(2) Seismic Behaviour of Steel Coupling Beams Linking Reinforced Concrete Shear Walls

Engineering Structures 27 (2005) 1024–1039

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Seismic behaviourof steel coupling beams linking reinforced concretshear walls

Wan-Shin Park, Hyun-Do Yun∗

School of Architecture, Chungnam National University, Daejeon 305-764, Republic of Korea

Received 3 November 2004; received in revised form 26 January 2005; accepted 15 February 2005Available online 9 April 2005

Abstract

Due to lack of information, current design methods to calculate embedment length for steel coupling beams linking concrete share tacit about cases in which the beams have connection details to the walls that include stud bolts and horizontal ties. In tanalyses were carried out to develop a model for calculating the embedment lengths of embedded steel sections. Five models for calculaembedment lengths in hybrid coupled walls are developed as variations of the Prestressed Concrete Institute guidelines for steel bracketattached to reinforced concrete columns. In addition, experimental studies on the steel coupling beam were carried out. Thevariables were the ratios of the coupling beam strength to the connection strength. Based on the test results, it is more advantageouthe coupling beams as shear-yielding members since a shear-criticalcoupling beam exhibits a more desirable mode of energy dissipathan a flexure-critical coupling beam.© 2005 Elsevier Ltd. All rights reserved.

Keywords: Steel coupling beams; Embedment lengths; Connection failure; Shear critical; Flexure critical

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1. Introduction

Properly designed hybrid coupled shear walls hamany desirable earthquake-resistant design features. Glateral stiffness and strength can be achieved. By couplingindividual walls, the lateral load resisting behaviour changto one in which overturning moments are resisted partialby an axial compression–tension couple across the wsystem ratherthan by the individual flexural action of thewalls. The beams that connect individual wall piers arereferred to as coupling beams. During major earthqualarge seismic forces are transferred between individual wallpiers through the coupling beams. In order for the hybcoupled shear walls to behave desirably the coupling beshould be stiff and strong and possess stable hysteresis.

Structural steel coupling beams provide a viable altnative to reinforced concrete coupling beams, particulawhere height restrictions do not permit the use of dee

∗ Corresponding author. Tel.: +82 42 821 6281; fax: +82 42 823 9467.E-mail address:[email protected] (H.-D. Yun).

0141-0296/$ - see front matter © 2005 Elsevier Ltd. All rights reserved.doi:10.1016/j.engstruct.2005.02.013

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reinforced concrete or composite coupling beams, or whthe required capacity and stiffness cannot be developeconomically by conventional reinforced concrete couplibeams. Previous researchers [1–3] have shown that the lat-eral stiffness and strength of concrete shear walls can benificantly increased by coupling the shear walls using ebedded steel beams. The critical design issues for a hycoupled shear walls are: (1) how to develop the modelcalculating the embedment lengths of steel coupling beataking into account the connection details; and (2) howdesign the steel coupling beams for the desired behaviouthe hybrid coupled shear wall system. No design methoare currently available for computing the required embement length of steel coupling beams, taking into accountcontribution of the auxiliary bars and the horizontal ties othe top and bottom flanges of an embedded steel section. Treliability of the model proposedin this study for computingthe required embedment length of steel coupling beamsevaluated. A flow chart of this research is shown inFig. 1.

This research develops a model for computing trequired embedment length of steel coupling beams, tak

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W.-S. Park, H.-D. Yun / Engineering Structures 27 (2005) 1024–1039 1025

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Notation

a distance from beam shear to face of wall(in mm)

A1 the area within the hysteresis loop of one halfcycle

A2 the area of the triangle defined by anequivalent elastic stiffness to the peak load andcorresponding deflection at each half cycle

be effective bearing width (in mm)bf beam flange width (in mm)Cb resultant concrete compressive force acting on

top and at the back of embedded steel section,(in N)

C f resultant concrete compressive force actingbelow and at the front of embedded steelsection, (in N)

d beam overall depth (in mm)e distance from face of wall to effective fixed

point of beam (in mm)fb bearing strength (in MPa)f ′c concrete compressive strength (in MPa)

f ′cc compressive strength of confined concrete

(in MPa)l distance from ram on beam to face of wall

(in mm)le embedment length (in mm)Lclear clear span of steel coupling beams (in mm)L effective clear span of steel coupling beams

(in mm)Mp plastic moment (in N mm)t f beam flange thickness (in mm)tw beam web thickness (in mm)twall thickness of wall (in mm)Vf shear corresponding to the moment capacities

(in N)Vn nominal shear capacities (in N)Vr connection strength (in N)Vu ultimate beam shear force (in N)β equivalent elastic damping coefficient(=

A1/2π A2)

into account the effects of the auxiliary bars and horizontties. In addition, it investigates the seismic behaviour of thsteel coupling beams in terms ofthe failure mechanism,hysteretic response, strength, stiffness, effective embedmlength, dissipated energy characteristics, and the equivaleelastic damping coefficient. The test results and discussionspresented in this paper provide background for desigguidelines of a hybrid coupled shear walls.

