Jetty design at Rio Ara rang uá

JetMtty deMSc. Proj

esignect, Hydr

WimIlsRo

Prof A.H. daLindino BIr. H.J.

Ir. W.J.M

at Riraulic Eng

Itajaí, BrazilOctober 2009

m van der Woe van den Bosoderik HeeremMaarten Ruij

Supervisors:

a Fontoura KleBenedet Oc. M Verhagen (TU. Peperkamp

Sponsors:

io Aragineering

9

oerdt sch ma s

: ein (UNIVALI)M.Sc (CPE) U Delft) (TU Delft)

arangg, TU Delf

guá ft

General notice to the reader: In the academic programme for Hydraulic Engineering we have in the 4th year (i.e. in the first year of the Master Programme) the requirement that students should do in a group of four to six persons a so-called "groupwork". It is also called "Master Project". During this groupwork they should make a full design of something. The work should be integral, starting with terms of reference, and ending with the real design. This can be a structure, but also it can be a harbour lay-out, a policy plan design, etc. The total time available for the project is in the order of two months and will provide 10 European Credits. It has to be practical and applied. It is certainly not an M.Sc. thesis assignment (the thesis work is individual, 6 months and more focussed on research or advanced design work on details). But it is also not an appren-ticeship, internship or traineeship where the student has to work together with a group of experienced people. For this groupwork they have to solve the problem on their own (of course with guidance). This report is the result of such a Master Project. This report has been assessed by staff of TU Delft. It has been provided with a passing mark (i.e. a mark between 6 and 10 on a scale of 10), and consequently considered sufficient for publication. However, this work has not been fully corrected by TU Delft staff and therefore should be considered as a product made in the framework of education, and not as a consultancy report made by TU Delft. The opinions presented in this report are neither the opinions of TU Delft, neither of the other sponsoring organisations. Department of Hydraulic Engineering Delft University of Technology

Jetty design at Araranguá river Final Report

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Preface In this report a design is made of a jetty in the river mouth of Rio Ararangua, Santa Catharina, Brazil. This research project was done at Universidade do Vale do Itajai (UNIVALI) from September to October 2009 and was commissioned by Coastal Planning and Engineering (CPE). The report is written by four students from the Delft University of Technology (TU Delft) studying at the faculty of Civil Engineering & Geosciences. Three of the students working on this project study the master variant “Hydraulic Engineering” and the fourth studies “Building engineering”. The project is done for elective course “CT4061 Multidisciplinary project”, which can be chosen to in the master program of Civil Engineering at the TU Delft. The accent of the course “Multidisciplinary project” lays at research and design in collaboration with students with other specialties. The fact that this project has been done in Brazil creates new challenges like communication difficulties, different circumstances and the use of different methods. The subject “design of a jetty” is derived in collaboration with ir. H.J. Verhagen (TU Delft), Prof. A.H. da Fontoura Klein (UNIVALI) and Lindino Benedet Oc. M.Sc. (CPE). During the project Lindino Benedet acted as the initiator, A.H. da F. Klein and ir. H.J. Verhagen both supervised our progress. The design derived from this project can be regarded as a conceptual design which can be used as a starting base. We would like to thank everybody that helped us during this project. Lindino Benedet, prof. Klein, ir. Verhagen and ir. W.J.M. Peperkamp for the supervision. Prof. Maria Elizabeth da Costa Gama for her guidance through UNIVALI and arranging our stay. All the students at the “Laboratorio Oceanographia Fisica” who helped us, especially João Dobrochinsky for his specific knowledge on the project. And off course our sponsors, who not only made this project possible, but also for their specific knowledge that was shared by e‐mail and even by a visit in Brazil. Itajai, Brazil 22 October 2009 Ilse van den Bosch Wim van der Woerdt Roderik Heerema Maarten Ruijs


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Abstract The hydraulic efficiency of the river Araranguá will be increased by the construction of jetties at the mouth of the river. The river tends to migrate from south to north which decreases the hydraulic efficiency and causes flooding upstream during events of high discharge. To stop the migration the construction of jetties at the mouth of river Araranguá is proposed. A design for a jetty system is available (INHP, 1993), but should be updated. Due to the construction of a jetty system longshore sediment transport will be interrupted and this will cause morphological changes. At Araranguá river a large and highly varying longshore sediment transport is present with a high energy wave climate and micro tidal range. These conditions cause increased morphological changes when constructing a jetty system. The objective of this research is to design a jetty system that prevents flooding upstream and induces as less morphological changes as possible. The jetty should be designed at the specified location, using available datasets from previous research and according to modern design standards. Also navigation through the channel should be possible. Using a Systems Engineering (SE) approach all requirements, design options and design steps have been identified. Furthermore a functional analysis was conducted to lists the required performance criteria of the system. Calculations and modeling of hydraulic, meteorological and sediment transport conditions form the basis of the boundary conditions used as an input for the conceptual design. Many assumptions needed for design choices are also based on availability of materials and knowledge and costs. In a trade off these assumptions and the boundary conditions defined the jetty type and –material and the sediment bypass system type. The conceptual design consists of a functional design in which channel cross section, jetty length and jetty orientation are defined. Also the structural design is done where detailed cross sectional dimensions are calculated, where slope angle, crest height, crest width and dimensions of the armourstone are most important. The channel width and depth are determined by navigational and morphological purposes and are 120 m and 6 m respectively. A weir jetty, which has been implemented in the SW arm of the design, was considered the best sediment bypass system after a multi criteria analysis. Because of the placing of the deposition basin waves could penetrate much further in the jetty channel and lengthening the NE arm to 630 m and a depth of 6,6 m was necessary. The length of the SW arm of the jetty from the previous is kept at 650 meter to a depth of 7,2 meter. The final result of the structural design is a statically stable rubble mound breakwater with granite as construction material. Rock has been applied in the range of 0,4 m (170 kg) to 1,0 m (2,7 tons) in diameter, with a core of quarry run material. In sections where the required rock mass exceeded the threshold value for economical construction (2,7 t), tetrapods have been applied to stabilize the slope. Tetrapods have been applied of 6,6 and 9,7 tons. The total use of material is about 42.000 tons of tetrapods and 490.000 tons of granite rocks. Overtopping requirements determine the crest freeboard and 0,3 meter is added for post construction settlement. The crest width depends on construction considerations. All other dimensions are directly related to the stone size and weight, which are dependent on the design wave heights. Due to the short time span, uncertainty of data and lack of knowledge and experience, important recommendations are made for future final design purposes. First of all, 3D and physical modelling of wave impact improves reliability of the design. Also several data sets, like soil conditions, have not been available and should be acquired. Finally, costs were not calculated, but assumptions were made based on reasoning. Cost calculation should be used in many decisions involved in jetty design and.


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Furthermore, it is possible to reduce the cost of the construction, by reducing the amount of material needed. Since the construction of the jetties at river Araranguá is subject to a very limited budget, a cost reduction is an important issue for further research on this jetty design. Safety margins can be reconsidered, which may lead to a lower volume of rock required. Also only navigation of smaller boats can be considered, which leads to a shallower channel and therefore the jetties have to be extended less far into the ocean. In the same fashion, overtopping requirements can be reconsiderd, which leads to a lower construction height and less material.


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Index JETTY DESIGN AT RIO ARARANGUÁ ....................................................................................................... 0

PREFACE ........................................................................................................................................................ 1

ABSTRACT ..................................................................................................................................................... 2

INDEX ............................................................................................................................................................ 4

LIST OF FIGURES ............................................................................................................................................ 8

LIST OF TABLES ............................................................................................................................................ 11

LIST OF SYMBOLS ........................................................................................................................................ 12

Roman symbols ............................................................................................................................................. 12 Greek symbols ............................................................................................................................................... 14

1. INTRODUCTION .................................................................................................................................... 16

2. PROJECT DESCRIPTION ......................................................................................................................... 17

2.1 AREA AND LOCATION .................................................................................................................................. 17 2.2 PROBLEM DESCRIPTION ............................................................................................................................... 18 2.3 OBJECTIVE ................................................................................................................................................ 19 2.4 PROJECT APPROACH .................................................................................................................................... 19 2.5 REQUIREMENTS ......................................................................................................................................... 20

3. FUNCTIONAL ANALYSIS ........................................................................................................................ 22

3.1 THE JETTY MUST PREVENT FLOODING UPSTREAM THE RIVER ................................................................................ 22 Cross‐sectional area channel ......................................................................................................................... 22 Length ........................................................................................................................................................... 22 Structure stability .......................................................................................................................................... 22

3.2 THE JETTY MUST HAVE A SEDIMENT BYPASS SYSTEM TO ALLOW FOR LONGSHORE SEDIMENT TRANSPORT ...................... 23 3.3 VESSELS MUST BE ABLE TO PASS THROUGH THE JETTY ......................................................................................... 23 3.4 THE JETTY MUST BE DESIGNED FOR A LIFETIME CYCLE OF 50 YEARS ....................................................................... 23 3.5 THE JETTY DESIGN MUST ALLOW FOR PEDESTRIANS TO WALK ON THE JETTY STRUCTURE ............................................ 24 Overtopping criteria ...................................................................................................................................... 24

3.6 THE JETTY MUST BE DESIGNED AT THE SPECIFIED LOCATION IN THE MOUTH OF THE ARARANGUA RIVER ........................ 24 3.7 THE CONSTRUCTION MATERIALS ARE LIMITED TO GRANITE BLOCKS AND PREFAB CONCRETE ELEMENTS ......................... 25 3.8 THE JETTY DESIGN MUST COMPLY WITH GLOBAL EN LOCAL REGULATIONS ............................................................... 26 3.9 THE JETTY MUST BE DESIGNED USING OF‐THE‐SHELF TECHNOLOGY ....................................................................... 26 3.10 THE JETTY MUST BE DESIGNED USING AVAILABLE DATASETS ................................................................................. 26

4. BOUNDARY CONDITIONS ..................................................................................................................... 27

4.1 HYDRAULIC AND METEOROLOGICAL BOUNDARY CONDITIONS .............................................................................. 27 Water levels .................................................................................................................................................. 27 River conditions ............................................................................................................................................. 29 Wind conditions ............................................................................................................................................ 29 Currents ......................................................................................................................................................... 30 Longshore sediment transport ...................................................................................................................... 30

4.2 WAVE CONDITIONS ..................................................................................................................................... 33 Wave data ..................................................................................................................................................... 33 Design life time and probability of damage .................................................................................................. 33 Estimation of design wave height ................................................................................................................. 34

4.3 WAVE PROPAGATION .................................................................................................................................. 36 Results ........................................................................................................................................................... 36

4.4 GEOTECHNICAL DATA .................................................................................................................................. 38 Soil bearing capacity ..................................................................................................................................... 38 Bathymetry .................................................................................................................................................... 38

4.5 OTHER BOUNDARY CONDITIONS .................................................................................................................... 40


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Material ......................................................................................................................................................... 40 Equipment ..................................................................................................................................................... 40 Labour ........................................................................................................................................................... 40 Political .......................................................................................................................................................... 40 Environmental considerations ....................................................................................................................... 41

5. SEDIMENT BYPASSING METHODS AND TECHNIQUES ............................................................................ 42

Natural Bypassing ......................................................................................................................................... 42 Fixed systems ................................................................................................................................................ 43 Mobile systems .............................................................................................................................................. 43 Semi‐mobile systems ..................................................................................................................................... 43 Equipment for sediment extraction ............................................................................................................... 43 Equipment for discharge ............................................................................................................................... 43 Weir Jetties .................................................................................................................................................... 44 Spur Jetties .................................................................................................................................................... 47 Sediment traps .............................................................................................................................................. 48 Channel wideners .......................................................................................................................................... 48 Flood shoals ................................................................................................................................................... 48 Ebb shoals ..................................................................................................................................................... 48

6. TRADE OFF ........................................................................................................................................... 49

6.1 SEDIMENT BYPASS SYSTEM TRADE OFF ........................................................................................................... 49 Design requirements ..................................................................................................................................... 49 Qualitative trade‐off for the choice of bypass system................................................................................... 50

6.2 BREAKWATER TYPE SELECTION ...................................................................................................................... 54 Description of main types .............................................................................................................................. 54 Selection type of breakwater ........................................................................................................................ 55 Description of mound subtypes ..................................................................................................................... 55 Selection of mound subtype .......................................................................................................................... 57

7. FUNCTIONAL DESIGN ........................................................................................................................... 59

7.1 PREVIOUS DESIGN ...................................................................................................................................... 59 7.2 JETTY LAY OUT FOR SHIPPING CONDITIONS ....................................................................................................... 60 Vessel characteristics .................................................................................................................................... 60 Channel width ............................................................................................................................................... 60 Channel depth ............................................................................................................................................... 61 Jetty length .................................................................................................................................................... 62 Jetty orientation ............................................................................................................................................ 62

7.3 MORPHOLOGICAL APPROACH ON CHANNEL DESIGN ........................................................................................... 64 Channel cross‐sectional area versus tidal prism............................................................................................ 65 River equilibrium states ................................................................................................................................. 66 Conclusion on morphological approach ........................................................................................................ 70

7.4 JETT LAY OUT FOR SAND BYPASS SYSTEM ......................................................................................................... 70 Design of the weir section ............................................................................................................................. 71 Design of the deposition basin ...................................................................................................................... 73 Sediment transport response to the design .................................................................................................. 74 Dredging operation cycles ............................................................................................................................. 75 Dredging equipment ..................................................................................................................................... 78 Dumpsite location ......................................................................................................................................... 79 Additional modification options .................................................................................................................... 79

7.5 SUMMARY AND CONCLUSION ON JETTY LAY OUT ............................................................................................... 80 Channel width and depth .............................................................................................................................. 80 Jetty length .................................................................................................................................................... 80 Jetty orientation ............................................................................................................................................ 81

8. STRUCTURAL DESIGN ........................................................................................................................... 82

8.1 CROSS SECTIONAL DESIGN ............................................................................................................................ 82


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Cross sectional dimensions ............................................................................................................................ 82 Mass of the armourstone .............................................................................................................................. 84 Toe structure ................................................................................................................................................. 87 Bedding layer ................................................................................................................................................ 88 Roundhead design ......................................................................................................................................... 89 Concrete Units ............................................................................................................................................... 90 Breakage ....................................................................................................................................................... 92 Weir section .................................................................................................................................................. 92 Transitions ..................................................................................................................................................... 93 Calculations and results ................................................................................................................................ 94

8.2 OVERTOPPING CALCULATION ...................................................................................................................... 100 Empirical method ........................................................................................................................................ 100 Neural Network (CLASH) ............................................................................................................................. 101 Remarks on overtopping calculations ......................................................................................................... 103

8.3 BEARING CAPACITY AND SETTLEMENT ........................................................................................................... 103 Bearing capacity calculations ...................................................................................................................... 103 Calculation of settlement ............................................................................................................................ 104

8.4 ROCK QUANTITIES .................................................................................................................................... 105 8.5 CONSTRUCTION ASPECTS ........................................................................................................................... 106 Land based construction ............................................................................................................................. 106 Water based construction ........................................................................................................................... 107

8.6 STRUCTURAL OVERVIEW ............................................................................................................................ 108

9. CONCLUSIONS .................................................................................................................................... 110

9.1 FUNCTIONAL DESIGN ................................................................................................................................. 110 Channel width and depth ............................................................................................................................ 111 Jetty length .................................................................................................................................................. 111 Jetty orientation .......................................................................................................................................... 111

9.2 STRUCTURAL DESIGN ................................................................................................................................. 112

10. RECOMMENDATIONS ..................................................................................................................... 113

10.1 BOUNDARY CONDITIONS ............................................................................................................................ 113 10.2 FUNCTIONAL DESIGN ................................................................................................................................. 114 Navigation ................................................................................................................................................... 114 Bypass system ............................................................................................................................................. 114 River equilibrium ......................................................................................................................................... 114

10.3 STRUCTURAL DESIGN ................................................................................................................................. 115 Overtopping ................................................................................................................................................ 115 Foundation .................................................................................................................................................. 115

10.4 COSTS .................................................................................................................................................... 116

11. APPENDICES ................................................................................................................................... 118

APPENDIX A: FIELDTRIP REPORT ............................................................................................................................. 119 APPENDIX B: PROJECT APPROACH ........................................................................................................................... 120 Project description and approach ............................................................................................................... 120 Project stages .............................................................................................................................................. 120 Work breakdown structure ......................................................................................................................... 122 Organizational structure ............................................................................................................................. 122

APPENDIX C: INCIDENTAL LOADS ............................................................................................................................. 126 Vessel collision ............................................................................................................................................. 126 Seismic loads ............................................................................................................................................... 126

APPENDIX D: TIDE TABLE ....................................................................................................................................... 127 APPENDIX E: CURRENTS ........................................................................................................................................ 128 APPENDIX F: SEDIMENT TRANSPORT ANALYSIS ........................................................................................................... 131 APPENDIX G: DETERMINING DESIGN WAVE HEIGHT ..................................................................................................... 135 Wave data ................................................................................................................................................... 135 Design life time and probability of damage ................................................................................................ 135


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Estimation of design wave height ............................................................................................................... 135 Wave statistics ............................................................................................................................................ 142

APPENDIX I: OVERTOPPING OUTPUTS ....................................................................................................................... 148 APPENDIX J: BEARING CAPACITY AND SETTLEMENT ...................................................................................................... 150 Calculation of the bearing capacity ............................................................................................................. 150 Calculation of settlement ............................................................................................................................ 152

APPENDIX K: WAVE PROPAGATION .......................................................................................................................... 155 Validation .................................................................................................................................................... 155 Design waves ............................................................................................................................................... 155 Other input .................................................................................................................................................. 155

APPENDIX L: ALTERNATIVE LAYOUTS ........................................................................................................................ 157 Alternative 1 ................................................................................................................................................ 157 Alternative 2 ................................................................................................................................................ 157

APPENDIX M: AUTOCAD DRAWINGS ........................................................................................................................ 160

13. REFERENCES ................................................................................................................................... 180


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List of figures Figure 1: Brazil (www.wikipedia.org) .................................................................................................................... 17 Figure 2: Santa Catarina and municipality Araranguá (wikimedia.org)................................................................. 18 Figure 3: Rio Araraguá (maps.google.com) ........................................................................................................... 18 Figure 4: Final Work Breakdown Structure ........................................................................................................... 20 Figure 5: Location of the jetty ............................................................................................................................... 25 Figure 6: available shapes for rubble mound jetty (Van Baars et al., 2008) ......................................................... 26 Figure 7: Reference levels Araranguá ................................................................................................................... 27 Figure 8: Seasonal variation of net sediment transport along the coast for the period of 1998 to 2008. (CPE, 2009: 76) ............................................................................................................................................................... 31 Figure 9: Annual variation of the net sediment transport along the coast for the period of 1998 to 2008. (CPE, 2009: 77) ............................................................................................................................................................... 31 Figure 10: Simulation of the net sediment transport rate in the Current Situation, for wave case 2 during low tide. (CPE, 2009: 43) .............................................................................................................................................. 32 Figure 11: Simulation of the bathymetry of after a 3 year simulation for Alternative 2. The thin black line is the course of the coastline in the Current Situation. Note the original mouth of River Araranguá closing itself off after three years. (CPE, 2009: 72) ......................................................................................................................... 33 Figure 12: Direction and magnitude of the waves (CPE, 2009:11)........................................................................ 34 Figure 13: Bathymetry region ............................................................................................................................... 39 Figure 14: Bathymetry Araranguá ......................................................................................................................... 39 Figure 15: an example of a natural bypass system ............................................................................................... 42 Figure 16: Typical elements of a weir jetty ........................................................................................................... 44 Figure 17: net longshore transport and offset ...................................................................................................... 46 Figure 18: Placement of a groin updrift ................................................................................................................ 47 Figure 19: An example of the use of spurs in a jetty system ................................................................................ 48 Figure 20: Rubble mound breakwater .................................................................................................................. 56 Figure 21: Breakwater with a berm ...................................................................................................................... 56 Figure 22: Berm breakwater ................................................................................................................................. 56 Figure 18: Jetty lay out as proposed in the CPE report ......................................................................................... 59 Figure 24: Redundant part of breakwater ............................................................................................................ 60 Figure 26: Vectors and magnitude of Hs: Wave case 2 (CPE, 2009 figure 26) ...................................................... 63 Figure 27: Diagram of occurrence versus Hs and direction (CPE, 2009 figure 6) .................................................. 63 Figure 28: Tidal flow pattern near coastal inlet (+= sedimentation area; ‐ = erosion area). (Van Rijn, 2005:5.4) 65 Figure 29: Top view of shortening river reach (Crosato, 2008: 48) ....................................................................... 67 Figure 30: Longterm variations caused by shortening of the river course (Crosato, 2008: 50) ............................ 68 Figure 31: Top view of a river with channel narrowing (Crosato, 2008: 39) ......................................................... 68 Figure 32: longterm variations caused by river narrowing (Crosato, 2008: 41).................................................... 69 Figure 33: Basin dimensions .................................................................................................................................. 74 Figure 34: Active and dead storage on a beach adjacent to a weir jetty system, Weggel (1981) ........................ 74 Figure 35: Determination of dead and active storage (Appendix N) .................................................................... 75 Figure 36: Use of mass curves to predict volumetric capacities for updrift beach and deposition basin (Weggel, 1981) ..................................................................................................................................................................... 76 Figure 37: Use of mass curve to determine the progression of the net sediment transport over time ............... 77 Figure 38: Jetty Lay out ......................................................................................................................................... 80 Figure 39: Rubble mound cross section (CIRIA, 2007:793) ................................................................................... 82 Figure 40: Recommend cross section zero‐moderate overtopping (U.S. Army corps of engineers, 2006:VI‐5‐115) .............................................................................................................................................................................. 83 Figure 41: critical area for damage on a roundhead (USACE, 2006: 2243) ........................................................... 90


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Figure 42: tetrapod dimensions (Verhagen et al., 2009:305) ............................................................................... 91 Figure 43: Design graph for low crested structures above SWL (Van der Meer, 1995:293) ................................. 93 Figure 44: typical transitions in concrete armour unit breakwaters (CIRIA, 2007: figure 6.22) ........................... 94 Figure 46: Parameters used for the NN modeling of wave overtopping discharge ((Van Gent et al., 2007) ..... 101 Figure 47: Lay out dimensions ............................................................................................................................ 110 Figure 48: Cross section locations ....................................................................................................................... 111 Figure 49: Work breakdown structure ................................................................................................................ 122 Figure 50: Worldwide avaraged seismic hazard (PIANC/MarCom 34, 2001) ...................................................... 126 Figure 51: Harmonic constants (FEMAR, 2000) for Region Araranguá ............................................................... 127 Figure 52: Current velocity and direction for wave 1 (NE‐SW) at high tide (CPE, 2009: 26) ............................... 128 Figure 53: Current velocities and directions for wave 1 (NE‐SW) at low tide (CPE, 2009: 27) ............................ 128 Figure 54: Current velocities and directions for wave 2 (SW‐NE) at high tide (CPE, 2009: 28) .......................... 129 Figure 55: Current velocities and direction for wave 2 (SW‐NE) at low tide (CPE, 2009: 28) ............................. 129 Figure 56: Current velocities and direction for wave 2 at low tide with jetties present (CPE, 2009: 35) ........... 130 Figure 57: Currents velocities and direction for wave 2 at high tide with jetties present (CPE, 2009: 35) ......... 130 Figure 58: Simulation of the net sediment transport rate for Alternative 2, wave case 2 during low tide. (CPE, 2009: 51) ............................................................................................................................................................. 131 Figure 59: Simulation of the net sediment transport rate for Alternative 2, wave case 2 during high tide. (CPE, 2009: 50) ............................................................................................................................................................. 132 Figure 60: Simulation of the relative change of net sediment transport rates for Alternative 2 compared to the Current Situation, for wave case 2 during low tide. (CPE, 2009: 51) .................................................................. 132 Figure 61: The erosion and sedimentation pattern after a 1 year simulation for Alternative 2. Green implies sedimentation, red implies erosion. (CPE, 2009: 68) .......................................................................................... 134 Figure 62: Direction and magnitude of the waves (CPE, 2009: 11) .................................................................... 137 Figure 63: Exponential exceedence graph .......................................................................................................... 140 Figure 64: Gumbel exceedence graph ................................................................................................................. 140 Figure 65: Weibull exceedence graph ................................................................................................................. 141 Figure 66: relation between gamma value and threshold, all waves ................................................................. 142 Figure 67: Tidal prism (P) versus jetty spacing (W) (U.S. Army corps of engineers, 2006: II‐6‐49) ..................... 146 Figure 68: Escoffier curve (d'Angremond and Pluijm‐van der Velden, 2001: 143) ............................................. 146 Figure 69: Iz distribution ..................................................................................................................................... 153 Figure 70: Bottom profile perpendicular to coastline ......................................................................................... 156 Figure 71: Alternative lay out 1 ........................................................................................................................... 158 Figure 72: Alternative layout 2 ............................................................................................................................ 159 Figure 73: ACAD drawing: Active and Dead storage ........................................................................................... 161 Figure 74: ACAD drawing: Basin dimensions ...................................................................................................... 162 Figure 75: ACAD drawing: Cross section 1 .......................................................................................................... 163 Figure 76: ACAD drawing: Cross section 2 .......................................................................................................... 164 Figure 77: ACAD drawing: Cross section 3 .......................................................................................................... 165 Figure 78: ACAD drawing: Cross section 4 .......................................................................................................... 166 Figure 79: ACAD drawing: Cross section 5 .......................................................................................................... 167 Figure 80: ACAD drawing: Cross section 6 .......................................................................................................... 168 Figure 81: ACAD drawing: Cross section 7 .......................................................................................................... 169 Figure 82: ACAD drawing: Cross section 8 .......................................................................................................... 170 Figure 83: ACAD drawing: Cross section 9 .......................................................................................................... 171 Figure 84: ACAD drawing: Cross section locations .............................................................................................. 172 Figure 85: ACAD drawing: Lay out dimensions ................................................................................................... 173 Figure 86: ACAD drawing: Roundhead North ..................................................................................................... 174 Figure 87: ACAD drawing: Roundhead South ...................................................................................................... 175


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Figure 88: ACAD drawing: Tetrapod plan (Dn=1,6m)........................................................................................... 176 Figure 89: ACAD drawing: Tetrapod cross section (Dn=1,6m) ............................................................................. 177 Figure 90: ACAD drawing: Tetrapod plan (Dn=1,4m)........................................................................................... 178 Figure 91: ACAD drawing: Tetrapod cross section (Dn=1,4m) ............................................................................. 179


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List of tables Table 1: Overtopping criteria ................................................................................................................................ 24 Table 2: Tidal levels ............................................................................................................................................... 28 Table 3: Design storm wave height, direction SSW .............................................................................................. 35 Table 4: Design storm wave height, direction ENE ............................................................................................... 35 Table 5: Input parameters SwanOne .................................................................................................................... 36 Table 6: Results wave propagation modelling with SwanOne .............................................................................. 37 Table 7: basic values for calculation ...................................................................................................................... 38 Table 8: Multi Criteria Analysis for Sand Bypass System .................................. Fout! Bladwijzer niet gedefinieerd. Table 9: Input and output parameters of river narrowing .................................................................................... 70 Table 10: Calculations of wave transmission for three cases ............................................................................... 72 Table 11: Cumulative net sediment transport volumes (S‐N) .............................................................................. 77 Table 12: Characteristics of Cutter Suction Dredges dependent on pipeline diameter ........................................ 78 Table 13: results of stone size calculation............................................................................................................. 97 Table 14: Tetrapod dimensions in meters ............................................................................................................ 98 Table 15: Final stone dimensions .......................................................................................................................... 99 Table 16: Cross section dimensions ...................................................................................................................... 99 Table 17: Input and results of empirical method ................................................................................................ 101 Table 18: Input parameters CLASH ..................................................................................................................... 102 Table 19: Final freeboard values ......................................................................................................................... 103 Table 20: Tetrapod quantities ............................................................................................................................. 105 Table 21: Rock quantities .................................................................................................................................... 105 Table 22: Comparison of the used amount of rock ............................................................................................. 106 Table 23: Final Cross section dimensions ............................................................................................................ 109 Table 24: Final cross section dimensions ............................................................................................................ 112 Table 25: Stone dimensions ................................................................................................................................ 112 Table 26: Wave data ........................................................................................................................................... 136 Table 27: Storm data, all waves, threshold = 1,5 m ............................................................................................ 139 Table 28: Design storm wave height, all waves .................................................................................................. 141 Table 29: Design storm wave height, direction SSW .......................................................................................... 141 Table 30:Design storm wave height, direction ENE ............................................................................................ 141 Table 31: wave steepness in a storm .................................................................................................................. 143 Table 32: Wave statistics ..................................................................................................................................... 144 Table 33: Optimalization freeboard for each cross section (H1/1) ....................................................................... 148 Table 34: Check overtopping for H1/475 wave conditions ..................................................................................... 149 Table 35: Optizimalization cross‐section 4 for H1/475 .......................................................................................... 149 Table 36: Comparison between wave propagation modelled in Delft3D by CPE and wave propagation modelled in SwanOne ......................................................................................................................................................... 155 Table 37: Waves to be modelled with SwanOne ................................................................................................ 155


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List of symbols

Roman symbols Symbol Description Unit

Ab bay area m2

ab bay amplitude M

Ac inlet cross‐sectional area m2

B Width M

B relative breakage %

B0 Average river width M

Bcore core width M

Beammax vessel width M

Bt toe width M

C cohesion strength kN/m2

C Concentration %

c’ weighted cohesion kN/m2

cpl plunging coefficient ‐

cs surging coefficient ‐

Ct coefficient of transmission ‐

D water depth M

D Draught M

D discharge pipeline diameter M

D characteristic armour unit length tetrapod M

d15 Stone diameter exceeded by 85% of stones M

d85 Stone diameter exceeded by 15% of stones M

dc factor for the depth of the foundation ‐

Df bedding layer thickness M

Dn nominal diameter of tetrapod M

Dn50 mean nominal diameter of armourstone M

dq factor for the depth of the foundation ‐

Dtoe50 mean nominal diameter of toe stone M

Du50 mean nominal diameter of underlayer stone M

dγ factor for the depth of the foundation ‐

E modulus of elasticity MPa

F Frequency Hz

G gravitational acceleration m/s2

Gc Crest width M

H water depth M

h0 Average river depth M

Hb water depth over toe berm M

Hi initial wave height M

hl leeward berm or shoulder height M

Hm0 significant wave height M

Hs significant wave height M


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Hs total water depth M

Hs,toe significant wave height at toe M

Hss significant storm wave height M

Ht wave height threshold value M

Ht transmitted wave height M

ht seaward toe level M

ib0 Bed slope ‐

ic factor for the horizontal load ‐

iq factor for the horizontal load ‐

Iz strain influence factor ‐

iγ factor for the horizontal load ‐

KD Hudson stability factor ‐

L wave length M

L Length M

LOAmax length over all, max M

M safety margin M

M armour unit mass Ton

M50 median mass of the armour units kg, t

Mc50 mean core stone mass Kg

Mt50 mean toe stone mass kg/t

Mu50 mean underlayer stone mass kg/t

N number of waves in a storm ‐

Nc factor for the bearing force ‐

Nod stability parameter ‐

Nq factor for the bearing force ‐

Ns number of storms ‐

Ns stability parameter ‐

Nγ factor for the bearing force ‐

P Probability %

P permeability of the structure ‐

P tidal prism m3

p’max maximum effective stress kN/m2

Q Discharge m3/s

Q probability of exceedance

q’ effective stress at foundation depth kN/m2

Q0 initial water discharge m3/s

QL longshore sediment transport to left m3/year

qmean Overtopping m3/s/m

Qnet net sediment transport m3/year

QR longshore sediment transport to right m3/year

R vertical motion due to wave response M

R radius of roundhead M

Rc crest height M

S characteristic concrete tensile strength Mpa

S0 initial sediment transport m3/s

sc factor for the shape of the foundation ‐


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Sd damage parameter ‐

Sl shoulder width M

smax Sinkage M

som deep water mean wave steepness %

sq factor for the shape of the foundation ‐

Ss shoulder width M

sγ factor for the shape of the foundation ‐

T wave period S

T tidal elevation M

ta armour layer thickness M

teff effective work time s/week

TL Lifetime Years

Tm mean wave period S

Tp peak wave period S

tu underlayer thickness M

tyears Time Years

U wave orbital velocity in shallow water m/s

U current magnitude m/s

V Volume m3

V flow velocity in the pipeline m/s

Vr flow velocities m/s

vw wind velocity m/s

W minimum channel width M

Wb bank clearance M

WBM basic width M

Wi additional width M

Wp separation distance M

X coordinate in SAD 1969 M

x’ coordinate local coordinate system M

Y coordinate in SAD 1969 M

y’ coordinate local coordinate system M

ΔZb lowering of bed profile M

Δzi height of every sub layer M

ΔZw lowering of water level M Greek symbols Symbol Description Unit

α slope angle Degrees

γ safety factor ‐

γ density consolidated kN/m3

γ’ density unconsolidated kN/m3

γ’ effective volumetric weight of the soil ‐

γb influence factor berm ‐

γf influence factor roughness elements on the slope ‐

γm material factor ‐


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γv influence factor wave wall ‐

γw density water kN/m3

γβ influence factor of oblique wave attack ‐

ξcr critical surf similarity parameter ‐

ξm

surf similarity parameter‐

ξm‐1,0 Breaker parameter ‐

Ρ Settlement M

ρc density of the concrete kg/m3

ρr density of the rock kg/m3

Σ pressure/stress kN/m2

Φ angle of internal friction Degrees

φ’ effective friction angle Degrees


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1. Introduction In periods of heavy rainfall and storms, the town of Araranguá in the southern part of the state Santa Catarina in Brazil tends to flood. This is caused by a shortcoming in hydraulic efficiency of the river Araranguá. The river tends to migrate from south to north and this migration decreases the hydraulic efficiency. During events of high discharge the river Ararangua causes flooding upstream. To stop the migration and increase the hydraulic efficiency the construction of jetties at the mouth of river Araranguá is proposed.

A design for a jetty system at this location is available, but outdated. Preceding research has shown that the jetties indeed cause lower water levels upstream. However, due to the construction of a jetty system longshore sediment transport will be interrupted. Because at this location a large net longshore sediment transport from south to north takes place, this interruption will cause morphological changes. At Araranguá river a large and highly varying longshore sediment transport is present with a high energy wave climate and micro tidal range. These conditions cause increased morphological changes when constructing a jetty system. The objective of this research is to design a jetty system at the predefined location in the mouth of river Araranguá. Both a jetty layout and structural design should be made. The jetty system should prevent flooding upstream and design methods are based on modern design standards. Furthermore a sand bypassing system must be applied to allow for longshore sediment transport to take place. In this project a structured approach is used conform to the Systems Engineering (SE) theory. This design approach is used to list all requirements, design options and steps to be taken for the jetty system design. Data are acquired from previous research and from the initiator of this assignment, the company Coastal Planning and Engineering. The data are used to define the boundary conditions and to calculate loads on the structure. An extensive literature research is done to list design possibilities and retrieve design methods for structural layout, dimensions and sediment bypass system. The most suitable options are worked out as a conceptual design of the jetty system. This report covers the entire design cycle for the jetties in the mouth of river Araranguá. Chapters 2 described the approach that was adopted and the functional analysis of chapter 3 lists the required performance of the system. All boundary conditions are put together in chapter 4. The choices for sediment bypass system, breakwater type and material that are made in the trade‐off are presented in chapter 5. Chapter 6 describes all steps for the design of the layout, sediment bypass system and other aspects that are present in the final functional design of the jetty. In the structural design of chapter 7 the detailed dimensions are calculated. Finally the conclusion list all outcomes of the design effort and recommendations are given on an all uncertainties and possible improvements.


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2. Project description In this chapter background information is given on the project; area and location of the project site in the first section, secondly a problem description and finally the objective of the project.

2.1 Area and location Brazil Brazil is the largest country in South‐America and the fifth‐largest country in the world (see Figure 1). It covers over 8,5 million km2 and borders almost every country on the continent. It has a coastline of over 7000 km along the Atlantic Ocean and an extensive river system of which all rivers are draining to this ocean. The climate in Brazil is generally speaking tropical, although further divisions can be made. The official language is Portuguese.

Figure 1: Brazil (www.wikipedia.org)

Nowadays, Brazil is a republic divided in 5 regions and 26 states, inhabited by 18 million people. The project area is located in the Santa Catarina state which is located in the most southern region (South Region). (Project group CF83, 2008)

Santa Catarina Santa Catarina has almost 600 km of coastline. The largest part of the population are descendants of European settlers. The inheritance of especially the German immigrants, is still visible in this state. The state capital is Florianópolis which is located on the Santa Catarina Island (See Figure 2).


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Figure 2: Santa Catarina and municipality Araranguá (wikimedia.org)

To investigate how jetty systems are constructed in Brazil and in particular the state of Santa Catarina a field trip has been made to see three different jetties in the area. During this fieldtrip also the region of river Araranguá has been visited and studied, more information is given below. The report on the findings of this field trip can be found in appendix A. River Araranguá The river Araranguá is located more or less 170 kilometer south of Florianópolis in a more remote area of the state Santa Catarina (see Figure 3). A fishing community can be found aside the migrating part of the river mouth 30km upstream the city Araranguá is located. There are no access road to the location where the jetties will be built, there is only a dune area.