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2. Prototype structure

2.1. Design of walls and coupling beams

The test specimens represented a subassembly at37th floor of an imaginary prototype structure, a 50-storethree-bay by eight-bay office assumed to be locatedUniform Building Code seismic zone 4 [4]. The structureis comprised of five reinforced concrete walls linked tform the central core, flat slab system, and perimeter stframes. The typical floor plan is shown inFig. 2. Thecoupling beams are assumed to be structural steel membembedded in concrete door lintels over doors into thboundary element and interfacing them with the boundaelement vertical bars and hoops. The design moment ashear in the 37th floor coupling beam are the largest of thosof any floor. Hence, the interface between the steel couplibeams and reinforced concrete walls is more important atthis location. The prototype structure was designed for thecombined effects of gravityand earthquake load accordingto the UBC provisions for concrete structures [4]. The wallswere proportioned and detailed following the UBC anseismic provisions of ACI 318-02 [5]. The steel couplingbeams were designed in accordance with the seismic designrequirements for link beams in eccentrically braced framof the Canadian steel design standard [6].

2.2. Required embedment length

Additional auxiliary bars and horizontal ties attacheto the beam flanges in the embedment can contribute tothe capacity of the connection. No specific guidelines aprovided for computing the required embedment lengttaking into account the contribution of the auxiliary barand the horizontal ties on the top and bottom flangethe embedment regions, but previous studies have showthe adequacy of four models proposed by Marcakis anMitchell [7], Mattock and Gaafar [8], Kent and Park [9], andMinami [10], respectively. The required embedment lengthof the steel coupling beams in hybrid coupled shear wawas evaluated using five methods that are variations of thPrestressed Concrete Institute guidelines for the designsteel beams used as brackets [11], as shown inFig. 3. In allcases, the concrete strain was assumed to vary linearly, wthe maximum concrete strain taken as 0.003 at the facethe wall.

The first model (model A) was that proposed by Marcakand Mitchell [7]. Using a slightly different assumed stressdistribution shown in Fig. 4(a) and assuming a rigid-body motion of the embedded steel section, the followinexpression may be used for determining the requireembedment length.

Vu = 0.85 f ′cb′(le − c)

1 + 3.6e(le−c)

(N). (1)

1026 W.-S. Park, H.-D. Yun / Engineering Structures 27 (2005) 1024–1039

Fig. 1. Flow chart of this research.

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This model accounts for spreading of the compressforces within the wall. As a result, the width of the effectivecompression zoneb′ may be greater thanthe width of thebeamb (or the beam flangeb f ).

The second model (model B) was that proposedMattock and Gaafar [8]. Using the distribution of bearingstresses shown inFig. 4(b), the required embedment lengtle can be computed from:

Vu = 0.9 fbβ1b f le

(0.58− 0.22β1

0.88+ ale

)(N). (2)

The required embedment length to develop a verticshear at a distance from the face of the wall is computfrom this equation.

l

In the third method (model C), the concrete stredistribution was modified to reflect the larger strengand stiffness due to the confinement provided by laramounts of transverse reinforcement in typical boundaelements. The modified Kent and Park model [9] was usedto represent theconcrete stress–strain relationship, and tassumed stress distribution was changed accordingly, aillustrated inFig. 4(c). The effective bearing width was takeas b f (twall/bf )

0.66, similar to the derivation of the modelproposed by Mattock and Gaafar. A closed-form expressfor computing the required embedment lengthle could notbe derived. The necessaryle for developing the computedshear and moment of the coupling beam at the 37th floor wcalculated for a number of different levels of confinemeexpressed in terms ofK = f ′

cc/ f ′c.


2inbd

Fig. 2. Typical plan of prototypestructure (dimension: mm).

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(a) Actual stresses.

(b) Assumed stresses and strains.

Fig. 3. Stresses and strains of steel coupling beam–wall connection.