Thirteen kilometers from the mouth of the river a gravel pit can be found, which produces granite. Also concrete is an easily accessible construction material in this region, because of neighbouring factories.

Figure 3: Rio Araraguá (maps.google.com)

2.2 Problem description The mouth of river Araranguá tends to migrate in northern direction over an distance of eight kilometers, due to the net sediment transport from south to north. This migration of approximately 100 m/year, causes a


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decrease of the hydraulic discharge efficiency of the river. This is not a problem during normal discharges, but in extreme conditions there are problems with flooding upstream. It is required to increase the hydraulic efficiency and stop the river mouth to migrate to prevent the hinterland from floods. To create this a jetty structure will be applied. At the bend of the river urbanization has developed, some fishermen settled there. This small fishing community has no access to the sea at the moment, because the river mouth is too shallow for their boats to pass. The construction of the jetty will create a possibility for the fishermen to access the sea, but will also cause the old river course in front of the village to shoal. In 1993 the Instituto de Pesquisas Hidroviarias (INPH) a state company has made a design for a jetty at river Araranguá, but this design is outdated and is known to have some flaws. The company Coastal Planning and Engineering (CPE) has done morphological research on the project for the situation without jetties and with different jetty designs. This morphological research confirmed that the jetties indeed cause lower water levels upstream. At this point CPE wants a new design for the jetties, which is up to date and should be in line with morphological findings. From the morphological study it was concluded that a sediment bypass system can be a major improvement to the old alternatives.

2.3 Objective

The objective is to design a jetty system at the predefined location at the mouth of river Araranguá to prevent flooding upstream. This design should be done conform up‐to‐date technology and standards. The design should include a sediment by‐pass system.

2.4 Project approach In this project a structured approach according to the Systems Engineering (SE) theory was used. In SE the project is structured in different stages in order to structure the work, keep overview and work efficient. A very important part of this is the technical project plan, or project approach. This plan is used as a guideline for the structure throughout the entire project. The plan is used to define the purpose and the method of the systems engineering effort. Part of the project plan are a requirement analysis, functional analysis, time schedule or planning and the Work Breakdown Structure (WBS). In appendix B, the project plan, as conducted in the beginning of the project, is given. All steps taken during the design process are explained in this appendix. As this is the final report, there is no need to present the entire approach here, especially since it is outdated. Tasks may be performed by different persons and entire activities might have been scrapped or created, as designing is an iterative process. Therefore only the final WBS is given here. The WBS is a result of the first investigation of all the effort needed to come to pre specified final result. It is used to define all activities that need to be done, from which a planning for the project can be made. The WBS is a living document, which means that it is constantly updated in order to function as an overview of the work to be done throughout the project. With the WBS as a basis, tasks are divided and the planning is updated. Figure 4 presents the final version of the WBS.


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Figure 4: Final Work Breakdown Structure

2.5 Requirements According to the systems engineering theory, a requirement analysis is conducted. The purpose of the requirement analysis is to define all requirements to the jetty system. As a starting point, the objective of our project, to design a jetty, is taken. In a later stage of the project, the functional analysis, these requirements will be allocated to elements of the jetty. In the functional analysis the requirements on the different jetty elements, for example the channel, will be quantified in terms of performance requirements. The requirements as derived from the requirement analysis define the following:

• What the system has to do

• How well the functions have to be performed

• The environment in which the system operates

• The constraints of the system

An important notice is that the system performance requirements do not quantify the performance of the system; they only state the performance quality of the system.

A well structured brainstorm session, according to systems engineering specifications, resulted in the following requirement on jetty as the total system. Requirements from customer objectives and performance requirements

1. The jetty must prevent flooding upstream the river 2. The jetty must have a sediment bypass system to allow for longshore sediment transport 3. Recreational‐ and fishing boats must be able to pass through the jetty 4. The jetty must be designed for a lifetime cycle of 50 years 5. The jetty design must allow for pedestrians to walk on the jetty structure


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Requirements from constraints and system environment 6. The jetty must be designed at the specified location in the mouth of the Araranguá river 7. The construction materials are limited to granite blocks and prefab concrete elements 8. The jetty design must comply with global en local regulations 9. The jetty must be designed using off‐the‐shelf technology 10. The jetty must be designed using available datasets


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3. Functional Analysis The purpose of the functional analysis is to transform the requirements that were identified through requirements analysis into a description of system functions that can be used to guide the design synthesis that follows. During concept‐ and detailed design it must be known what the system must do, how well, and what constraints will limit design flexibility. The required system functions need to quantify all these properties. In this project, the purpose is to design a jetty based on existing technology. From literature many aspects of jetty design can be identified, which will help to allocate all different requirements to parts of the jetty. Through this step (the functional analysis) performance requirements can be defined on the elements of the jetty. In literature the requirements on different functions of a jetty are presented in terms of regulations and standards. In this functional analysis all performance parameters are defined that come forth from these regulations and standards and allocate them to elements of our jetty design. Starting from the requirements as derived from the requirement analysis, performance parameters are allocated per requirement. Below the performance requirements are indicated per system requirement.

3.1 The Jetty must prevent flooding upstream the river To prevent flooding the jetty keeps the river from migrating, so the construction of the jetty on its own is already the measure to prevent flooding. But there are also other criteria to be met to comply to this requirement; the cross‐sectional area of the channel; the length of the jetty; stability of jetty construction. Cross‐sect ional area channel The cross sectional area of the channel has to comply with different criteria:

• Cross‐sectional area should be sufficient for navigational purposes (see also section 3.3)

• Cross‐sectional area equilibrium can help minimizing sedimentation

• Cross‐sectional area needs to accommodate extreme discharges of river Araranguá; without exceeding maximum water levels upstream

Length ‘The appropriate seaward extend of jetties is a compromise between navigation benefits gained by longer jetties and reduced natural sand bypassing to the downdrift beaches offered by shorter jetties. Furthermore construction costs will increase with jetty length, not only because of additional material, but also because stable armor stone sizes increase as jetties are built into deeper water’(Hughes and Kraus, 2006). For the jetty length therefore the following criteria have to be met (Van de Graaff, 2006):

• the length of the jetties should extend beyond the littoral transport zone, to reduce sedimentation at the entrance of the channel due to longshore bypassing to a minimum

• The length of the jetty will affect the tidal currents, the longer the structure the more impact, the contraction of the currents just near the entrance hamper a safe sailing of vessels and might lead to scour (stability) at the front of the jetties

Structure stabil i ty The structure stability should be such that a design storm (specified later) can be encounterd without loss of function. The amount of damage has to be specified.


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3.2 The jetty must have a sediment bypass system to allow for

longshore sediment transport According to the Coastal Engineering Manual (USACE, 2006), the sediment bypass system must comply with the following requirements:

• The sediment bypass system must ensure that (more or less) the same net amount of sand is transported from south to north as was the case in the original situation, either continuously or discontinuously, dependent on the choice of the type of bypass system

• The total volume of sand that has to be mechanically transferred must be minimized

• The sediment bypass system must not allow for sediment entering the navigation channel and cause additional shoaling

• The sediment bypass system must be able to withstand a design storm i.e. not loose its function

3.3 Vessels must be able to pass through the jetty Although the channel only has to be designed for small fishing boats and will probably not be critical a few requirements have to be regarded.

• The navigation channel should have minimal cross sectional area for navigation

• The navigation channel should protect against high current velocities

• The navigation channel should protect against waves and wave breaking

• The ebb shoals and sedimentation of the channel must not obstruct navigation

• The navigation channel should have the smallest possible length and minimize bends

• The navigation channel should have a minimum cross‐current and cross‐winds

• The navigation channel should have a small angle with the waves for better navigability.

3.4 The jetty must be designed for a lifetime cycle of 50 years During its design life time, a jetty is exposed to several storms. The jetty design can never be made such that the jetty is able to withstand all storms that it can possibly encounter during its design lifetime. A design storm, from which the loads in the Ultimate Limit State (ULS) are calculated, needs therefore to be estimated from wave data.

To determine the design storm, a probability must be chosen that the design storm is encountered during the operational life of the jetty. For this probability holds: “The probability of failure of a structure is partly a financial problem (the extra cost of lowering the probability of failure has to be lower than the capitalized cost of failure), and partly depends on non‐monetary values, such as loss of life, ecological damage, etc.”(Verhagen et al., 2009). In our case failure of the jetty structure does not mean that the jetty is immediately unable to fulfill its primary function, which is flood control. Therefore no lives are threatened directly by failure of the jetty. Because the jetty does not function as a port, no (or negligible) financial damage will be suffered when the jetty is unavailable. Financial damage only includes repair of the structure, which can be carried out relatively easy. Summarized this means that damage done to the jetty by a storm it encounter is relatively small in terms of both monetary and non‐monetary values.


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A thorough optimization of the cost of repair versus the cost of lowering the probability of failure for the jetty design is beyond the scope of this research project. Therefore the probability of failure is estimated with reference to literature. Acceptable values for the probability of failure of a jetty lay between 5% and 20%, where it is noted that an economic choice for most breakwaters will be around 5 % (Verhagen et al., 2009). In our case, relatively small damage is done by failure of the jetty. However, it must be noted that necessary repair is always costly and the probability should be kept as low as possible. For our jetty design a probability of jetty failure as high as 10% during its operational lifetime, seems a reasonable economic estimation and will be used throughout the design. The jetty lifetime, together with the probability of failure, determine the design storm for this jetty. The design storm is one of the boundary conditions that will determine wave loads in the ULS. Also incidental loads are regarded in appendix C; seismic loads and vessel collision. As explained in this appendix these loads can be neglected in this project.

3.5 The jetty design must allow for pedestrians to walk on the jetty structure

To comply with this requirement the overtopping criteria have to be determined that will allow pedestrians under normal conditions. This will not be necessary during extreme design conditions, because it is assumed that people will not go fishing or be on the jetty during these circumstances. Also the criteria for overtopping under extreme design conditions are determined in this chapter, because this will have impact on the stability of the structure. Overtopping criteria The overtopping criteria are retrieved from PIANC guidelines (PIANC, 2001) and the overtopping manual (EurOtop, 2007). For overtopping criteria the mean overtopping discharge is often used to judge allowable overtopping, however this will never describe the real behavior of wave overtopping. The major part of the overtopping discharge during storm conditions is only from a small number of waves, so the amount of overtopping discharge (m3/s/m) from a single wave during the peak of the storm can be much larger (100x) than the mean overtopping discharge. This is a restriction in the way of calculating overtopping. The amount of overtopping which we want to accept is set in two levels. As said before pedestrians aren’t considered to be on the jetties during extreme events. Some small damage is accepted during these extreme events. The average overtopping discharges are set so in comparison of PIANC guidelines (PIANC, 2001) and data given in the overtopping manual (EurOtop, 2007). In the table below the criteria are given.

Table 1: Overtopping criteria

Condition state Mean overtopping discharges

(liters/s/m) Normal service conditions 0,1 Extreme design conditions 10

3.6 The jetty must be designed at the specified location in the mouth of the Ararangua river

The location is predetermined by Coastal Planning and Engineering (CPE), depicted in Figure 5. This location is


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chosen because it is the shortest way to the sea and therefore will decrease the potential of flooding upstream

the most. The exact location is defined as: 28°55´39 S , 49°20´47 E

Figure 5: Location of the jetty

Due to erosion/sedimentation patterns discovered further down the river (CPE, 2009) it has been suggested to do more research on the location of the jetty. This research will not be done during this project and the jetty will be located at the location that is defined above.

3.7 The construction materials are limited to granite blocks and prefab concrete elements

One of the demands from the client in order to keep the jetty design affordable is to use construction materials that can be found in a close proximity of the construction site. In this case these materials are granite rocks from the nearby granite pit and prefab elements from a nearby construction company. Also, in relative close proximity a construction plant for tetrapodes (see below) is located. When using prefab elements, several options are available. First of all, a caisson breakwater can be constructed in a desired shape. For rubble mound breakwaters, several options are available as shown in Figure 6. The most common shapes are Rock, Tetrapode, Accropode (which is the same as Core‐loc) and Xbloc (Van Baars et al., 2008) . It must be noted that Accropode, Core‐loc and Xbloc are trademarked, which means that use of these items can be more costly.


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Figure 6: available shapes for rubble mound jetty (Van Baars et al., 2008)

3.8 The jetty design must comply with global en local regulations The use of reliable literature to retrieve design methods, models and formulas should ensure that this requirement is met. For structural safety, one should work with partial safety factors for both load and material. In the calculation for the design conditions, a probabilistic calculation is made for the wave conditions. Because of this no partial safety factor for the load is required as an acceptable safety margin has already been included. For the granite as well as the concrete a material factor for the strength of γm = 1,1 is used, as will be described in the chapter on boundary conditions. The material factor is multiplied by the stability factor, Ns, to incorporate the safety factor throughout the design (More information on this matter in chapter 7). The use of Design High Water (DHW) should take care of the safety factor on water levels (more in chapter 4). This water level includes all components that increases water levels.

3.9 The jetty must be designed using of‐the‐shelf technology This requirement is a constraint that means that no new inventions should be done. No laboratory research or measurements on site shall be conducted during this design project and research will be limited to literature study. The aim of this project is to design a jetty according to modern standards. Literature study will establish a knowledge basis of modern technology. Eventually the design process will lead to a design according to modern standards using available (or off‐the‐shelf) technology.

3.10 The jetty must be designed using available datasets Due to the short period of time available for this project, further data gathering is not or hardly possible and is therefore omitted. Used datasets (wave, current, soil and climate conditions) are the ones provided by CPE and UNIVALI. We will make an effort to validate the data by comparing with other information sources.


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4. Boundary conditions In this chapter all boundary conditions will be defined, divided in five paragraphs; hydraulic and meteorological boundary conditions, wave conditions, geotechnical data and other boundary conditions. Most of the data are extracted from the report of CPE in which the morphological and numerical modeling is done on the construction of jetties at Araranguá river (CPE, 2009). For the setting of the boundary conditions we will assume that this data are correct and this data from the numerical modeling will be used.

4.1 Hydraulic and Meteorological boundary conditions In this paragraph all data concerning hydraulic and meteorological conditions are given. This is divided in different paragraphs; tide, river conditions, storm surge, wind, sea level rise and sediment transport.

All values for water levels in this report were reduced to the same vertical datum (IBGE). IBGE is the value for Mean Sea Level used in Brazil named after the institution that monitors the sea level, in order to fulfill the demands of national geodesy. The full name of the institution: Instituto Brasileiro de Geografia e Estatica. In Brazil you also have another vertical reference level used for nautical maps, DHN (Diretoria de hidrografia e navegacao), which can be compared with Chart Datum. This level is all throughout Brazil the same as the lowest tidal level (MLLW). And then you also have local reference levels RN (Referencias de Nivél). In Figure 7 all reference levels for the Araranguá area are shown.

Figure 7: Reference levels Araranguá

Water levels Water levels are of great importance for the design of a jetty system, as they will determine several dimensions of the structures. The water levels depends on different influences; the tide, sea level rise and storm surge. With this different levels we can determine the design high water (DHW) for our system, which is the extreme water level for designing. Tidal levels For the tidal difference at River Araranguá the same value will be used as in the numerical modelling study done by CPE (2009). This tidal regime is in turn the same value that was used in the previous design study for different jetty alternatives at the river mouth (INHP, 1993). The values of the harmonic constants, amplitude and phase, were used as a basis for defining these tidal characteristics and derived from (FEMAR, 2000), a validation and calibration done by CPE corresponds well with the predicted progression of the tide by FEMAR (shown in appendix D). The harmonic constants were measured at coordinates 28 º 55.2 'S / 49 º 20.5' W, which is near the Araranguá river channel. Important parameters:

• Tidal period: 11 h 50 min


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• Maximum tidal range: 0,89 m

Table 2 gives all tidal levels, referenced to IBGE.

Table 2: Tidal levels

Description Notation Reference IBGE Mean Level ML ‐0,11 mMean Lower Low Water MLLW ‐0,32 mMean Higher High Water MHHW 0,12 mMean High Low Water MHLW ‐0,18 mMean Low High Water MLHW ‐0,06 m Highest Astronomical Tide HAT 0,44 mLowest astronomical Tide LAT ‐0,56 m

An increase in the value of the tidal range will only be caused by weather conditions when cold fronts enter the region. Other meteorological influences are treated in the next section on storm surge. Storm surge Beside the tidal conditions other variations in water level are due to meteorological factors, such as wind and atmospheric pressure, causing storm surges. The southeastern coast of Brazil is frequently affected by meteorological disturbances such as cold fronts, which are sometimes associated with intense tropical cyclones. These disturbances cause oscillations on the sea surface, generating low‐frequency motions (DE OLIVEIRA et al., 2009). Research done on storm surge (Truccolo, 1998) in São Francisco do Sul, is in the southern part of Santa Catarina, showed a surge of 0,8 meter and has a occurrence probability of 2% per year and lasts two days. Assumed is that the storm surge near Araranguá will be in the same order: 0,8 m Future sea level rise Many different opinions about future sea level rise exist, however it is clear that future sea level rise is very import to take into account for flood defense structures. According to the Rock Manual (CIRIA, 2007:335) ‘the present consensus is that in future the rate of rise will probably increase to about 5 mm/year’. The vagueness of this statement shows that no one knows what the sea level rise will be. In general it is accepted that a mean sea level rise will cause the same increase in all water levels, including extreme values for calculation. Despite the uncertainty, newly designed structures currently take a safety margin into account for future sea level rise. Different governments use different approaches, ranging from a description of contingency responses to a certain value of increase of the water level during the design lifetime (CIRIA, 2007). In this project we will take into account a future sea level rise of 60 cm per century, which means 30 cm for a design lifetime of 50 years. This is the same value as adopted by the Dutch government. Design High Water (DHW) The highest design water level is determined by different parameters that cause higher water levels than under normal conditions. This DHW will be used to determine the crest height for the jetties. To determine the design level on the structure not only separate cases have to be looked at, but load combinations should be taken into account. The best way to determine these load cases is to use a joint probability analysis to discover what load cases are likely to happen at the same time, but mostly it turns out to be impractical or impossible. Therefore it is omitted in this design project. The highest design water level is given by the water depth in front of the structure relative to still water level (SWL), which is the same as mean sea level (MSL). Storm surge, Highest Astronomical Tide (HAT) and sea level


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rise (SLR) are added to SWL. The combination of these values will give us the most severe situation for our jetty in the 50 years design life time. In summary from the previous sections: SWL: ‐0,11 m IBGE Storm surge: 0,8 m HAT: 0,44 m SLR: 0,30 m The Design High Water (DHW) can then be calculated for every cross‐section: SWL + 1,54 m = 1,43 m IBGE For overtopping calculations the DHW is used. All components should be included for overtopping according to what is stated in the rock manual (CIRIA, 2007): if the water level at the toe of a structure is needed in a formula, then all components should be included in the design water level. For the wave propagation, modeled in SwanOne, the SLR in the calculation for Hs is left out, because the equilibrium profile of the beach adepts to sea level rise and the depths will probably not increase with 30 cm in total (Van de Graaff, 2006). Furthermore including all components in all calculation will cause a too conservative design. River condit ions Useful data for the conditions in river Araranguá are divided in discharge, depth and sediment transport Discharge The discharge of the river Araranguá is monitored daily at different stations of the National Water Agency (ANA). These are historical records from 1943 till 2004. The discharge data used in the report are extracted from two stations upstream without tidal influence, the Taquaruçu and Forquilhinha station. Extracted from the data available from ANA the discharges used by CPE (2009) are: Average discharge: Q = 42 m3/s Discharge during extreme events: Q = 800 m3/s This extreme discharge is assumed to last for two days during modeling. For flow velocities in the river a difference can be seen during low and high tide: Vr = 0,6 m/s (ebb) Vr= 0,4 m/s (flood) Cross section The average depth of the channel is 5 m, reaching up to 8,5 m upstream. The migrating part of the river mouth is shallower, between 1 and 3m. This last fact is seen as the cause of the flooding upstream during extreme events. The mean estuarine width is about 170 m near the mouth, decreasing upstream, and is about 80 m at the town of Araranguá; average width is 140m. (D'Aquino et al., 2009) Sediment transport Given the small discharge rate and the small quantities compared to the large longshore sediment transport the fluvial sediment transport is neglected. Wind condit ions The wind conditions in this area can be divided in two governing directions:


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Dominant wind direction: vw = 12 m/s NNE / ENE Wind with greater intensity: vw= 23,7 m/s with direction SW and SSW Currents Waves approaching the coast at an angle generate longshore currents (especially in the surf zone where the waves break). Additionally, the tidal range generates ebb currents and flood currents. The parameters for current velocities and directions are derived from the numerical modelling study done by CPE (2009). The figures and explanation on this can be found in appendix E.

In summary the longshore currents are: Currents during ebb: 0,42 m/s Current during flood: 1,0 m/s And for the maximimum current velocity near the structure: Currents near the head of southern structure: 1,0 m/s Current on other parts of the southern structure: 0,4 m/s Currents near the head of northern structure: 0,8 m/s Currents near other part of northern structure: 0,4 m/s Note that these are the currents that will occur with the lay out used by CPE. It is assumed that the currents in our design will be in the same order. Longshore sediment transport The sediment bypass system as an integral part of the jetty design has to ensure that the same net amount of sediment will be transported as in the original situation. However, in the case of Araranguá, this will prove to be not as simple. The reason for this is the rather large seasonal and annual fluctuations in the amount and direction of the sediment transport. As part of their research INHP (1993) used data of waves from the Ocean Waves Statistics ‐ Area 44 Marsden, and applied the formula of CERC / DELFT to each of the 48 sampled waves there, reaching:

• Transport from SW to NE: 1.654.153 m3/year

• Transport NE to SW: 1.187.143 m3/year

• Net Transport: 467.019.50 m3/year SW to NE CPE has quantified the seasonal and annual variation of the magnitude of sediment transport along the coast by using the method of energy flow. Therein, 70 wave cases used represent the wave climate in the region, resulting in two figures. Figure 8 gives the seasonal variation of the net sediment transport for the period 1998 to 2008 and Figure 9 gives the annual variation of the net sediment transport for the same period.


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Figure 8: Seasonal variation of net sediment transport along the coast for the period of 1998 to 2008. (CPE,

2009: 76)

Figure 9: Annual variation of the net sediment transport along the coast for the period of 1998 to 2008. (CPE,

2009: 77)

The oscillation in the seasonal and annual amount of net sediment transport is directly related to the characteristics of the waves that were analyzed. In Figure 8, a clear predominance can be seen in the SW‐NE direction. This can be explained by the cold fronts passing through the region during autumn and winter (in the southern hemisphere), bringing in strong winds and waves from the southern quadrant. During spring and summer, waves come predominantly from the east/north‐east, resulting in a possible reversal of the net sediment transport. The pattern of the annual amount of net sediment transport, as shown in Figure 9, is remarkable, with reversed directions in the years 1998, 2001, 2004 and 2008. This variation is associated with the variability of wave characteristics in the region.


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The simulations in the report made by CPE (2009) indicate that the net sediment transport rate is, conclusively, in the SW‐NE direction, but annual or seasonal inversions are possible. The obtained value for the amount of net sediment transport over a period of 10 years is 550.000m3/year, which is close to the value found by (INHP, 1993) of 467,019.50 m3/year. However, due to the high variability some recommendations have been made by CPE:

• When bypassing the net amount of sediment of 550.000m3/year a margin of +/‐25% should be taken into account due to the variability in wave climate

• Bypassing processes should preferably be done shortly after the summer months (March) and extend up to the start of winter (May)

• The collected sand to be bypassed should be deposited at least 800m NE from the jetty in order to ensure its continued course along the coast and avoid that it returns to the navigation channel

In Figure 10, the path of the sediment transport and magnitudes, expressed in m3/m/s, is given for a dominant wave case from the southern quadrant in the Current situation, as follows from the numerical model by CPE (2009). The path of the sediment transport is typically over the sand bar parallel to the coastline. The pattern gives an indication for the width of the breaker zone.

Figure 10: Simulation of the net sediment transport rate in the Current Situation, for wave case 2 during low

tide. (CPE, 2009: 43)

CPE has performed numerical simulations for three different layouts for the jetty system. The best evaluated layout is called Alternative 2 and is described in Appendix F. In this alternative the river flow is redirected by the jetty orientation. The passage with the old channel is kept open for the fishing community and the opening at the end of the old river mouth is kept open as well. Figure 11 shows the influence of the sediment transport against the structure after a simulation of three years. Updrift accretion is up to 200m seaward from the cross section of the jetty with the original coastline and erosion takes place downdrift. The navigation channel starts to shoal significantly after three years.


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Figure 11: Simulation of the bathymetry of after a 3 year simulation for Alternative 2. The thin black line is the course of the coastline in the Current Situation. Note the original mouth of River Araranguá closing itself

off after three years. (CPE, 2009: 72) More information on longshore sediment transport result from numerical modelling can be found in Appendix F.

4.2 Wave conditions In this paragraph it will be clarified how different design wave heights used to calculate the different loads on the jetty structure have been derived. Wave data The wave data used in this project is a series of eleven years, 1998‐2008, obtained at the break of the continental shelf, close to the project site. Data was acquired by the National Oceanic and Atmospheric Administration (NOAA) using the program WavewatchIII (Tolman, 2009) and then translated by CPE. Wavewatch III simulates processes of generation / propagation of waves in deep water, based on meteorological data reanalyzed. The series contain the significant wave height (Hs), the peak period (Tp) and the predominant direction of the waves every three hours. Design l i fe t ime and probabil ity of damage As described in the functional analysis, the jetty will be designed for a lifetime of 50 years, with a probability of serious damage during its lifetime of 10%. The probability of serious damage during the lifetime of the construction is given by the Poisson distribution (Verhagen et al., 2009):

)*exp(1 LTfp −−= (1)

Where

• p probability of occurrence of the design storm one or more times in period TL

• TL lifetime of the breakwater in years

• f average frequency of the design storm per year


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Using TL = 50 years and p = 10%, (1) becomes:

1 *ln(1 0.1) 1/ 47550

f per year= − − = (2)

Furthermore a 1/50 year storm is estimated for structural design purposes. Estimation of design wave height This section will cover the steps taken and analysis that has been carried out for the estimation of the design wave height needed to estimate the loads in the Ultimate Limit State (ULS). In this chapter, the results of the analysis are given, with a brief description of the methods. A thorough description of the considerations and calculation methods is given in Appendix G. Method for analysis To come to an estimation of a design storm, an extrapolation was made from the wave data provided by CPE. Therefore an analysis of all the storms in the available data is made. According to the Peak over Threshold (PoT) method a storm is defined as a period in which the significant wave height (Hs) is higher than a certain threshold value (Ht). The highest significant wave height (Hss) in the storm defines the magnitude of the storm. The result of the PoT analysis is a database of all storms during the period in which the measurements (or simulation) took place. When the number of storms has been identified, an extrapolation must be made to determine the significant design wave height. Extrapolation of the storm data to produce a design wave height for a 1/475 (once in 475), 1/50 and 1/1 year storm have been done using the Gumbel and the Weibull distribution. Both are statistical methods that produce reliable predictions of future wave heights. A larger threshold value, Ht, produces fewer storms in the database, but more reliable results of the extrapolation. Therefore the threshold value will be adopted where there are at least 100 storms in the database. Based on the wave directions, which are presented in Figure 12, a choice was made to perform analysis on wave data for groups of waves from main directions SSW and ENE. Wave heights from ENE are much smaller than SSW wave heights and SSW waves are unable to hit the trunk of the NE arm of the jetty. Therefore, considering these waves separately leads to a lower design wave on the NE arm of the jetty, which leads to a smaller structure and cost savings.

Figure 12: Direction and magnitude of the waves (CPE, 2009:11)

Results of analysis Analysis for different values of Ht results in the following values for the design wave height, presented in different tables for the 2 different cases. Here, only the wave heights for a 1/475 year storm are presented, for


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the other values is referred to Appendix G. Table 3 gives the design wave heights for the waves from SSW. The waves from ENE are described by the parameters in Table 4.

Table 3: Design storm wave height, direction SSW

Ht 1,5 2,0 2,5 3 3,5 Ns 61,27 50,91 35,00 23,36 15,27

Hss,1/475 Exponential 10,32 10,00 9,71 9,43 9,14 Hss,1/475 Gumbel 9,91 9,68 9,40 9,14 8,89 Hss,1/475 Weibull 8,61 8,45 8,34 8,21 8,20

Table 4: Design storm wave height, direction ENE

Ht 1,5 2,0 2,5 Ns 42,73 23,00 10,64

Hss,1/475 Exponential 6,39 6,17 6,01 Hss,1/475 Gumbel 6,23 6,00 5,84 Hss,1/475 Weibull 5,58 5,45 5,38

Conclusion Two design wave heights that have to be established, where a conservative average value of the Gumbel and Weibull extrapolations will be used as a value for the detailed design of the jetty. It must be noted that this value will always be an estimation, for there are no exact values in wave height estimation. This means that the design wave height will be:

• SSW waves 8,7 m

• ENE waves 5,7 m The different heights are used for different purposes:

• SSW waves will be used in the structural design of the south arm of the jetty, as well as the roundheads of both arms of the jetty

• ENE waves are used in the structural design of the north arm of the jetty Wave statistics For further analysis of the wave propagation to the jetty structure, SwanOne (TUDelft) is used to calculate wave characteristics at the toe of the structure. SwanOne transforms the input parameters significant wave height, Hs and the peak period, Tp into a Jonswap wave spectrum. For the estimation of the peak period of the design storm, the wave steepness is estimated. The wave steepness is extrapolated for the design wave height, to be converted back to Tp of the design storm. For this an analysis has been made of the wave data (see appendix G). Table 5 presents the input parameters to the SwanOne model.


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Table 5: Input parameters SwanOne

Wave Direction Hs (m) Tp (s)

Hss,1/475 SSW 8,7 11,6Hss,1/50 SSW 7,5 10,7Hss,1/1 SSW 5,5 10,4Hss,1/475 ENE 5,7 10,5Hss,1/50 ENE 4,9 9,9Hss,1/1 ENE 3,6 9,0Hs All 2,95 9,0

4.3 Wave propagation To determine which wave loads on the structure have to be regarded the wave propagation has to be examined. In a previous chapter the design waves for deep water were calculated. This is the input into the 1d model SwanOne (TUDelft) which is used to predict the wave propagation and uses the deep water waves to predict the propagation of the waves in shallow waters. In appendix L the use of the program is valid be validated with waves extrapolated in the CPE report with the 3D model of Swan, here also a larger overview is given of the input data of Table 5 used to calculate the wave propagation. In this chapter only the results of the modeling are shown. Results The result of the wave modeling done with SwanOne is given below in Table 6. The results are given for the waves from Table 5 at various depths. These depths correspond with different parts of the breakwaters. This differentiation is done because it is expected that the landward side of the breakwater can be designed with smaller cross sections due to smaller wave heights.


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Table 6: Results wave propagation modelling with SwanOne

Input Results

Deep water wave height

Hs (m)

Peak period Tp (s)

Direction Water

depth (m)

Wave height Hs (m)

Mean period Tm (s)

Hss,1/475 8,7 11,6 SSW

1,0 1,3 7,3 2,0 1,8 7,6 3,0 2,3 7,8 5,0 3,3 8,3 7,0 4,2 8,8

Hss,1/50 7,5 10,7 SSW

1,0 1,3 6,8 2,0 1,8 7,3 3,0 2,3 7,6 5,0 3,2 8,0 7,0 4,1 8,4

Hss,1/1 5,5 10,4 SSW 3,0 1,8 6,9 5,0 2,6 7,4 7,0 2,9 7,8

Hss,1/475 5,7 10,5 ENE

1,0 1,5 6,1 2,0 2,1 7,2 3,0 2,3 7,6 5,0 2,9 7,8 7,0 3,3 7,6

Hss,1/50 4,9 9,9 ENE

1,0 1,4 6,5 2,0 1,8 7,0 3,0 2,1 7,3 5,0 2,6 7,2 7,0 2,7 6,8

Hss,1/1 3,6 9,0 ENE 3,0 1,7 7,3 5,0 1,9 7,6 7,0 2,0 7,5

Hss,1/475 8,7 11,6 SSE 5,0 3,3 8,2 7,0 4,3 8,7

Hss,1/50 7,5 10,7 SSE 5,0 3,3 7,8 7,0 4,3 8,4

The above design waves are used to determine all cross sectional dimensions.


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4.4 Geotechnical data In this chapter all data concerning soil and geotechnical information are given. Soi l bearing capacity It has become apparent that only very little information is available about the soil characteristics at the location of construction. No Cone Penetration Tests (CPT) are available that are near enough to give a reliable prediction of the bearing capacity. Furthermore it is beyond the scope of this project to carry out measurements at the project location. The only known variable is the diameter of the sand, which is 0,18 mm. To be able to perform calculations on the design, certain assumptions have to be made. After a meeting with CPE the following assumptions were agreed upon:

• The soil is assumed to be only close packed (or dense) sand

• No differential settlement will take place

• Values for calculation are estimated

To avoid differential settlements measures should be taken in the cross‐section design. The assumption that the soil consists of only dense sand means that the bearing capacity of the sand is high enough to carry the load induced by the jetty. However this must be verified with calculations. Furthermore it means that the soil is cohesionless , because it consists of more than 80% sand (USACE, 2004a). This also means that settlement only consists of initial settlement and creep is not a factor. The assumptions for characteristic values for the homogeneous soil (dense sand) and other assumptions that form the basis for the calculations are summarized in Table 7. Basis for these values are the known grain size.

Table 7: basic values for calculation

Description Symbol Value Angle of internal friction φ (degrees) 35 Cohesion strength c(kN/m2) 0 Modulus of elasticity E (MPa) 175 Density unconsolidated γ’ (kN/m3) 18 Density consolidated γ (kN/m3) 20 Density water γw (kN/m

3) 10 Bedding layer thickness Df (m) To be determined

Bathymetry Multiple bathymetry data is available for the river, the breaker zone in front of the river mouth and for the surrounding sea. FUNDESPA (2004), a Brazilian foundation which does research for the oceanography section of the Sao Paolo University, did some bathymetry measurements in 2004 for the river and the breaker zone. In 2008 measurements were done on the bathymetry of the Araranguá river and the breaker zone in more detail (see Figure 14). The oceans bathymetry is obtained using a digital nautical chart of the Directorate of Hydrography and Navigation (DHN), see Figure 13. Because the FUNDESPA data is only available in maps and dates from 2004, while the CPE data dates from 2009 and is available digitally, the CPE data is preferred and the FUNDESPA data is only used to check for uncertainties or fill in any blanks. Bathymetry of the ocean is only available by the DHN charts, so these will be used. The bathymetry is used in this project for modeling in SwanOne. Therefore the data has to be ordered properly to have a good insight in which data is available. Both ArcGIS and Surfer are used to comply with this. The


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result is a file where all the different data is overlaid in one map. When specific data is needed a quick look at this map will give insight in which data to use. All the data is referred to with regard to South American Datum 1969 (SAD 1969) for longitude and latitude. With regard to depth all digitized data (both CPE and DHN bathymetry) are reduced to the same vertical datum, in this case IBGE. The FUNDESPA maps are measured from RN which is a local reference level.