In model D, a uniform concrete bearing stress equalf ′c was assumed over theembedment length, as shownFig. 4(d). This value was proposed by Minami [10] for

earing stress in cases that are similar to the stressesevelop on the beam flanges. Using this assumption,

ate

following expression can be derived.

le = Vu

4be f ′c

±√

3V2u

16b2e f ′2

c+ aVu

be f ′c

(N). (3)

Using the effective bearing widthbe as for model C, therequired embedment length using model D becomes 161for specimen SBVRT, which issignificantly less than thevaluescalculated using models A, B, or C. This can battributed to the assumed size of the concrete bearing stabove and below the embedded steel section. The value bymodel D was deemed to be too small and model D wasfollowed for the final design.

The fifth model (model E) was that proposed in thstudy. Based on the observation of the test results fromprevious study [12], stud bolts and horizontal ties on thtop and bottom flange of an embedded steel coupling beasection were specified in an effort to improve the stiffnesand to improve the transfer of the flange-bearing force tosurrounding concrete. The contribution of the auxiliary barsand the horizontal ties are not considered in the previous fmodels. Model E was used to calculate embedment lengtaking into account the contribution of the auxiliary bars athe horizontal ties, as shown inFig. 4(e). Taking momentsabout the centre of action ofCb, the required embedmentlengthle can be computed from:

Vu(proposed) = fbβ1ble

(0.58− 0.22β1

0.88+ a/ le

)

+2(0.88− a/ le)

n∑i=1

Asi fsi

0.88+ a/ le(N), (4a)


(a) Model A (Marcakis and Mitchell 1980). (b) Model B (Mattock and Gaafar1982).

(c) Model C (Kent and Park 1982).

(d) Model D (Minami 1985). (e) Model E proposed by this study.

Fig. 4. Proposed models for calculating embedment length.

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fb = 4.5√

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(t

b

)0.60

(MPa). (4b)

The variation ofle in terms ofK is plotted inFig. 5. Therequired embedment clearly increases as the confinemenreduced. As shown inFig. 5(a), for the amount of transversereinforcement in the boundary element (specimen SBVRat the 37th floor, the value ofle computed by model Eproposed in this study is 396 mm, which is slightly morthan 369 mm, the value computed by model C(K =1.20). As shown inFig. 5(b), for shear-critical steel beams(specimen SCF), the required embedment lengths calculausing the models A, B, C(K = 1.0), D, and E are foundto be527 mm, 294 mm, 306 mm, 117 mm, and 286 mmrespectively. As shown inFig. 5(c), for flexure-critical steelbeams (specimen FCF), the required embedment lengcalculated using the models A, B, C, D, and E range betwe84% and 115% of those for shear-critical steel beams. Tembedment lengths were calculated based on the Modeproposed in this study that considered the contribution of tauxiliary bars and horizontal ties.

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3. Experimental program

Standard similitude concepts were followed to establithe size and dimensions of reinforcing bars for the tesspecimens. The overall wall and beam dimensionsspecimens are summarized inFig. 6. The testsubassembliesconsisted of one half of the length of the coupling beaat the 37th floor. The test variables and details used in tstudy are summarized inTables 1and2. The specimens werecast vertically, and typical construction joints in the waaround the connection were not reproduced. Ready-miconcrete with a specified 28-day compressive strength30.0 MPa was used for each of the three specimens.maximum size of concrete aggregate chosen was 15 mto ensure good compaction of the concrete in the tspecimens. The slump of the concrete was 145 mm. Feach batch, 100× 200 mm cylinders were constructed tmeasure the compressive strength of the concrete. Concstrength and elastic modulus were tested using the metdefined in the ASTM standards. The horizontal and verticreinforcement consisted of 13 mm diameter deformed barThe reinforcing steel used for all walls was obtained froa single batch of steel for each bar diameter, and thrspecimens were tested from each diameter of reinforcing


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(a) SBVRT specimen (Connection failure).

(b) SCF specimen (Shear critical).

(c) FCF specimen (Flexural critical).

Fig. 5. Comparisons of computed embedment length.

used. Tension tests were conducted on full-sized bar samin accordance with ASTM Standard A370 to determine tyield strength, ultimate strength, and total elongation. Theobserved material properties are reported inTables 3and4. A schematic diagram of the testapparatus is shown inFig. 7. The testspecimens were loaded with two servocontrolled actuators, a 1000 kN ram to apply load to the wa

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and a 2000 kN ram to load the steel coupling beam. Boactuators were controlled using a computer-based controlThe displacement of all specimens was controlled to follosimilar displacement histories with progressively increasingamplitude. The loading history for connection failure, shearand flexure-critical specimens are shown inFig. 8(a)–(c),respectively. The data were acquired from the load on tactuators, the deflection and rotation of the steel couplinbeams, the strain in the longitudinal reinforcing bars anstud bolts in the embedmentregion, and the strain onthe flanges and web of the steel coupling beams. A totof 36 data channels of instrumentation were used. Tinstruments monitored loads in the actuators, deflections arotation of the coupling beam, shear distortion of the stecoupling beam, strains on selected longitudinal reinforcinbars in the boundary element, strains in the flanges aweb of the coupling beam, overall wall curvature, and locwall distortion in the vicinity of the embedded beam. Theexperimental results are used to evaluate the performanof structural steel coupling beams compared with other tedata.