Figure 13: Bathymetry region

Figure 14: Bathymetry Araranguá


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4.5 Other boundary conditions In this paragraph all boundary conditions that aren’t included in previous paragraphs are described, but are nevertheless important for our design. Material Materials are limited to the use of granite from the gravel pit 13 km from the building location and the use of concrete, which is easily accessible. The design should only consider these materials, because it will not be a good economical choice to design a structure with materials that are not easily accessible. For granite there is a threshold for safety, environmental and economical reasons, so there will be no blocks available heavier than this threshold. The threshold value is 2700 kg, which corresponds to an average rock diameter of 1,0 meter. This value is etsimated from field research. For concrete there are several prefab elements on the market, some designs are outdated but still used. This is because newer elements mostly all have trademarks. CPE therefore recommends to design with concrete blocks that don’t have trademarks, so they can be used without interference of other companies. In this case it was assumed that tetrapodes which are produced in a nearby factory are the most economical design option. Equipment The equipment involved with jetty structures include equipment for construction of the jetty itself and equipment used for maintenance of the entire jetty system. The use of equipment can be limited by either financial factors or availability. Equipment for construction includes trucks for granite, other means of construction material transport, and land‐based equipment and waterborne equipment for placing armourstone on rockfill structures. The exact method of construction of the final jetty system design is beyond the scope of this project. Equipment for construction of a rubble mound breakwater is available, as these have been constructed in the area in the past. The larger part of system maintenance is concerned with bypassing sediment. Shoaling of the navigation channel between the jetties has to be either prevented or monitored. The choice of sediment bypass system is, amongst other criteria, dependent on the availability of equipment. For instance, dredging operation with use of big hoppers with a large transport capacity is not reasonable because they are too big for the project area and there probably are none in Brazil. Twelve inch cutterheads are available, but they have little operating power. Labour For labour conditions we can assume that there is at least enough knowledge to construct rubble mound structures and to use tetrapods (or prefab elements) in a construction. This assumption is based on the field investigation in the neighborhood. Polit ical The political system with respect to the application of hydraulic structures is quite different from the Netherlands. There is no central authority that is responsible for the construction, maintenance and monitoring of national hydraulic works, like ‘Rijkswaterstaat’. Instead, local and regional governments decide on whether or not a certain project has to be carried out. The process of application of hydraulic structures is limited by politics for a number of reasons that also can interrelate to some extent:

• Nearly every 4 years the local and regional government officials change posts. Therefore the governing officials give higher priority to short term projects

• A hydraulic structure project could have been high on the agenda of the current government, but can just as easily be skipped by the next


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• Priority is given to other investments, because of limited governmental budget

• Priority is given to other investments with a larger economic value

• Priority is given to other investments because of personal gain by prominent government officials

• In a large country like Brazil the national government has a limited influence and ability to control local and regional policy

Because of the political considerations described above, one can imagine that channel dredging will only be done when there is an economical interest. This economical interest could be posed by the fishermen who need to navigate in the channel. Especially when future development will bring more economic values to the area, such as a marina, the economic interest can guarantee that the channel will be dredged. . Environmental considerat ions Placing a jetty near the river mouth will have influence on the environment. Dune migration, salt intrusion, disturbance of animals and vegetation and erosion/ sedimentation will have to be considered. In this subchapter it will be explained what the exact influence of the jetty is on these issues. Dune migration Dunes in the area tend to migrate due to sand transport caused by wind. A possibility is that the channel will consume amounts of sand when dunes migrate, but this sand cannot return in the dune system. The dunes could encounter degradation because of this and eventually they could disappear totally. Also extra sediment that will enter the channel and cause more shoaling. For environmental considerations and because of shoaling of the channel it could be wise to do calculations on this matter in the future. Salt intrusion More salt intrusion could occur due to the deepening of the river mouth and the shortening of the river reach. This could be a problem for the production of drinking and irrigation water, for wildlife and vegetation. A research done on salt intrusion in Rio Araranguá for the current situation concluded that the salt intrusion reached up to 30 kilometers upstream (D'Aquino et al., 2009), this will be probably further after construction of the jetties. During this project no study on salt intrusion will be done, but when the above consequences are of any importance further investigation is needed on this subject. Wildlife and vegetation Building a jetty could have implications on the local wildlife and vegetation during the construction period and after completion. Due to the disappearance of the dunes, salt intrusion could also disturb them. An ecological study has to be done to find out what the exact effects of the jetty will be on the wildlife and vegetation, but is beyond the scope of this project. Sedimentation and erosion Sedimentation and erosion of the coast will occur due to interruption of the longshore transport. For this reason a sediment bypass system is designed. The bypass system that will be designed will contain mechanical bypassing; dredging and nourishment of the foreshore downstream. Erosion and sedimentation can affect the environment as well the effects of dredging and nourishment. A study on this matter is beyond the scope of this project.


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5. Sediment Bypassing Methods and Techniques

Coastal inlets tend to act as a sink for sedimentation. This can result in siltation of the channel which in turn results in a risk of flooding of the area upstream during times of high river discharge. The first objective for the design of the Ararangua River jetties is to increase the hydraulic efficiency of the inlet. However, the jetties will completely change the natural shoaling, channel migration and longshore sediment bypass cycles of the inlet. In order to maintain the original net sediment transport rates for the future situation it is essential to make the sediment bypass system an integral part of the jetty design. In this chapter the most commonly known artificial sediment bypass methods and their accompanied techniques are briefly discussed. Special attention is given to weir and spur jetties, because of their relative complexity and the fact that it seems to be the most likely solution in the early stage of the project. To complete our understanding of this topic some background information on natural bypassing is also given. The information in this chapter is extracted from Seabergh & Kraus (2003) and the Coastal Engineering Manual (U.S. Army corps of engineers, 2006). Natural Bypassing There are three ways of natural bypassing:

• Bar bypassing

• Sediment moves along bars and ebb‐tidal shoals

• Normally large gross longshore sediment transport rate

• Tidal bypassing

• Sediment enters on flood current and exits on the other side of the channel with ebb current

• Normally large tidal prism

• Probably smaller gross longshore sediment transport rate

• Episodic Bypassing

• Caused by collapsing ebb‐tidal shoal

• Normally large tidal prism and easy wave climate

• Can be triggered by storms

Figure 15: an example of a natural bypass system

In case of the Araranguá river bar bypassing will probably be the case, because of the heavy wave climate and small tidal prism.


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Fixed systems Fixed systems are stationary dredging systems designed, built and operated for a specific location. Components of a typical fixed system are the pump and pump motor; a housing to protect the pump and pump motor from waves, spray, and surge; a boom or brace that supports the intake(s); the dredge crater (created by the intake); a discharge pipeline; and booster pumps (depending upon the requisite diameter and length of the discharge pipeline and capacity of the pump). Fluidizers can be used to increase the sand intake capability. Fixed systems are used when a continuous or high frequency sand transfer is needed. Mobile systems Mobile systems can be floating dredge plants or shore‐based plants mounted upon a vehicle. When using floating dredge plants greater quantities of sediment can be transferred and there is maximum mobility. On the other hand, the area to be dredged has to be reasonably protected by waves and easily accessible by the plant. Shore‐based plants can transfer smaller quantities of sediment, there must be a way to access the beach and the plant is limited to sand transfer from along the shore. However, it will not be hindered by weather or wave climate. Semi‐mobile systems Semi‐mobile systems include pumps or eductors that are deployed at some location in the inlet for a certain period of time and then moved to another location. Components of these systems are the dredging equipment and a temporary or dedicated pipeline or other means to discharge the sand. Use of these systems can be considered in either continuous or periodic sand transfer. Equipment for sediment extract ion

a) Dredges: A difference is made between mechanical and hydraulic (suction) dredges. Mechanical dredges place the material upon a barge, hopper or other storage area from where it is transported to the dump site. Hydraulic dredges discharge a mix of sand and water to a hopper, scow or pipeline. Both can be incorporated to either, fixed or mobile, continuous or periodic, bypassing strategies. Hopper dredges can directly recover sediment from the seabed and store it in an onboard hopper for subsequent disposal. A difference in characteristics of the dredged sand with the original sand at dump site can be an issue.

b) Jet pumps and eductors: Hydraulically powered pumps with no moving parts. A high velocity (low‐pressure) jet entrains the surrounding fluid and a mixture of water and sand is forced through the discharge line. Jet pumps can be incorporated to either fixed or mobile sand bypass systems.

c) Submersible pumps: Electrically or hydraulically driven pumps lowered directly into the material to be transferred. They offer the potential to pump material at higher solids contents than jet pumps or conventional dredges, and require neither a clear water intake, supply pump nor supply lines. Risk of clogging the discharge line is fairly high due to the capability of producing high solids‐content slurries.

d) Fluidizers: Fluidizers can be used to help any of the previously mentioned dredging systems to increase the mobility of the seabed sediments and potentially increase the uptake and production rate of the bypass system. It injects clear water in the vicinity of the suction intake. Limited prototype experience and chance of clogging are potential problems for the use of fluidizers

Equipment for discharge

a) Hydraulic (pipeline) discharge: For modest transport distances, direct hydraulic transfer by fixed, temporary, floating, or submerged pipeline can be used.


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b) Hopper dredges, barges and scows: A distinction is made between closed‐hull and open‐hull. Closed‐hull dredges, barges or scows can only be filled and emptied from the top so the sediment must be pumped out or mechanically transferred for either nearshore or beach disposal. Open‐hull means the dredge, barge or scow can be open from the bottom in order to drop the sediment directly on the seabed. It can also discharge the sediment to the beach or shallow waters by pump‐out or mechanical transfer.

c) Mechanical, land‐based discharge: Truck transport (if adequate roads and beach access locations are available). Suitable only when transport of small quantities of sand is needed. It should be considered as a method of last resort.

d) Offloaders and rainbow dredges: Offloaders are hydraulic (pump and pipeline) or mechanical (belt‐conveyor) systems integral to the hopper, barge, or scow, that are capable of transferring the sediment to the nearshore or beach. Rainbow dredge discharge is a hydraulic offloading system whereby a shallow‐water hopper dredge offloads sediment by a bow‐mounted pipeline that discharges a jet of slurry in an arc toward the shoreline. Mild beach slopes and deep hopper drafts can be a limiting factor for the application of this technique.

e) Hydrocyclones: A hydrocyclone separates coarse‐ and fine‐grained particles within slurry by using centrifugal force. Its potential application is to isolate coarser‐grained, beach‐compatible sediment from unsuitable sediments (silt and clay) dredged during maintenance or expansion of waterways. Because of their reasonably modest capacities multiple hydrocyclones will be needed. Application in coastal engineering is mostly untested.

Weir Jett ies A weir jetty system is a method for bypassing sediment at coastal inlets. The weir section of a jetty system is a depressed region that permits waves and the longshore current generated by wind, waves, and tide to transport sediment moving along the coast to enter a deposition basin located in the lee of the weir, thereby reducing the amount of sediment entering the navigation channel. As the deposition basin gradually fills up, dredging is needed on a regular basis. The dredged sand can be used for nourishment of the beach down drift (e.g. a shadow zone).

Figure 16: Typical elements of a weir jetty


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Additional benefits of a weir in a jetty system are:

• Its function as a breakwater

• The availability of a semi‐protected area for dredging the deposition basin

• The possibility of flood currents entering over the weir during flood flow and the subsequent channelling of ebb flows out the navigation channel between the jetties, promoting net seaward sediment flushing

• A possible shorter jetty length than conventional jetties, because of a minimization of sediment transport along the outside of the jetty (Seabergh and Lane, 1977)

Variations in weir jetty systems depend on:

• Structure orientation

• Bathymetry

• Presence of bottom features such as shoals and rock reefs Jetty weirs have been constructed from concrete sheet piles, wooden piles, and rubble rock. The typical crest elevation is between low water up to about mid‐tide. Rubble rock weirs are more easily adjusted later on, if necessary, and encounter less wave reflection. A highly permeable jetty offers an even greater wave protection. Weir elevation with respect to inlet hydrodynamics is a subtle, but key parameter for the timing of maximum flows and flow volume over the weir. Inlets with a small bay tide compared to the ocean tide range have maximum currents in the inlet at high and low water. If the inlet bay tide range is nearly as great as the ocean tide range, then maximum flood and ebb currents occur at mid‐tide level. In this case a lower weir is possible. Weir elevation is determined from:

• Tide range

• Wave height

• Inlet bay response

• Magnitude of left‐ and right‐ directed longshore sediment movement Weir length is a trade‐off between wave protection in the channel and deposition basin area and the possibility of diverting significant amounts of sediment seaward. Local beach slope is also a factor in determining how far seaward the weir should extend. Flatter nearshore slopes require longer weirs to prevent too much sediment from bypassing the weir with a potential to enter the channel. Problems of weir jetty systems are, for example, scour along the structure, tidal currents migrating through the deposition basin, and sediment bypassing the basin near the shoreline. A (weir) jetty system usually consists of two jetties on either side of the inlet or navigation channel, e.g. a north jetty and a south jetty (off course dependent on the orientation of the coastline). A typical response in a single‐jettied system is that the ebb shoal collapses in towards the single jetty, forcing the channel against the structure. This can result in scour along the toe of the jetty and may eventually cause jetty failure. Therefore this response must be taken into account during construction when for instance the construction of the north jetty is completed, whilst the south jetty has yet to be constructed. Other recorded causes of failure of weir jetty systems in the past include:

• A wrongly chosen location of the deposition basin with respect to the position of where the sand moves over the weir, typically at the swash zone.

• Arising problems for navigation through the inlet due to increased wave heights and crosscurrents in the entrance.


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• A greater amount of sand passing over the weir than expected, resulting in a significant amount of sand deposition in the navigation channel.

• Redirection of the navigation channel due to unexpected spit migration. There are several possible modifications for a weir jetty system when it has proven to be inefficient:

• Construct an impoundment basin training structure to function as a groin in halting encroachment of a spit to the channel and to direct longshore sediment transport into the basin.

• Raising the weir to mean sea level or to the same crest height as the seaward part of the jetty.

• If the dredging schedule is irregular or the transport episodic, a revetment wall can be constructed at the back of the basin for confinement. This could however be sensitive to scour and could be a hazard to navigation.

• To prevent sediment bypassing the deposition basin another possible solution could be to cover a portion of the landward edge of the weir with stone, creating an offset with respect to the basin ().

Figure 17: net longshore transport and offset

In a situation where the longshore transport is not unidirectional (so it varies from right‐to‐left and from left‐to‐right), ideally, the net longshore transport is captured in the deposition basin (see Figure 17 ). A portion of the left‐to‐right movement must be prevented from entering the deposition basin. Offsetting the weir seaward of the original shoreline may be beneficial with regard to the creation of a temporary fillet region that fills when the longshore sediment transport is directed from right‐to‐left. After the fillet has grown to the shoreward edge of the weir, sediment will begin entering the deposition basin. If right‐side jetty orientation will permit wave approach to move sediment out of the fillet region back up‐coast, this is possible. If the right‐side jetty creates a large shadow zone so that sediment does not move left to right out of the fillet, a possible up‐coast groin may hold the sediment (Figure 18).


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Figure 18: Placement of a groin updrift

Spur Jett ies A jetty spur may be defined as a relatively small structure added to the main breakwater. The spur will typically be perpendicular to the jetty, but can be oriented at an angle up to 45 degrees. The spur can be placed on both sides of the breakwater; on the inside to redirect large currents to prevent scouring and/or jetty instability and on the outside of the breakwater to prevent sediment from entering the inlet due to the redirection of the longshore current. In this case only the latter will be discussed. Spurs:

• Can alter the direction of the longshore current and can contain or divert the sediment

• Also provide wave height reduction

• Could also be used on the down‐drift breakwater when feedback is expected

• Used in a weir‐jetty system can reduce the wave heights for dredging operations in the deposition basin.

When designing the spur you will have to take the following into consideration (The information below is retrieved by a little empirical research in 2003. This information can be outdated or not correct):

• Spur location

• Spurs are usually located at 75% (60‐100%) of the length of the breakwater from the shoreline.

• Spur length

• The spur length divided by distance from spur to local average shoreline (S/L) should be smaller then 0,4. Any bigger and the shoreline will reach the spur and this will eventually cause sediment transport seaward, which is what you want to avoid.

• Spur elevation

• Spurs usually have the same height as the breakwater, but could also be constructed as a reef‐spur, which will probably reduce costs. The reef‐spur hasn’t been investigated properly.

• Spur material

• Spurs are usually made of the same material as the breakwater.


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Figure 19: An example of the use of spurs in a jetty system

Sediment traps Specific areas of an inlet system can be designated, designed, or dredged to serve as a deposition basin to trap sediment, either with or without weir jetties. The sand trap should be located in areas where natural wave and current processes will readily transport sand into the trap, but not out of it. The location should be a (semi‐) protected area to make periodic operation of a dredge or bypass pump possible and not hinder navigation. Traps can be located on the downdrift side of a weir section, or at other locations where chronic shoaling is observed (e.g., at the seaward ends of jetties, at interior spits, across the flood shoals, etc.), or at locations where shoaling is induced by the construction of breakwaters or other structures. Traps can be located within the inlet (interior traps) or outside the inlet jetties (exterior traps). Channel wideners Channel wideners serve four purposes:

• They decrease the frequency of requisite maintenance dredging (which, in turn, reduces dredging costs by reducing the number of dredge mobilizations.

• They improve the reliability of safe navigation.

• They provide a designated basin into which littoral sediment can deposit and be less likely to mix with non‐beach‐compatible sediments of the channel.

• They increase the quantity of littoral sediments available for a given dredge job and is thereby more economically attractive than beach disposal dredging practices.

F lood shoals Interior flood shoals can be used as a source of sand for inlet‐adjacent beach nourishment. However, its use is limited by differences in characteristics of the sand of the flood shoal. Grain size is often different from the finer beach sand and there might be restrictions to dredging activity because of presence of environmentally sensitive resources. The typical thin, widespread flood shoals require the use of smaller dredging equipment. Extensive removal of sand from the flood shoal will likely lead to a more hydraulically efficient inlet and will probably increase the rate of interior shoaling. As long as the geometry of the shoal borrow area is designed so as to act as a trap and the location does not draw sand from the adjacent beaches, this is not problematic.

Ebb shoals Like flood shoals, ebb shoals can offer an attractive opportunity as a sand source. At inlets where the ebb shoal functions as a natural bypass for longshore sediment transport care must be taken with respect to the amount of dredged material and the location (preferably, sand is borrowed from the seaward edge of the sand bar). When done improperly severe erosion can occur in the inlet. Also in this case extensive removal will increase the hydraulic efficiency of the inlet.


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6. Trade off The trade off is used to pick the best design for the jetty system. This design can be worked out in detail later on in the design process. The objective of doing a trade off is that all design decisions have been made and that the detailed design phase will only be used to optimize the jetty design. The method for trade off, as well as the criteria, depends on what is known and how easy it is to make a selection. In this section a Multi Criteria Analysis is used to select the most suitable type of sand bypass system, and a balancing of advantages and disadvantages is made to choose the type of jetty construction material. With the outcomes, a jetty design should be possible to make in a way that it will meet all requirements.

6.1 Sediment Bypass System trade off This paragraph is dedicated to the selection of an adequate sediment bypass system for the River Araranguá jetties. The various bypass systems and techniques as described in the previous chapter will be reconsidered and the best solution will be chosen. The steps that will be taken in order to find the most appropriate system, are retrieved from the Coastal Engineering Manual (USACE, 2006), functioning as a guideline in the design process. The first step is to define the design requirements for this site specific project. Then, based on the qualities and characteristics of the different bypass system types a choice will be made, by using the Multi Criteria Analysis method. In the MCA the four main system types are; fixed systems, mobile systems, semi‐mobile system and other systems. Other systems include; weir jetties, channel wideners, sediment traps and offshore breakwaters, since these systems have the common characteristic of receiving sediment by natural inflow, whereas the first three mechanically collect sediment. Finally, a single system remains as the best option for the Araranguá River. Design requirements For proper engineering design of an inlet improvement and bypassing plan, it is necessary to:

a. Define the problem. b. Determine the need for jetty improvements or other structure/channel modifications. c. Define the requisite quantities and expected seasonal/annual variations thereof. d. Define the appropriate physical method(s) for the work. e. Define the location(s) of sediment removal. f. Define the appropriate location(s) for sediment placement.