4. Experimental results

4.1. Damage and crack pattern

Fig. 9 shows the failure modes for specimens SBVRT,SCF, and FCF. In specimen SBVRT, initial vertical crackinat the steel coupling beam flange–concrete interface belthe embedded bottom flange was observed at a rotationangle of about 0.017 rad. Inclined cracks located at thflange–concrete interface extended from the flange acrothe inner face of the wall to the side faces of the wall atrotational angle of about 0.019rad. Localized spalling andcrushing of the concrete along the top and bottom flangesthe coupling beam at the front of the compression zone wasinitially observed at a rotational angle of 0.043 rad, as showin Fig. 9(a). In specimen SCF, severe web buckling in thclear span of the steel coupling beam led to its final rupturas shown inFig. 9(b). Specimen FCF, having a clear spaof 1200 mm, was designed and detailed as a flexure-criticcoupling beam. The response of specimen FCF was notaless stiff than that of specimen SCF. In addition, the stebeam remained elastic in shear throughout the test. Unlishear yielding, which occurs uniformly over the entire lengtof a coupling beam, the flexural hinge propagates away frothe region of critical moment, as shown inFig. 9(c). Fromthe observed failure modes, shear critical failure was moreasonable for rehabilitation or retrofitting when considerinthe degree of building damage.

4.2. Hysteresis response

The hysteretic responses of specimens SBVRT, SCF, aFCF are presented inFig. 10. Specimen SBVRT did notexhibit stable spindle-type hysteretic loops due to prematu


Fig. 6. Details of steel coupling beams (unit: mm).

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embedment region failure before the web and flange ofsteel coupling beam yielded. As shown inFig. 10(a), thehysteretic behaviour of specimen SBVRT shows a moderlevel of pinching, which is attributed to bearing failure ithe beam–wall connection region. In addition, specimenSBVRT showed a sudden decrease in strength betweenfirst and second cycles at the rotational angle of 0.043 rad.

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e

However, specimen SCF exhibited very large, stable loothroughout the test with little strength or stiffness decayevident, as shown inFig. 10(b). Compared with specimenSCF, specimen FCF exhibited a more unstable and pincheresponse, as shown inFig. 10(c). This isattributed to localconcrete bearing failure by buckling of the compressiflange, near the shear wall face.

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.-D.Y

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Table 1Test variables

Researcher Detail l e l/H f ′c fh fv fyw fy f Loading Predicted

(mm) – (MPa) (MPa) (MPa) (MPa) (MPa) history failure mode

This study300 3.43 30.0 398 398 219.5 225.4 C CF300 3.43 30.0 398 398 219.5 225.4 C SCF300 3.43 30.0 398 398 219.5 225.4 C FCF

Shahrooz [13]864 1.17 35.0 475 475 234 234 C FCF762 1.17 35.0 475 475 234 234 C FCF508 1.17 35.0 475 475 234 234 C FCF

Harries [14]

600 3.46 25.9 458 410 320 372 C CF600 3.46 43.1 458 410 309/276 295 C SCF500 1.29 32.9 447 437 403 378 C SCF600 3.46 35.0 447 437 403 378 C FCF

CF: Connection failure; FCF: Fle

Specimen H B tw t f lname (mm) (mm) (mm) (mm) (mm)

SBVRF 350 175 7 11 800SCF 244 175 7 11 600FCF 244 175 7 11 1200

Wall 1 457 203 25 25 534Wall 2 457 203 25 25 534Wall 3 457 203 25 25 534

S1 347 135 5 19 1200S2 347 135 8 19 1200S3 349 127 6 8 450S4 349 127 8 8 1200

xural critical failure; SCF: Shear critical failure.


Table 2Details of test specimens

Specimens Item

Stud bolts Horizontal tiesWall reinforcements Eccentricity of vertical load

RemarkIn wall In connections e (mm)

SBVRT 12-φ19 4-HD10 HD13@230 HD13@230 +150 Connection failurel/(Mn/Vn) = 1.8

SCF 12-φ19 4-HD10 HD13@230 HD13@230 +150 Shear criticall/(Mn/Vn) = 1.4

FCF 12-φ19 4-HD10 HD13@230 HD19@100 +150 Flexure criticall/(Mn/Vn) = 2.8

Table 3Average concrete compressive strengths

Compressive strength Ultimate strain Slump Elastic modulus Poisson’s(MPa) (µ) (mm) (GPa) ratio

30.0 2340 150 26.2 0.11

At the time of testing.

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The relationship between normalized measured load androtational angle is shown inFig. 11. Specimens SBVRT,SCF, and FCF developed maximum capacities equal402.1, 283.1, and 211.8 kN, respectively at the ultimate loin the compression cycles (beam pushed down).Fig. 11(a)shows values ofVn (test)/Vn (anal.) for specimens SBVRT,SCF, and FCF of1.07, 1.27, and 1.06, respectively. Threserved strength of specimen SCF is 17% and 20% larthan that of SBVRT and FCF, respectively. This can battributed to the full shear yielding occurring in the cleaspan of the steel coupling beam without significant distrein the embedded region. In particular, specimen FCF did nodevelop a substantially larger strength than the theoreticavalue because of premature lateral buckling. Compared withthe test results of Shahrooz [13] and Harries [14], rotationalangles greater than those corresponding to ultimate loadfor the shear critical steel coupling beams showed mostable response, without significant strength degradatithan connection failure and flexure critical beams, as shownin Fig. 11(b).

4.3. Stiffness characteristics

Fig. 12 shows thedegradation of peak-to-peak stiffnesplotted against rotation angle. The initial peak-to-peastiffness of specimens SBVRT, SCF, and FCF rangebetween 19.81 and 49.82 kN/mm, but at a rotation angleof 0.063 rad, the stiffness values of all specimens wesimilar, with values within the range 3.51–6.29 kN/mm, asshown in Fig. 12(a). For the rotation angle of 0.014 radthe stiffness degradation of specimens SBVRT, SCF, aFCF were 74.7%, 93.9%, and 94.3% of the initial stiffnesrespectively. The overall stiffness degradation was alsmaller for specimen FCF compared to specimens SBVRT

r

,

d

and SCF. Stiffness decay was more evident in specimeSBVRT. Compared with the test result of Shahrooz [13]and Harries [14], the rate of stiffness degradation of shearcritical steel coupling beams, having a shorter clear spathan flexure critical beam, is larger than that for connectionfailure and flexure critical beams. However, the rate ofstiffness degradation for shear critical steel coupling beamsis smaller than that of connection failure and flexure criticabeams as the rotation angle increases, as shown inFig. 12(b).

4.4. Effective fixed point

Accounting for flexural and shearing deformations, andassuming that the steel coupling beam is fixed at the faof wall, the initial stiffness values of specimens SBVRT,SCF, and FCF are 154.30 kN/mm, 83.22 kN/mm, and52.54 kN/mm, respectively. The measured initial stiffnessvalues range between 32.3% and 38.5% of the theoreticallcalculated values. Hence, the steel coupling beams wereclearly not fixed at the face of the walls. The apparent lacof fixity may also be seen from the deflection profiles inFig. 13(a), in which profiles corresponding to a rotationaangle at yield load are shown. It should be noted that as thyield rotational angles are different for each specimen, thetip displacements are not identical when a certain rotationangle is reached. For specimen SBVRT at point D, whicwas approximately 5 mm outside the face of the wall, amaximum deflection of 1.68 mm can be seen fromFig. 13(a)(beam push down). This amount of deflection, which isapproximately two-fifths of the corresponding value at thtip, is substantial for a point that is typically assumed tohave negligible deformation. For specimens SCF and FCadditional strain gauges were installed to detect the beamovement along the embedment length (inside the wall


Table 4Mechanical properties of steel

Type ItemYield strength Yield strain Elastic modulus Ultimate strengthf y (MPa) εy (×10−6) Es (GPa) fsu (MPa)

Reinforcement10 mm diameter deformed bar 398 2325 171.2 56613 mm diameter deformed bar 400 2533 157.9 555

SteelBeam flange 339 1682 201.2 461Beam web 352 1827 192.7 489

Stud bolts 19 mm diameter deformed bar 362 1701 215.8 449

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Fig. 7. Test setup.