The latter four requirements are, ultimately, interrelated. a. Define the problem The jetty system design for River Araranguá investigated by CPE (2009), by the name of ‘Alternative 2’, does not include a sediment bypass system. As a result, channel shoaling, updrift accretion and downdrift erosion occurs. A sediment bypass system can solve these unwanted effects by:

• Restoring the original net longshore sediment transport rate

• Prevent shoaling of the navigation channel or present an acceptable solution to shoaling (i.e. dredging activities)

b. Determine the need for jetty improvements or other structure/channel modifications Alternative 2 contradicts with the following main requirements of this jetty design project:


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• The jetty must prevent flooding upstream the river; after just a few years the channel depth will decrease due to shoaling, resulting in a decrease in hydraulic efficiency and an increased risk of river flooding further upstream .

• The jetty must have a sediment bypass system to allow for longshore sediment transport; to pose a solution to the unwanted effects mentioned above.

• Recreational‐ and fishing boats must be able to pass through the jetty; shoaling may hinder navigation due to a decrease in channel depth.

c. Define the requisite quantities and expected seasonal/annual variations thereof As described in chapter 4 in the paragraph on longshore sediment transport, discussing the sediment transport boundary conditions:

• The seasonal/annual variations are given in Figure 8 and Figure 9

• The net sediment transport rate is 550.000 m3/year with a margin of +/‐25% because of these variations

d. Define the appropriate physical method(s) for the work The appropriate physical methods are a fixed system, a mobile system, a semi‐mobile system or another system. In the second part of this chapter a MCA will be done in order to determine the best option for this jetty system. e. Define the location(s) of sediment removal The locations of sediment removal can be, depending on the final bypass system method:

• The SW side of the southern jetty (predominant wave direction from the southern quadrant, updrift)

• A designated location for a deposition basin or sand trap

• The navigation channel itself f. Define the appropriate location for sediment placement Following from the recommendations of CPE, the appropriate location for sediment placement will be approximately 800m NW of the northern jetty, just beyond the shadow zone (southern quadrant). Deposition should typically be in the surf zone where wave breaking occurs. In this manner the shore will find a new equilibrium in a natural way. Qualitat ive trade‐off for the choice of bypass system The aim in this phase of the design is to use the qualitative background information to make a reasonable choice for the best suitable bypass system for the River Araranguá jetty design. The arguments made are based on technical, economical and circumstantial criteria. General classes of bypass operations include:

a. Interior vs. exterior traps b. Periodic vs. continuous c. System type (fixed, semi‐mobile, mobile or other)

a. Interior vs. exterior traps First of all, the navigation channel of Alternative 2 acts as an interior trap, because bypassing and wave action cause the channel to shoal. The updrift accretion against the SW jetty, as a result of the dominant wave action from the southern quadrant, can be seen as an exterior trap. When the wave action is from the north‐eastern quadrant during summer, some accretion takes place at the north‐eastern side of the NE jetty. Ultimately,


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sedimentation takes place at the tip of the SW jetty connecting this shoal with the ebb‐shoal (see Figure 20, chapter 1.3). b. Periodic vs. continuous Because of the high variability of the seasonal and annual sediment transport rates and directions, either periodic dredging has to take place or continuous pumping by a dedicated plant during a designated period of time in the year, presumably winter. Continuous operations have an intercepting character and periodic operations have a collecting character. c. System type There are a variety of bypass systems applicable to jetty designs (see appendix H). These systems can be divided into four categories. In short, fixed systems are stationary dredging plants mounted upon a structure, semi‐mobile systems are dredging plants that can be moved from one location to the other, mobile systems are floating dredge plants or shore‐based plants mounted upon a vehicle, other systems include weir jetties, sediment traps, channel wideners or offshore breakwaters.

Multi Criteria Analysis

In order to decide on which system type is best suitable for River Araranguá a Multi Criteria Analysis (MCA) has been done. Each of the four main system types is judged on a number of criteria and is awarded with points from 1 up to 5, according to their performance in relation to the other system. All the criteria have a certain weighing factor, dependent on its relative importance, expressed in a percentage. All together, the weighing factors of the criteria together add up to 100%. The list of criteria is given below with a short description. Judgements are based on the systems’ weakness towards those criteria. In other words, the system is judged on its dependency on:

1. Frequency; the systems’ capability of dealing with high frequency sand supply during extreme conditions.

2. Sediment transport rate capacity; techniques that often comply with a certain system might decrease the transport rate capacity to some extent.

3. Mobility; flexibility of multiple areas that can be dredged. 4. Accessibility; accessibility with respect to method of construction and maintenance works. 5. Deployment conditions (downtime); dependency on weather (wave) conditions or seasonal

deployment 6. Sediment characteristics; dependency on the degree of high solid contents in the debris. 7. Transport distance; techniques that often comply with a certain system might have a larger

dependency on transport distance. 8. Bottom features; performance with respect to a large variability in bottom features. 9. Flexibility in case of bad performance; ability to adjust or alter the system if it does not meet up to

the requirements. 10. Negative effect on navigation; operations might hinder navigation. 11. Long term costs; qualitative estimation of costs (construction, operation and maintenance) of a

certain system relative to the other and with respect to jetty structure costs. 12. Application; successful achievements in the past, i.e. certainty of success.

The result of the MCA is given in Table 8. All the criteria have been awarded points which in turn have been multiplied by a weighing factor and added up to obtain a final score. Before drawing conclusions from this table, some clarifying information is needed to understand the logic.


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Important additional notes;

• Land‐based mobile systems are left out of consideration here, because these systems are limited to sand intake from along the beach and need easy access to the intake and deposition location. This cannot be achieved in the vicinity of River Araranguá.

• Sand traps and offshore breakwaters are also left out of consideration; because these ‘other’ bypass systems require the construction of additional breakwater structures to the jetty system. The aim is to reduce the construction material as much as possible, in order to reduce costs.

Table 8: Multi Criteria Analysis for Sand Bypass System

Criteria nr. Weighing factor Fixed Semi‐mobile Mobile Other 1 15.0% 5 3 1 4 2 5.0% 4 4 3 5 3 5.0% 1 4 5 3 4 15.0% 3 4 4 3 5 10.0% 2 3 3 4 6 2.5% 2 2 3 5 7 2.5% 4 4 3 5 8 2.5% 5 3 4 3 9 10.0% 2 3 4 2 10 2.5% 5 2 3 4 11 15.0% 3 3 4 4 12 15.0% 4 2 4 3

Score 100% 3.3 3.075 3.375 3.525 Explanation towards chosen weighing factors The weighing factors are chosen in accordance to importance of the criterion. The importance of a certain criterion is mainly dependent on the circumstances at a specific location and is therefore different for every inlet. In this case, the weighing factors describe the importance of a criterion with respect to the circumstances at River Araranguá. Circumstances that characterise the Araranguá inlet are; a high energy wave climate, small tidal range, remoteness, small shipping activity, mild shoreline slopes, fine sand and the need for high hydraulic efficiency. Explanation towards the points awarded, per criterion, based on the theory retrieved from the Coastal Engineering Manual (USACE, 2006) and (Scholten, 2002). Only the systems’ weakness towards the criterion is pointed out.

1. Frequency; floating mobile systems are more efficient when dealing with constant, small scale (some 10.000 cu/yr) sand supply. At Araranguá, the sand supply is much larger with great seasonal variation. Semi‐mobile systems are less easily relocated to the exact location of sand deposition during high frequency sand supply in storm conditions.

2. Sediment transport rate capacity; again, mobile systems will probably prove to be less efficient with respect to the large amount of sand to be bypassed at Araranguá. Pipeline discharge frequently used in combination with fixed and semi‐mobile systems can transport the required sediment transport rate, but is not unlimited.

3. Mobility; fixed systems are limited to one intake location. Weir jetties and channel wideners cover a limited area from where the sand can be dredged. In this case, sand bypassing the deposition basin is unwanted.


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4. Accessibility; fixed systems are mounted on the jetty structure and therefore require a more sophisticated construction method. Maintenance operations for the submerged parts can be difficult. Construction of the weir jetty system might require offshore operated equipment, because of limitations by shore‐based infrastructure at the River Araranguá inlet. Maintenance might have to be done from offshore as well.

5. Deployment conditions (downtime); in the situation of Araranguá, choosing a fixed system will result in a structure that will only have to operate for several months a year. Semi‐mobile and mobile systems are more vulnerable to wave conditions and weather circumstances.

6. Sediment characteristics; all the systems requiring sophisticated intake equipment and discharge pipelines have an increased risk of clogging due to a possible high solids content in the debris.

7. Transport distance; all systems can easily cover the required transport distance of approximately 800m. Mobile system operations might have a more difficult procedure, because of constantly varying distance. Pipeline discharge is more costly, but reasonable for short distances.

8. Bottom features; semi‐mobile systems need extensive research to find the right location to be dredged. At weir jetties and channel wideners it is very hard to estimate the exact amount and location of where the sand deposits.

9. Flexibility in case of bad performance; when having determined the wrong intake location or calculated the wrong sediment transport rate capacity required, steering towards improvement will prove to be difficult, especially for fixed systems. When at a weir jetty or channel widener the sand bypasses the deposition basin, failure is obvious and steering requires drastic measurements.

10. Negative effect on navigation; mobile and semi‐mobile systems will hinder navigation most, because their operations mainly take place in the navigation channel.

11. Long term costs; fixed and semi‐mobile systems require a large investment and are costly to maintain. Weir jetties have large initial costs, but are more attractive in the long run.

12. Application; some techniques used in combination with semi‐mobile systems have not yet been applied extensively in practice and are therefore more vulnerable to failure. Quite some recordings have been made where weir jetties or channel wideners fail because of sand bypassing the deposition basin. Fixed and mobile systems have sometimes proven to be incapable of providing the promised transport capacities.

Conclusion from the MCA As can be seen in Table 8, ‘other systems’ has acquired the highest score. On a qualitative basis, the best option for the sand bypass system will be either the weir jetty or channel wideners. However, from a practical perspective, it can be said that channel wideners will prove to be a difficult solution for several reasons:

• Implementing the channel wideners in the Alternative 2 jetty design of CPE will bring uncertainties on whether the sand will actually deposit in the specified areas, because simulation shows an unevenly distributed shoaling pattern. This process is much better controlled when using a weir.

• When implementing channel wideners you will have to tamper with the cross‐sectional area of the channel, possibly reducing the hydraulic efficiency.

• Choosing channel wideners will require a considerable amount of research on morphological processes and numerical modelling to prove its feasibility.

To conclude, a weir jetty design offers the best solution for a sediment bypass system at River Araranguá.


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6.2 Breakwater type selection For jetty or breakwater design two main types can be distinguished, the mound types and monolithic types. Beside the main types considered in this chapter others exist. First composite types which combine a monolithic element with a low‐crested mound type composed of stable loose elements. In fact, there is an abundance of composite breakwater designs that combine a rigid element and a flexible structure. These types of structures will be included as subtype of a monolithic type of structure. The second are special types, which are all unconventional breakwaters that have been implemented or their use has been proposed (floating, hydraulic, piles). These special types appear always to be either unfeasible or uneconomic. Therefore these will not further be considered as options for this jetty design. The two main types will be discussed in this chapter with their advantages and disadvantages, where after one of the types is selected. The type selected can then be divided in different subtypes; also for a subtype a selection will be made. All information on types and selection is retrieved from the lecture notes, Breakwaters and Closure dams (Verhagen et al., 2009) Descript ion of main types Here the two main types are described; mound types and monolithic types. Mound types Mound types of breakwaters are in fact just simply large heaps of loose elements, such as gravel and quarry stone or concrete blocks. The stability of the exposed slope of the mound depends on the ratio between load and strength i.e. wave height (H) versus size and the relative density of the elements (Δd). On one extreme, for example, is a gravel beach that is subject to continuous changes in the equilibrium profile as the wave characteristics change and also due to longshore transport. On the other extreme, for example, is the ‘statically stable breakwater’, where the weight of the elements in the outer armour layer is sufficient to withstand the wave forces. Between these two extremes is the ‘berm breakwater’, where the size of the armour is not sufficient to guarantee stability under all conditions, but where some extra quantity of material is provided so that the slope of the structure can reshape between given limits. Advantages of the rubble mounds types are:

• Simple construction

• Withstands unequal settlements

• Large ratio between initial damage and collapse

• Many guidelines available for the designer

Disadvantages of the rubble mound are:

• Dependence on the availability of adequate quarry

• Large quantity of material required in deeper water

• Large space requirement

• Difficult to use as a quay wall

Monolithic type Monolithic breakwaters have a cross‐section which acts as one solid block. Types of monolithic structures include caissons, a block wall, or a masonry structure. Monolithic breakwaters can be constructed in large parts and assembled on site. This leads to very short construction times on site. On the other hand the monolithic types are subject to brittle failure; relatively small damage can cause the entire structure to fail and reparation is expensive. Advantages of the monolithic breakwater are:


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• Short construction time on location

• Can function as quay wall

• Economical use of material in deeper water

Disadvantages of the monolithic breakwater are:

• Sensitivity to poor foundation conditions (settlements and liquefaction)

• Uncertainty about wave loads in breaking waves

• Complete and sudden failure when overloaded

• Reflection against vertical wall

• Limited support for the designer from guidelines and literature

Select ion type of breakwater The main differences between the mound and the monolithic types of breakwaters are caused by the interaction between the structure and the subsoil and by the behavior at failure. The mound‐type structures can be considered flexible (i.e. they can follow uneven settlement of the foundation layers), whereas monolithic structures require a solid foundation that can cope with high and often dynamic loads. The behavior of the structures when close to failure is also quite different. When a critical load value is exceeded, a monolithic structure will lose stability at once, whereas a mound type of structure will fail more gradually as elements from the armour layer are displaced. However, because of the sloped construction, the footprint of a rubble mound breakwater is much larger. Where construction restrictions related to depth or environmental issues are a concern a vertical wall breakwater may be the better options After the advantages and disadvantages of the two types have been compared a selection can be made. For the following reasons the selected type will be the rubble mound type:

• Rubble mound is simple type of construction that is applied widely in the area

• There is a high energy wave climate at our location and the monolithic construction has many uncertainties concerning wave loads

• Some damage is allowed on our structure, therefore it is more interesting to use the more flexible rubble mound structure

• There is no need to use the structure as a quay wall

• There are no limitations on the available space

• Granite for construction is available from a nearby gravel pit

• There is a large support of guidelines (which is in line with objectives of the project)

• Rubble mound is able to withstand unequal settlements

Descript ion of mound subtypes The mound type can be subdivided in subtypes, in this part three different subtypes of the mound types are shortly explained. These are the rubble mound, breakwater with a berm and a berm breakwater. Material use is divided in either natural rocks or prefabricated concrete units. Rubble mound A conventional rubble mound breakwater is required to be almost statically stable for the design wave conditions (See Figure 20). This type of breakwater can also be constructed with concrete units as armour layer.


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Figure 20: Rubble mound breakwater

Breakwater with a berm There is a distinct difference between a breakwater with a berm and a berm breakwater. As the name says, a breakwater with a berm is a breakwater with a clear berm (Figure 21). The function of this berm is to reduce wave load on the main armour and to decrease overtopping over the breakwater.

Figure 21: Breakwater with a berm

Berm breakwater A berm breakwater is also a breakwater with a berm, but is designed in such a way that during storms the berm may deform into a more stable profile. These breakwaters are also called dynamically stable breakwaters. Reshaping of the form of the breakwater is shown in Figure 22.

Figure 22: Berm breakwater

Berm breakwaters may be divided into three types (PIANC/MarCom 40, 2003)

• Statically stable non‐reshaping structures. In this condition only some stones are allowed to move similar to conventional rubble mound breakwaters.

• Statically stable reshaped structures. In this condition the profile is allowed to reshape into a profile, which is stable and where the individual stones are also stable.

• Dynamically stable reshaped structures. In this condition the profile is reshaped into a stable profile, but the individual stones may move up and down the slope.

Concrete units For each of the above subtypes, it holds that a choice have to be made between the use of natural rocks or prefabricated concrete elements. A number of considerations are important when making the choice between the two options Generally, the use of quarry stone will be cheaper than the use of concrete blocks, even if the availability of quarry stone is limited. A problem of using quarry stone is the fact that it is a natural material, so its quality and properties are governed by nature. This means that neither density nor maximum size can be selected freely by


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the designer. The main option for the designer who is facing problems with the stability of quarry stone armour is reduction of the slope. This decision is accompanied by a big increase in the volume of material required. At a certain point, the step towards concrete armour units becomes inevitable, because in this project a threshold value for the volume and mass of granite exists. The question that arises is whether to use simple blocks, cubes (or similar shapes) or more complicated shapes that rely on their interlocking capabilities. Nevertheless, if a decision to use the more complicated units is made, the utmost care must be taken to avoid breakage. For different types of units three approaches are distinguished:

• The use of units that obtain their resistance mainly by their weight;

• The use of units that obtain their resistance mainly by interlocking between elements;

• The use of units that obtain their resistance mainly by friction between elements

Select ion of mound subtype It is not possible to select at this stage the most suitable subtype for construction. Although no selection can be made at this stage a scheme can be given that will function as a guide in the selection of subtype during the design process. Selection subtype The procedure to select the most appropriate type of breakwater is mainly done on cost considerations. Every design starts with the design of a conventional rubble mound breakwater following the Van der Meer method. If calculated armourstone exceeds the largest stone threshold value another path have to be taken. Now three options are available. First the designer has the tool to reduce the slope of the structure, this would imply that more is space required and a larger amount of stones. Secondly it can be investigated whether it is economically more attractive to design a statically stable non‐reshaping berm breakwater. This would reduce the use of (more costly) large stones, but will increase the total amount of stones. If the threshold value of stones is even too low for this application then a wider and even more voluminous berm breakwater can be designed in the form of a dynamically stable structure or statically stable reshaped type. The last option is to investigate the costs when using concrete units. When in the design of a conventional rubble mound type the threshold value is exceeded concrete units should be applied. Here after more information on the design with concrete elements is given. Trade off mound type selection No economical analysis can be done during this project to select which of the three options should be followed in choosing subtype mound breakwater. The last option with the use of concrete units is selected for the following reasons:

• All jetty constructions in the area are build in this way; so most experience is with the construction of conventional types

• The time span of the project is too short to make an economical analysis; so for now we will assume that the use of concrete units is cheaper instead of constructing a more voluminous berm breakwater with the use of more quarry stone.

• Concrete is easy available


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Now that a subtype with the use of concrete units is selected a second decision has to be made on the type of units that will be used. The selected concrete unit is a tetrapode:

• CPE prefers a design with concrete elements that are not patented or trademarked

• Tetrapods are produced in the area

• Tetrapods are used as armour units on all breakwaters in the neighbourhood

The decision is mainly based on the first argument. A tetrapod is a concrete unit with four legs that dissipates the wave energy by allowing water to flow around it. For Tetrapods units resistance depends mainly on its weight, but they have a certain degree of interlocking as well. Selected type The designing process is proceeded with the design of a conventional rubble mound breakwater with armour layer consisting of quarry material, i.e. granite rocks. Tetrapods will be applied when threshold value of 2,7 tons for stones is exceeded.