At the yield rotational angle, beam vertical deflections agreat as 0.002 mm were recorded at a point 50 mm insithe wall from the face (point C). Movement at point A,almost 300 mm from the face of the wall, was also recordeafter the beams yielded.Fig. 13(a) shows that the steelcoupling beam clearly displaced vertically by as much a0.002 mm at point C. Based on three test results, the effectfixed point of steel coupling beams with auxiliary bars isinside the wall at about 1/6–1/5 of the embedment length.These values are substantially smaller than those(1/4–1/3)

found in Shahrooz [13] without auxiliary bars, as shownin Fig. 13(b). This can be attributed to the increased fixitydegree of steel coupling beam at the face of the shear wdueto the presence of the auxiliary bars.

4.5. Energy absorption and dissipation

A graph of the cumulative dissipated energy is plottedin Fig. 14(a). The real energy absorptions of specimenSBVRT and FCF were virtually identical until a rotationalangle of 0.052 rad was reached. Specimens SBVRT and Fwere able to dissipate about 8.5 and 10 times their yieenergy through a rotational angle of 0.060 rad, respectiveThe change in failure mode from flexural to shear hingin

(a) Load history 1 (Specimen SBVRT).

(b) Load history 2 (Specimen SCF).

(c) Load history 3 (Specimen FCF).

Fig. 8. Displacement history.

resulting from the improvements made to specimen Sresults in a significant increase in energy absorption capato about 28 times the specimen’s yielding energy. Atrotational angle of 0.051, 0.052, and 0.051 rad, cumulatenergy values were equal to 33.98, 226.9, and 37.4 kJspecimens SBVRT, SCF, and FCF, respectively.

Fig. 15 shows the equivalent elastic damping coefficienβ for specimens SBVRT, SCF, and FCF, and for thotested by Shiu [15]. A larger damping coefficient reflects agreater ability to dissipate energy. The maximum valueβ, representing elasto-plastic hysteretic behaviour, is ab2/2π = 31.8%. The short shear span of specimen SCF ashowed greater energy absorption ability than specimSBVRT and FCF, as shown inFig. 15(a). Two specimens,tested by Shiu [15], were chosen as representative


d5

esg

(a) Specimen SBVRT.

(b) Specimen SCF.

(c) Specimen FCF.

Fig. 9. Cracking pattern.

the response of conventionally and diagonally reinforceconcrete coupling beams with a span-to-depth ratio of 2.Compared with the test result of Shiu [15], specimen SCFexhibited greater energy absorption capability than either thdiagonally or conventionally reinforced concrete specimenThe response of diagonally reinforced concrete couplin

.

.

(a) Specimen SBVRT.

(b) Specimen SCF.

(c) Specimen FCF.

Fig. 10. Load–rotational angle hysteretic loops.


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d

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ion)

(a) Specimens SBVRT, SCF, and FCF.

(b) Comparisons of test results and other data.

Fig. 11. Strength characteristics.

beams conducted by Paulay [16] with smaller span-to-depthratios showed little improvement in hysteretic dampinover those withlarger ratios. The damping coefficients oPaulay’sbeams, whose span-to-depth ratios were 1.29 a1.02, are only marginally higher than Shiu’s(l/d = 2.5)

beams shown inFig. 15(a). Specimen SCF, with a spanto-depth ratio of 2.3 exhibited significantly greater energyabsorption ability than the comparable diagonally reinforcespecimens.

Fig. 15(b) shows the equivalent elastic damping coefcientβ of all specimens, that of Harries [11], and that of thesteel shear link with a span-to-depth ratio of 3.7 testedMalley and Popov [17]. It is clear that interms of rotationangle and energy absorption, the response of specimensVRT and FCF is poorer than the response of the steel shlink. However, the response of specimen SCF exceedsresponse of the shear link at a rotation angle of 0.041 rad,though it has a shorter shear link, giving it a significantly beter response than the one tested by Popov [17]. However, theresponse of specimen S2 of Harries [14] closely approaches

-



Fig. 12. Stiffness characteristics.

the response of the steel shear link. Based on theresults, if steel coupling beams are designed and detaileyield in shear, they are able to dissipate greater amountenergy and are more ductile than flexure critical steel copling beams, conventionally and diagonally reinforced cocrete coupling beams.

4.6. Strains and stresses

(1) Concrete strain inthe connection regionFig. 16 shows the compressive strain in the connecti

region for specimens SBVRT, SCF, and FCF. Three gawere located below the embedded steel coupling beamaddition, two gages were located within the beam flanwidth and one gage outside the beam width. In all casthe maximum concrete compressive strain occurred bethe center of the embedded steel coupling beam. Arotational angle corresponding to the ultimate load, theconcrete strain below the outer edge (point B) of the sectvaried from about 29% of the centerline strain (point C


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(a) Displacement profile.

(b) Model of effective fixed point.

Fig. 13. Beam displacement profile (outside and inside wall).

for specimen SBVRT to about 20% of the centerline strafor specimen SCF, as shown inFig. 16(a) and (b). Thisdifference must have resulted from the behaviour of tsteel coupling beams. However, as the load increased furththe strain below the outer edge of the embedded sectionincreased more rapidly for specimen SBVRT, as mayseen inFig. 16(a). In the case of specimen SCF, the concrecompressive strains at ultimate are estimated to vary froabout 2469 micro strains below the center to about 3micro strains below the outer edge of the bottom flangas shown inFig. 16(b). This increase in strain below thecenter of the flange relative to that below the edge of tflange is probably due to diagonal strutting action occurriin the concrete between the upper face of the bottom flanand the face of the web of embedded steel coupling beaSince specimens FCF had FBPs to mobilize the concrstrut between the flanges, the inner concrete gage (pointat ultimate picked up compressive strain at almost four timthe rate of theouter gage (point D), as shown inFig. 16(c).

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,

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Fig. 14. Cumulative dissipated energy.

(2) Embedded steel web stressesFig. 17 shows the distribution of stresses in the steel w

of specimens SBVRT, SCF, and FCF. The strains in the sweb were measured at three different locations by meanrosette strain gauges in order to determine the distributionstresses. It can be seen that the distribution of stressesnot follow a symmetrical pattern, with the higher stresserecorded in the front half of the embedment length. Itnoted that, in specimen SBVRT, significant deteriorationof the connection started to take place although yieldiof the steel web had not occurred. That deterioration wcharacterized by a rapid increase in the width of diagocracks and spalling of the concrete. In specimen SCF,shear stresses near the back face (point A) of the connectionwere approximately 15% lower than those measured at tfront face (point C), as shown inFig. 17(b). Comparedwith specimen FCF, the web strains of specimen SCF wsensitive to the changes in the failure mode, as shown inFig. 17(b) and (c). Based on the measured strains, shyielding of the embedded steel web in all specimens didoccur during the test. This canbe attributed to the fact that


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(a) Comparisons of test results and RC coupling beams.

(b) Comparisons of test result and steel coupling beams.

Fig. 15. Equivalent elastic damping coefficient(β).

concrete compressive struts due to the face bearing plhelped to reduce the participation of the embedded steelw

(3) Stud bolts stressesStrain gage readings for stud bolts adjacent to

connection are shown inFig. 18. In a steel couplingbeam–wall connection, because of the change in lothrough the connection, high bond stresses are imposedthe vertical stud bolts passing through the connection. Ifslippage takes place in these bars, the longitudinal strashould change from tension at the top (or bottom) of tconnection to compression at the bottom (or top). Whthe connection was loaded as shown inFig. 18, the uppergage recorded tension and the lower recorded compressIn specimen SBVRT, the measured stresses of stud bbelow the bottom flange were slightly less than those abovethe top flange, probably because of tension in the concand an unequal stress distribution of force between allstudbolts, as shown inFig. 18(a). It is worth emphasizingthat the measured stud bolt strains indicate that slippage overtical stud bolts below the bottom flange can occur af

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(a) Specimen SBVRT.

(b) Specimen SCF.

(c) Specimen FCF.

Fig. 16. Strains of concrete.


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(a) Specimen SBVRT.

(b) Specimen SCF.

(c) Specimen FCF.

Fig. 17. Stresses in embedded steel web.

a very few cycles and before the yield strength of the stubolts is reached. Similar behaviour has been reportedcolumn and beam bars in composite joints [18] due to a lossof bond transfer along vertical bars through the connectiregion. In specimens SCF and FCF, the measured streof stud bolts below the bottom flange showed similar valu

r

es

(a) Specimen SBVRT.

(b) Specimen SCF.

(c) Specimen FCF.

Fig. 18. Stresses in stud bolts.

in comparison to those above the topflange, as shown inFig. 18(b) and (c).


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5. Conclusions

The following conclusions were derived from the resultof the analysis and experiments in this study on the stecoupling beams in a hybrid wall system:

1. Existing models for calculating the embedment lengthdo not adequately consider the effect of connection detaiModel E proposed in this study can be reliably used tcompute the required embedment length of steel couplinbeams considering the effects of connection detailsauxiliary bars and horizontal ties in a hybrid coupled sheawall system.

2. In specimen SBVRT, localized spalling and crushingof the concrete along the top and bottom flanges of thcoupling beam, at the front of the compression zone, waobserved at failure. In specimen SCF, severe web buckliin the clear span of the steel coupling beam led to its finrupture. Specimen FCF remained elastic in shear throughothe test, and failure occurred uniformly over the entire lengtof the coupling beam, with the flexural hinge propagatinaway from the region of critical moment with increasingrotational angles.