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7. Functional Design In this chapter the functional design of the jetty system will be covered, divided in three main subjects. First the jetty lay‐out; the integral lay‐out will be determined based on shipping boundary conditions. The second topic is channel design; this is an evaluation of the first lay‐out designed for the channel, based on morphological principles. In the third paragraph, it is shown how the sediment bypass system is implemented in the design. At the end of the chapter a summary and conclusion will be given about the functional design.

7.1 Previous design The layout as proposed in the report of CPE(2009) is the main starting point of our jetty lay out (see Figure 18). The origin of this design lies in a design made by INHP (1993)and is modified by CPE. With new information available the lay out will be updated. In this chapter it will be explained what features of the jetty lay out will be changed or stay the same. In this report the channel has a width of 150m and a depth of 6m.

Figure 23: Jetty lay out as proposed in the CPE report

The jetty has been designed to lay between the following coordinates (SAD 1969): Xmin = 660800 m Xmax = 661800 m Ymin = 6798000 m Ymax = 6799400 m Curved closure of channel In the CPE report (2009) a curved extra breakwater is placed in the river in front of the canal that leads to the small fishing community (see Figure 24). It’s function is to stop sediment transport and water flow to this canal, but will allow navigation access. Numerical modeling studies (CPE, 2009) showed that without this extra curved breakwater there was no flow and sediment transport to the canal. Therefore this extra curved breakwater is redundant and will not be applied in the design.


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Figure 24: Redundant part of breakwater

7.2 Jetty lay out for shipping conditions The jetty lay out depends on the design vessels characteristics. It will play an important role in determining the width, depth, length and orientation of the jetty. Vessel characterist ics In the CPE report (2009) the jetty was designed for 5m draught vessels and a channel depth of 6m. In contemplation with CPE it was decided that the channel will be designed for smaller vessels. CPE stated that small fishing vessels as well as small and large recreational vessels should be able to enter the river. It was stated that de maximum draught of an entering vessel should be 4m. The maximum dimensions of a vessel entering the channel will be: LOAmax = 50m, Beammax = 10 m and draught = 4m. The maximum wave height that is allowed for a ship with these dimensions is 1,5m. The vessels cannot sail out when a significant wave height of 1,5m or higher occurs between the breakwaters. This value is derived from the vessel dimensions in collaboration with prof. ir. H. Ligteringen, professor at the department of “Hydraulic Engineering” at the University of Technology Delft. Channel width The channel width is mainly dependant on shipping and inlet stability. In this section only width for navigation purposes will be calculated, where the next section covers the morphological inlet stability. The minimal width of the channel for navigation purposes is determined with formulas (3) and (4) (Ligteringen, 2007) One way channel:

2BM i BW W W W= + +∑ (3)

Two way channel:

( )2 BM i B PW W W W W= + + +∑ (4)

Where:


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BMW = Basic width due to draught > 1,25 * channel depth = 1,6B

iW = additional width, dependant on multiple variables

• prevailing cross winds 15‐33 kn = 0,4B

• prevailing cross current 0.8 kn = 1,0B

• prevailing long current <0,4kn = 0,1B

• prevailing wave height 1‐3m = 1,0B

• Aids to navigation are good = 0,1B

• Soft seabed characteristics = 0,1B

• Cargo hazard = 0

iW = 2,7B (total)

BW = Bank clearance due to hard embankment = 1,0B

PW = separation distance due to ships speed = 1,2B

This will result in the following minimum widths, for a one way channel, formula (5):

1.6 2.7 2 1.0 6.3 63W B B B B m= + + = =i (5)

Two way channel, formula (6):

( )2 1.6 2.7 1.0 1.2 11.8 118W B B B B B m= + + + = = (6)

In summary for navigation purposes the channel should have a minimum width of 63m for a one way channel and 118m for a two way channel. In both cases the design width as in the CPE report is more than sufficient. Channel depth The dimensions of the channel will depend on the draught of the characteristic vessels (4m). This channel depth will be addressed below. The channel is designed for an acceptable downtime for ships with a large draft, this is at MLLW (corresponding to ‐0,32m IBGE). When choosing a higher acceptable downtime or smaller design vessels less dredging is necessary and will minimize costs. It can be reconsidered if these values are in accordance with the actual navigation, but for now the design is done using dimension as stated above. The minimal depth of the channel is determined with formula (7) (Ligteringen, 2007)

maxd D T s r m= − + + + (7)

Where: D = draught T = tidal elevation below which entry is not allowed smax = sinkage due to squad and trim r = vertical motion due to wave response m = safety margin The sinkage in the formula is estimated at 0,5m. The vertical motion is Hs/2, with this wave characteristic thus 0,75m. The safety margin depends on the soil, which in thjs case is a sandy bottom and for this the margin is 0,5m. Now the following draught results from formula (7): d = 4 – (‐0,321) + 0,5 + 0,75 + 0,5 = 6,07m with regard to IBGE


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The channel designed in the CPE report is designed for 5m draught vessels and has a depth of 6m. The dimensions of the vessels changed to 4m draught, but the minimum design depth of the channel is unchanged at 6m depth. This is due to more strict safety regulations and because the channel is designed for an acceptable downtime for ships with a large draft when the tide is below MLLW. Jetty length The jetty will have to extend seaward to at least the depth of the navigation channel and outside the surfzone; to protect the channel from sediment intrusion, to shelter from breaking waves and to protect from longshore current under non storm conditions. Making the jetty longer will maximize navigation possibilities due to maintaining the minimum required depth and will keep waves from breaking. On the other hand a long jetty has negative effects. It will make manoeuvring for large ships more difficult, it will be more expensive building a larger breakwater in deeper water and it will interrupt the longshore transport, causing sedimentation on one side of the jetty and erosion on the other. A trade‐off has to be made between the above criteria. CPE modelling (2009) showed that sediment enters the channel and shoaling will happen in front of the jetty. Making the jetty longer can (partly) diminish this. The bathymetry shows that when you extend more seaward the depth (after >7m) increases fast. This means that costs will increase fast when making the jetty longer. CPE stated that a “shorter” breakwater is preferred over a “longer” breakwater. The reason is partly political; CPE expects that when the channel shoals the government will invest in dredging equipment, but when there is accretion or erosion further up or down the shoreline, the government will not invest because of lack of financial/political gains. For the above reason it is decided that for navigational purposes the jetty length can be the same as in the alternative used by CPE. Jetty orientat ion In this section jetty orientation will be treated for shipping conditions, this will be done on three criteria; waves, currents and wind. Waves The wave penetration in the channel should be diminished by the breakwaters to the maximum wave height of 1,5 m, because of navigation restrictions of the vessels (see vessel characteristics). In the CPE report (2009) two wave cases (wave case 1: Hs=1.5m, Tp=6.8s, Angle=62° ~ ENE and wave case 2: Hs=4.3m, Tp=10.7s, angle=197° ~ SSE) have been examined on wave penetration in Delft3D (WL and Delft Hydraulics, 2006). The results (see Figure 25 and Figure 26) show that in both cases the waves are strongly diminished. Figure 27 shows the occurrence of waves when regarding wave height and direction.

Figure 25: Vectors and magnitude of Hs: Wave case 1 (CPE, 2009 figure 24)


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Figure 26: Vectors and magnitude of Hs: Wave case 2 (CPE, 2009 figure 26)

Figure 27: Diagram of occurrence versus Hs and direction (CPE, 2009 figure 6)

When wave case 2 is regarded a relatively high wave (Hs = 4,3m) with a low occurrence is used. Figure 26 shows that the wave height is reduced from 2.4m in front of the jetty to 1.6m at the entrance and to 0.8m in the channel. Therefore it’s safe to say that the channel is protected by the breakwaters from 95,2% of the southern waves or more (probability of a SE, SSE, SE, S, SSW, SW, WSW wave with Hs > 4m = 4,8%, see Figure 27). When wave case 1 is regarded (Hs=1,5m with a high occurrence) the wave height is reduced from 0.8m in front of the jetty to 0,5m at the entrance and to 0.3m in the channel. The probability of a NE, ENE and E wave with Hs < 1,5m is 32% (see Figure 27). This means that 100% ‐ 32% = 68% of the waves have a Hs > 1,5m: this is not a governing wave. In the timeframe of this project it is not possible to do modeling efforts to find out which deep water wave corresponds with a wave with Hs = 1,5m between the breakwaters. This because a 1d model doesn’t take into account any diffraction and such and 3D modeling will be too time‐consuming. For now an estimation has to be made, which has to be checked in the future in a 3D model. For now, the assumption is made that a deep water wave of 2,5m will diminish to a wave height of 1,5m at the jetty mouth and lower in the channel. Then the channel is protected by the breakwaters from 86,0% of the northeastern waves or more (probability of a NE, ENE, E wave with Hs > 2,5 = 14,0%, see Figure 27).


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The downtime or the probability of waves occurring that will result in a wave in the jetty that’s larger then Hs = 1,5m are in the range 5‐14%. For the mouth of river Ararangua this is an acceptable value. To find out what the optimal alignment is for the jetty with respect to the waves some more 3D modeling has to be done. The alignment of the jetty will not change from the jetty designed in the CPE report. Remarks

• The northeastern maximum deep water wave should be calculated using a 3D model.

• Northeastern and southern waves have been investigated, Eastern waves should also be modeled in 3D.

• Wave impact when ships leave the protection of the breakwater takes place under an angle. In case 1 this is under a small angle from the longitudal direction of the navigation channel, but in case 2 it is a large angle. This is a good reason to extend the southwestern breakwater more than the northeastern breakwater.

• Modeling can also be done for the jetty under another angle, to see if another alignment will decrease the occurrence of waves in the channel further.

Currents When looking at navigation currents should be taken into account. Longitudal currents are of no importance, but cross currents could interfere with navigation. Therefore a maximum for the cross current is set at 2kn = 3,7m/s. This value is derived from the vessel dimensions in collaboration with prof. ir. H. Ligteringen, professor at the department of “Hydraulic Engineering” at the University of Technology Delft. In the CPE report (2009) a study has been made on the influence of the jetty on the currents. This shows that the maximum current velocity in any of the normative situations is below 1,2m/s. This means that the current criteria is not a limiting factor. Wind The wind with greater intensity (23,7m/s, SW) will hit the vessels in longitudal direction, the dominant wind (12m/s, NE) will hit the vessels at 90°. The dominant wind is not a big interference with navigation because of it’s low intensity and the wind with greater intensity is not because of it’s longitudal direction.

7.3 Morphological approach on channel design To evaluate the foregoing design issues on the lay‐out based on shipping criteria, here the design of the channel from another point of view is evaluated. This will be done on morphological grounds, with the minimal depth and width for shipping as a given. This evaluation is done while varying in width and maintaining the minimum shipping depth, because width cannot be adjusted after construction. To determine the minimal channel dimensions between the jetties it will first be discussed why this is an important design issue and then the ways to approach this problem. When you construct jetties along the coast to stabilize coastal inlets or a river mouth, as is the case at River Araranguá, this will have an impact on the hydrodynamics of the area. In Figure 28 flow patterns, sedimentation and erosion areas are shown.


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Figure 28: Tidal flow pattern near coastal inlet (+= sedimentation area; ‐ = erosion area). (Van Rijn, 2005:5.4) Jetty systems are dominated by the tidal range, which is micro tidal at Araranguá river (<2m), by waves which have high wave energy (Hs>1,5m) and by the river tendency to adjust to its equilibrium position . These factors are important, because waves tend to shoal the channel between the jetties and tidal flows will take sediment out of the channel. Therefore, at river Araranguá it is clear that in the channel sedimentation might be a problem. There are more factors that are unfavorable for the channel stability. First, the large longshore sediment transport will create shoals and bars to bypass sand naturally along this interruption. Secondly, river discharge can help flushing the channel, but the mean discharge of the river Araranguá is low (42 m3/s) almost 90% of the time. Only during high discharge events the river helps flushing out sediment. The third factor is that in this area dunes tend to mitigate from south to north causing the channel to interrupt the dune migration and part of this sediment will move into the channel. The numerical simulation done by CPE (2009) shows indeed the sedimentation that was assumed after the foregoing enumeration after three years (see boundary conditions Figure 11). To reduce the impact on the morphology careful engineering is needed. To reduce the sediment transport into the channel by longshore sediment transport a sediment bypass system will be designed, but to get a more natural system you can also optimize the cross‐sectional area. For example when you reduce the cross‐sectional area the flow velocities will increase and cause scour, but excessive scour must be prevented to maintain the stability of the structure. In this chapter we will investigate if we can optimize the initial cross‐sectional area which was used by CPE for the numerical modeling (2009). The cross‐sectional area used had a width of 150 meters and a depth of 6 meters. To investigate the dimensions of the cross‐sectional area we will consider two approaches; the relationship between cross‐sectional area versus tidal prism and a river equilibrium state analysis. Channel cross‐sect ional area versus t idal prism Determining the design channel width can be done using the empirical relationship of the cross‐sectional area with tidal prism. O’Brien (1931) was the first to determine this empirical relationship, Jarret (1976) continued on the O’Brien and extended it for various coastal areas. This relationship is used in calculations to determine the channel equilibrium width. First the criteria on the stability of inlets proposed by Escoffier (1940) are used and second the graph on tidal prism versus jetty spacing of O’Brien is used to calculate the channel equilibrium width. Both of the methods were tested, but no result was obtained. Both methods couldn’t find equilibrium for which the inlet could be stable. The main reason for this is that these methods are based on inlets with a tidal basin behind the jetty. At Araranguá, there is not a kind of basin or lagoon behind the jetty structure, but a river. The tide has to travel up into the hinterland to create a tidal prism and there is a continuous inflow of fresh water. We can conclude that the tidal prism at river Araranguá is too small for this calculation methods.


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The calculations made using the Escoffier and O’Brien method can be found in Appendix I River equil ibrium states The equilibrium of the cross‐ sectional area of the channel is influenced by the long term river equilibrium states. The construction of the jetty influences the river equilibrium state in three ways; the construction will shorten the river length, the channel between the jetties will be deepened by dredging and the spacing between the jetties can be varied. All the long term responses of the river profile to this influences will be reflected in this paragraph. First general initial parameters that are used for all calculations are given, where the first two are extracted from boundary conditions: Average river width: B0=140m Average river depth: h0=5 m Bed slope: ib0= 5*10

‐5

Power law, exponent n=5 The value for n is always bigger than three, in our case of Rio Araranguá n=5 is used, the value for n set by Englund and Hansen. The bed slope is assumed, no data are available on this. Dredging The channel has to be dredged first to create sufficient depth for navigation purposes. For the excavation of a trench in a river, it is clear that it will cause no differences in the river profile on long term scales. The excavated area will be filled up with sediment from upstream. Short term effects are erosion near the end of the dredged area. Shortening of river The jetty structure will shorten the river length to improve hydraulic efficiency. To calculate the effects of this improvement the lecture notes on river morphology are used (Crosato, 2008). In the calculation of the long term equilibrium of the river profile the width of the new channel is the same as the river width, this is more or less the case. The initial water depth (design water depth) is equal to the normal water depth considering the initial slope of the shortened part (higher than in the original channel). The sediment forming the bed of the new channel is assumed to be the same as in the old channel, so no sedimentation from seaside is taken into account. Also the sediment transport formula and the Chézy coefficient are the same in both situations. A top view of a shortening in a river reach is shown in Figure 29.


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Figure 29: Top view of shortening river reach (Crosato, 2008: 48)

The parameters in Figure 24 are in the case of the jetty construction at Rio Araranguá: L0= 5750 m L1=1250 m L2=0 ΔL= 4500 On short term the shortened reach has a higher longitudinal bed slope, higher velocity and smaller water depth. Erosion will start to take place near the downstream end of the shortened reach (end of jetty structure) and sedimentation near the downstream end (beginning of jetty structure). The shortened channel has initially higher sediment transport capacity than the old channel. Erosion and sedimentation also progress upstream until the river channel has reached the same slope and water depth as before the shortening, but every point is now closer to the river mouth and this causes a lowering of the bed and water levels upstream of the jetty (Figure 30). We can calculate the long term response with equations 8 and 9:

0( , )bz L L iΔ ∞ = Δ ⋅ (8)

0( , ) ( , )w bz L z L L iΔ ∞ = Δ ∞ = Δ ⋅ (9)

Where ΔZb=lowering of bed profile ΔZw=lowering of water level


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Figure 30: Longterm variations caused by shortening of the river course (Crosato, 2008: 50)

The calculations give us:

54500 5 10 0,450,45

b

w b

zz z

−Δ = ⋅ ⋅ =Δ = Δ =

The channel depth will decrease with 45cm. Narrowing With the jetty design we can keep the jetty spacing as wide as the original average river width or the spacing can be narrower to induce a deeper equilibrium depth between the jetties. A remark is that the original lay‐out is wider than the average channel width, so the opposite will happen, a lower depth between the jetty will form than the initial river depth. To calculate what happens when the spacing between the jetties is more narrow than the original river width the lecture notes on morphological response are used again as well as the lecture notes on River Engineering (De Vriend et al., 2006). In Figure 31 the top view of channel narrowing is given, showing a good overview of parameters.

Figure 31: Top view of a river with channel narrowing (Crosato, 2008: 39)

The analysis which will be performed is valid for low Froude numbers, lower than 0,8. The equation of Froude numbers is given below, with input of parameters at Rio Araranguá

0,6 0,119,81 3

uFrg h

= = =⋅ ⋅

(10)


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B0, the initial length is reduced to B1=B0‐ΔB in the calculation as shown in Figure 31. The increase of flow velocity and water depth in the narrowed part are short term consequences. To calculate long term response the following relations are valid in the narrowed part:

0 01 1

0 0

;Q Sq sB B B B

= =− Δ − Δ

(11)

Where, Q0=Initial water discharge S0=Initial sediment transport So

11

0 11 0 0

1 0 0

1

nn ns q Bh h h

s q B

−⎛ ⎞−⎜ ⎟⎝ ⎠⎛ ⎞⎛ ⎞ Δ

= ⋅ = −⎜ ⎟⎜ ⎟⎝ ⎠ ⎝ ⎠

(12)

and

33

011 0 0

0 1 0

1

nn n

b b bsq Bi i i

q s B

−

⎛ ⎞⎛ ⎞ Δ= ⋅ = −⎜ ⎟⎜ ⎟

⎝ ⎠ ⎝ ⎠ (13)

Where in equation 12 and 13, s=sediment transport Derived from the relationships set in equation 11‐13 the narrowing of the river width leads to an increase of the water depth and a reduction of the longitudinal bed slope, as is shown in Figure 32. Furthermore river narrowing can cause undesirable long‐term changes, it can cause bed degradation upstream of the narrowed part. This causes problems to foundations of structures along the river and increases the chance of bank failure. More bed degradation upstream will be caused, when the narrowing is longer or the bigger the width difference there is.