3. Specimen SBVRT did not exhibit stable spindle-typhysteretic loops due to premature embedment region failubefore the web and flange of the steel coupling beayielded. However, specimens SCF and FCF exhibited molarge, stable loops throughoutthe test with little strength orstiffness decay evident.

4. Accounting for flexural and shearing deformations, anassuming that the steel coupling beam is fixed at the facewall, the initial stiffness values computed for specimens SBVRT, SCF, and FCF were significantly larger than the measured initial stiffness. The effective point of fixity of the steecoupling beam with the auxiliary bars and horizontal ties iinside the wall at about 1/5 to 1/6 of the embedment length.

5. Specimen SCF showed greater energy absorptiability than specimens SBVRT, FCF, and diagonally and thconventionally reinforced concrete coupling beams tested byothers. Therefore, in order for the desired behaviour of thhybrid coupled shear wall system to be attained, it is moadvantageous to design the coupling beams as shear-yieldmembers since a shear-critical coupling beam exhibits amore desirable mode of energy dissipation than a flexurcritical coupling beam.

Acknowledgements

The authors are thankful to the Ministry of Construction

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and Transportation (Project No. C104A1020001-04A02000210) for sponsoring this project.

References

[1] El-Tawil S, Kuenzli C. Pushover of hybrid coupled walls. II: Analysiand behaviour. Journal of the Structural Division ASCE 2002;128(10):1282–9.

[2] Harries KA. Ductility and deformability of coupling beams inreinforced concrete shear walls. Earthquake Spectra 2001;17457–78.

[3] Gong B, Harries KA, ShahroozBM. Behaviour and design ofreinforced concrete, steel, and steel–concrete coupling beamsEarthquake Spectra 2000;16(4):775–99.

[4] Uniform Building Code, Volume 2. Structural engineering desigprovisions. Whittier (CA): International Conference of BuildingOfficials; 1994.

[5] ACI Committee 318. Building code requirements for reinforceconcrete (ACI 318–02) and commentary (ACI 318R–02). FarmingtHills (MI): American Concrete Institute; 2002.

[6] Design of Concrete Structures (CSA A 23.3–94). Rexdale (CanadCanadian Standards Association; 1994.

[7] Marcakis K, Mitchell D. Precast concrete connections with embeddsteel members. Journal Prestressed Concrete Institute 1980;2586–116.

[8] Mattock AH, Gaafar GH. Strength of embedded steel sectionsbrackets. ACI Journal 1982;79(9):83–93.

[9] Park R, Priestly MJN, Grill WD. Ductility of square confined concretcolumns. Journal of the Structural Division ASCE 1982;108(4929–50.

[10] Minami K. Beam to column stresstransfer in composite structurescomposite and mixed construction. In: Roeder C, editor. New Yo(NY): ASCE; 1985.

[11] Manual on design of connections for precast prestressed concChicago (IL): Prestressed Concrete Institute; 2004.

[12] Yun HD, Park WS, Han BC, Hwang SK. Shear strength of steecoupling beam connections embedded in reinforced concrete shearwalls. Journal of the Architectural Institute of Korea 2004;20(543–50.

[13] Shahrooz BM, Remmetter MA, Quin F. Seismic design andperformance of composite coupled walls. Journal of the StructuDivision ASCE 1993;19(11):3291–309.

[14] Harries KA. Seismic design and retrofit of coupled walls usinstructural steel. Ph.D. thesis. Montreal: McGill University; 1995,p. 229.

[15] Shiu KN, Barney GB, Fiorato AE, Corley WG. Reversed load tests oreinforced concrete coupling beams. In: Central American conferencon earthquake engineering. 1978, p. 239–49.

[16] Paulay T. Coupling beams of reinforced concrete shear walls. Jouof the Structural Division ASCE 1971;97(ST3):843–62.

[17] Malley JO, Popov EP. Designconsiderations for shear linksin eccentrically braced frames. Report No. UCB/EERC–83/2Berkeley: Earthquake Engineering Research Center; 1983, p. 62.

[18] ASCE Task Committee on Design Criteria for Composite Structuin Steel and Concrete. Guidelines for design of joints between sbeams and reinforced concrete columns. Journal of the StructEngineering ASCE 1994;120(8):2330–57.

Documents

(2) Seismic Behaviour of Steel Coupling Beams Linking Reinforced Concrete Shear Walls