Figure 32: longterm variations caused by river narrowing (Crosato, 2008: 41)

Calculations for the most efficient spacing between the jetties will be done for 4 different cases, decreasing the width between the jetties with 10 meters every step. In Table 9 the input of the width is given and the results


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of new depths and slopes. Calculations were done down to a width of 80 meters, because this is a width that is the minimal width that we can allow for shipping conditions in one way.

Table 9: Input and output parameters of river narrowing

ΔB B1 h1(5m) Ib1 10 130 5,30 4,85*10‐5

20 120 5,66 4,70*10‐5

30 110 6,06 4,54*10‐5

40 100 6,54 4,37*10‐5

50 90 7,11 4,19*10‐5

60 80 7,81 4,00*10‐5

Conclusion on morphological approach For the first approach, the relationship between the cross‐sectional area and the tidal prism, all methods used are based on inlets with a tidal basin behind the inlet. At Araranguá, there is not a kind of basin or lagoon behind the jetty structure, but a river. The tide has to travel up into the hinterland to create a tidal prism and there is a continuous inflow of fresh water. We can conclude that the tidal prism at river Araranguá is too small for these calculation methods. For the river equilibrium states on long term there are results. The dredging of the channel will not have impact in the long term. The shortening of Rio Ararangua will cause a decrease in bed level of 0,45m and therefore water level of 0,45m. The narrowing of the jetties will cause a local decrease in bed level between the jetties, as shown in Table 9. Optimal design is a channel width of 120 meters, which will more or less cause an increase of depth of 90 centimeter in total. This optimal channel width will cause the highest river equilibrium depth between the jetties, which can allow shipping in two ways and does not differ too much from the initial average river width. The narrowing can’t differ too much from the initial average width, because it can cause degradation of the bed upstream. The interaction between both shortening and narrowing is not exactly known. In the shortening equations is for example assumed that width is the same, which is not the case when you narrow a part with the jetty spacing. It is not known if both measures will increase their impact or in contrary weaken the impact of each other. The above values are calculated for two breakwaters perpendicular to each other. But in this case a weir section is involved. The section next to the 6m deep channel could be (when empty) up to 12 meters deep. Only at the end of the jetty the channel width is 120m. The methods used in this chapter will probably will be influenced by the wider and partly deeper parts. The amount of influence is unknown. No alternative calculation methods are done.

7.4 Jett lay out for sand bypass system In this chapter the chosen sediment bypass system, i.e. the weir jetty system, will be designed. The entire process of sediment bypassing can be divided in three parts; sediment collection, transportation and deposition. In this case, sediment collection means the process where wave action and subsequent longshore currents guide sediment flow over the weir section where it enters the deposition basin. Transportation of the collected sediment takes place by means of dredging operation and discharge. Finally, the sediment to be bypassed is deposited in a designated location downdrift to ensure a continued longshore sediment transport.


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Design of the weir sect ion Design of the low‐crested structure (or weir section) that interrupts the regular shape of the jetty is concerned with the location of the deposition basin with respect to the shoreline, interior channel alignment, jetty orientation and typical sediment pathways; weir elevation and its controls on wave transmission and flow; weir length; composition of the weir; and weir function with respect to longshore sediment transport. Detailed design guidance on weir jetty systems is given by Weggel (1981). Since this literature has not been available during this project, the design is based on information on weir jetties from Seabergh (2002), the Coastal Engineering Manual (USACE, 2006) and case studies from Seabergh et al. (2002), which often use Weggel (1981) as a reference. Therefore this design can be considered preliminary. Weir elevation The weir elevation is dependent on wave height, tide range, inlet bay response and the magnitude of sediment transport. The crest height must be high enough to ensure a safe passage for navigation through the channel and adequate circumstances for dredging operation in the basin area. On the other hand, the crest height must be low enough to allow for the required amount of sediment to enter the basin. In general, weir elevation is at MTL (Mean Tide Level) in areas with a small tidal range (0,6‐1,5m), like the coasts of the Atlantic. The maximum tidal range at River Araranguá is 0,89m. When designing the crest height at SWL, this is at ‐0,11m IBGE (see Figure 7). Weir length The length of the weir is a trade‐off between wave protection in the channel and basin area and the amount of sediment diverted offshore along the jetty at the seaward side of the breaker zone. If the wave climate is highly variable, the weir length should extend further into the ocean in order to capture a large percentage of the breaker zone where heavy transport over the weir will occur. Another consideration is the amount of flow in the system. If high ebb dominance of flow is required in the navigation channel the weir should be longer. However, if the ebb currents are directed towards the weir, the channel could be pulled through the basin causing dispersal of the sediments coming over the weir. Additionally, ebb flow predominance can provide scouring ability in the navigation channel. The seaward end of the weir should not be too close to the navigation channel. At River Araranguá mild beach slopes and a highly variable wave climate result in a fairly wide breaker zone. The width of the breaker zone gives an indication for the length of the weir. The largest part of the wave action occurs along the stretch of the breaker zone from approximately 150m to 370m from the shoreline (see Figure 12). A first estimation of the weir length is therefore 220m. In fact, the actual weir length in the final design will be somewhat larger, because it is placed under an angle with respect to the shoreline. This can be seen in the technical drawings of the layout in Appendix N. The final lay out is highly dependent on the positioning of the weir section and deposition basin. Several alternative lay outs have been considered (see Appendix M and Figure 28) and from these alternatives a lay out has been chosen that reduces negative effect the most. Considerations for the choice of the final design are based on hydraulic efficiency, space required, breaker zone percentages and construction material reduction. The negative side of this layout is relatively high exposure to wave energy. When doing more research the other alternatives should also be evaluated. As a consequence of the orientation of the weir section in this layout, the northern jetty arm will have to be longer than in Alternative 2, lay out by CPE (2009) in order to prevent waves entering the deposition basin from the east‐northeast. Therefore, the northern jetty arm extends 630m seaward from the shoreline.


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Wave propagation over the weir Structures with low crests will transmit wave energy over the structure, causing wave action behind the structure. In this design the weir section is low crested. For safe navigation and dredging operations the wave energy transmitted over this section and the subsequent transmitted wave heights have to be calculated. This is done according to the Rock Manual (CIRIA, 2007). The severity of wave transmission is described by the coefficient of transmission, Ct, defined in terms of the incident and transmitted wave heights. This coefficient is defined below:

tt

i

HC H= (14)

The transmitting performance of low‐crested structures depends on; freeboard of the crest height (Rc), crest width (B), roughness, slope and permeability of the structure, wave height, wave period and water depth. In the rock manual (CIRIA, 2007) a simplified prediction method is given to determine the wave transmission coefficient. The relationship gives a very simple description, but can be sufficient for a preliminary estimate of performance. These equations can be used for oblique wave attack up to 70°. The angle of the weir with respect to the shoreline is based on optimalisation of the deposition basin dimensions.

2,0 1,13c

s

RH− < < − : 0,8tC = (15)

1,13 1,2c

s

RH− < < : 0, 46 0,3 c

ts

RC H= − (16)

1,2 2,0c

s

RH< < : 0,1tC = (17)

For wave transmission three cases are selected to test the performance of the weir structure design. The cases represent different water levels that can be present and will influence the value of Rc. For all cases the incoming wave height for a once in a year storm is used, at a depth of 3, which is the depth where the deposition basin is most exposed. Once a year storm conditions are taken, because it is assumed that neither dredging nor navigation will take place during more severe storms. Furthermore, it must be noted that these wave parameters are based on the predominant wave direction from the south, because dredging operations will have to be carried out under these conditions, in Table 10.

Table 10: Calculations of wave transmission for three cases

Input Output

Rc (m) Hs Ct (‐) Ht (m)

Normal water level conditions 0 1,8 0,46 0,83 Highest astronomical tide (HAT) present ‐0,44 1,8 0,53 0,95 HAT and sea level rise (SLR) present ‐0,74 1,8 0,58 1,0

A Cutter suction dredge with a dredging capacity of 60.000 m3/wk (a relatively small suction dredge), cannot work during storm conditions. It can only operate in waves lower than 0,5 meters. Possibly, it can work during ‘normal water level conditions’, dependent on the percentage of transmitted waves exceeding 0,5m, which can be quite small. In this case, the CSD has to stop working when high waves approach from the south. However, applying a larger cutter suction dredge, that can withstand higher waves, is a more appreciable situation.


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Design of the deposit ion basin The dimensions of the deposition basin are based on the sediment transport rate into the basin and the frequency of required dredging operation. In other words, the seasonal net transport rates determine the time it takes for the deposition basin to fill and subsequently the amount and frequency of dredging to be done. Under normal conditions the most sediment transport over the weir takes place at the landward side of the weir near the shoreline. Maximum sediment transport caused by wave action occurs in the breaker zone, typically in the middle of the weir. Basin dimensions According to the literature, the recommended procedure is to design for some amount greater than the yearly net, but less than the yearly gros transport directed towards the basin. Multiply that amount by 2 and take that amount as the required volume for the basin. Applying this rule to calculate the volume for the Araranguá deposition basin will result in dimensions that are out of proportion to the overall design of the system. The weir will have to be very long resulting in large exposure to channel navigation and dredging operation and either the basin will be too deep or too wide. In this case, a more logical approach, with respect to the highly variable wave climate, is to base the dimensions on the maximum net transport rate from south to north in one year. In the period from March until September very large amounts of sediment are transported in this direction. Therefore, dredging will have to be done every year for a longer period of time nonetheless. The maximum sediment transport rate is in May with a value of 270.000m3/month with a variance of 240.000m3/month (Figure 8, chapter boundary conditions). Because of the large variance, in their report, CPE (2009) has recommended to take a margin of 25% for the net sediment transport rates. Following the procedure stated above the required volume of the basin becomes: V = 1,25 * 270.000 * 2 = 675.000m3 The shoreward edge of the basin is placed at the location of the low‐water line of the adjacent beach. The water level at MLLW is IBGE ‐0,32m, which is approximately 40m from the adjacent shoreline. The seaward edge of the basin should reach as far as the end of the breaker zone where the wave energy is highest. The distance to this point corresponds with the seaward end of the weir section. This is to ensure that a large percentage of the breaker zone, where the highest transport rates take place, is covered. The length of the basin then becomes: L = 370 – 40 = 330m For the depth of the basin a value of 12m is chosen, which is twice the depth of the navigation channel. This depth is based on the abilities of the available dredging equipment and on basin depths of comparable projects in case studies. The area of the basin then becomes 65.250m2

.

The width of the basin at the shoreward edge can then be optimised (see Figure 33). Because the basin can not immediately border the navigation channel a width of 50m is taken into account for a transition zone between the beginning of the weir and the intersection with the seaward part of the jetty. The optimal shoreward basin width is 251m.


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Figure 33: Basin dimensions

Sediment transport response to the design According to Weggel (1981) the optimum weir jetty system allows only for the net sediment transport (Qnet=QR+QL) to enter the deposition basin order to reach ultimate bypassing to the downdrift beach (see Figure 34). Assuming that the net transport is directed to the right, and neglecting local, inlet induced transport, the total amount of sand moving towards the weir is QR. Therefore, an amount equal to (QR‐ Qnet) must be temporarily retained updrift to replace the sand trapped by the downdrift jetty in case of a longshore transport reversal (QL). This is called active storage.

Figure 34: Active and dead storage on a beach adjacent to a weir jetty system, Weggel (1981)

The numerical modeling study done by CPE (2009) has shown very large seasonal and annual fluctuations in transport rates and directions.


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To have an idea of the sediment transport response to the designed system the updrift accretion in the design of Alternative 2 by CPE (2009)has been further investigated. Using the accretion pattern from the numerical modeling (see Figure 11, chapter 4) and extrapolating this to the new jetty layout gives a prediction of what will happen in the new system. Part of the design is the offset from the shoreline with a length of 150m. The seaward end of this offset is partly in the breaker zone. The function of the offset is to allow accretion up to the tip and intersection with the weir section in order to create a dead storage zone behind the shadow zone of the southern jetty tip. This in turn allows for a reasonable amount of active storage updrift. If the offset is made shorter, part of the active storage will enter the deposition basin, which is what you want to avoid. If the offset is made longer, the weir will not cover enough width of the breaker zone, resulting in sediment bypassing along the tip of the southern jetty offshore or into the navigation channel (as was the case in the simulation of CPE (2009) where the channel started to shoal after 3 years). The extrapolation to determine the active storage and dead storage pattern is given in Figure 35. In the numerical simulation a pattern was found where the accretion at the landward side of the jetty reached up to 200m seaward after 3 years. So the duration of accretion reaching the tip of the offset will be more or less 2,3 years. From the geometry of this figure can be derived that the active storage covers approximately 16% of the total accretion after 2,3 years. At this stage it is hard to predict whether this active storage will be sufficient in order to allow only the net transport to enter the deposition basin. Further numerical modeling is required to be able to fully appreciate the effects of the bypass system design.

Figure 35: Determination of dead and active storage (Appendix N)

Dredging operat ion cycles The Coastal Engineering manual (USACE, 2006) describes a method adapted from Weggel (1981) to estimate the volumetric capacity requirements for the ideal system from a mass curve. The mass curve is developed


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from a plot of the cumulative longshore transport rate versus time (see Figure 36). In this case the method is used to determine the required active storage and the required storage in the deposition basin.

Figure 36: Use of mass curves to predict volumetric capacities for updrift beach and deposition basin

(Weggel, 1981) Weggel’s method can also be applied to the Figure 8 (chapter 4) of the seasonal net transport variation. However, in this case it cannot be used to determine requirements for the active storage and deposition basin volume, but it gives an indication of the progression of the net sediment transport over time in the original situation. The result is shown in Figure 37. In the figure can be seen that starting in January the net transport gradually starts to increase to positive values (meaning transport from south to north) between March/April and reaches its maximum after 8 months. At this point the net transport starts to decrease again. To convert these results to the new system and obtaining valuable information only the positive values and the values with an increasing domain must be looked at, since there is no negative (N‐S) net transport rate when there is a jetty system. The figure can tell us that in the period from approximately halfway April up to September the total net sediment transport volume is 698.000m3. Comparing this value with the acquired value for the deposition basin of 675.000m3 it can be concluded that dredging operation will definitely have to begin earlier than in September. The volumes of the net sediment transport for the months April to August are given in Table 11.


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Cumulative Net Sediment Transport

-300000-200000-100000

0100000200000300000400000500000600000700000800000

0 2 4 6 8 10 12

Month

Inte

grat

ed N

et T

rans

port

in m

3

Average Cumulative NetTransport

Cumulative Net Transport

Figure 37: Use of mass curve to determine the progression of the net sediment transport over time

Table 11: Cumulative net sediment transport volumes (S‐N)

Month Volume in m3

April 80000May 330000June 525000July 648000August 698000

For optimal use of equipment dredging operations should start when the deposition basin is full. At the end of July the volume is 648.000m3. Dredging should at least begin after this month and continued for several months depending on the equipment capacities. Using a small Cutter Suction Dredge (CSD) for example, with a capacity of 60.000m3/week the dredging cycle will have to be as follows: Starting after July the dredging duration will be: 648.000 / 60.000 ≈ 11 weeks Dredging continues, whilst the sediment will simultaneously continue to enter the basin. The additional volume entering from August until the beginning of September is 50.000m3. This remaining volume requires the following dredging duration: 50.000 / 60.000 ≈ 1 week So in total, when using a small capacity CSD, dredging operations will have to take place for more or less 12 weeks or three months, starting in August and ending in November. This immediately leads to a risk, because the months of October and November are considered to be the storm period. This means that higher than average waves will be encountered and therefore downtime can be more than expected. However, since the wave direction during this period is such that most waves are directed to the north arm of the jetty structure, the deposition basin is not subject to high waves. It is therefore concluded that this period is suitable for dredging operations.


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This length of this period can be reduced significantly by using a CSD with a larger dredging capacity. Other dredging equipment is discussed in the next section. Dredging equipment This section about dredging requirements, procedures and equipment is associated with the help of two sponsors of this project, the dredging contractors DEME and Boskalis and their websites (DEME, 2009) and (Boskalis, 2009). For the Araranguá weir jetty system design, typical dredging equipment will have to be either a Cutter Suction Dredge or a Trailing Suction Hopper Dredge. Dredging capacity, draft, length and availability determine the choice of the best type of equipment. A larger dredging capacity will result in a shorter dredging cycle. The used vessel must have a shallow draft in order to access the 6m deep navigation channel and the shoaled deposition basin. The length determines the manoeuvrability within the perimeter of the deposition basin. Together with the availability of the required equipment in the direct vicinity, these factors determine to what extent the costs of the overall operation can be minimised. Normally, the workability of a CSD is smaller than 1m of significant wave height, whilst Hoppers can operate up to 2m sea state. Another important consideration is the probability of the frequency of occurrence of the significant wave. For instance, if the height of the transmitting wave exceeds the highest wave in which the CSD can operate only 20% of the time, the overall workability of the dredge is still 80%. The 20% downtime can then be used for maintenance operations. For a first approximation of a Cutter Suction Dredge production a diameter of the discharge pipeline can be used of 500, 800 or 900mm. The flow velocity in the pipeline is 4m/s and the sand concentration 25%. The CSD works effectively for 110 hours per week, but spends 80% of the time on actual dredging. The rest of the time is spent on readjusting the CSD to a new dredging position. Time loss due to weather conditions is not taken into account. The production per week is given by:

21 0,804 effd v c tπ ⋅ ⋅ ⋅ ⋅

Where: d = Discharge pipeline diameter [m] v = Flow velocity in the pipeline [m/s] c = Concentration [%] teff = Effective work time [s/week] The results for the three diameters are given in Table 12. From the table can be derived that the dredging cycle reduces significantly when using a dredge with a larger pipeline diameter. The larger CSD will also have no problems working under conditions where the transmitting wave corresponds to the earlier found values of Ht (0,83‐1,0m). On the other hand, with a length of 100m, manoeuvrability can be a restriction.

Table 12: Characteristics of Cutter Suction Dredges dependent on pipeline diameter

Diameter Draft (m) Length (m) Max dredge depth (m) Max wave height (m) Capacity m3/wk Cycle (weeks)

d = 500mm 2,16 44,2 16 0 > 0.5 62203,53 12

d = 750mm 2,05 ‐ 2,35 48 ‐ 67,6 16 – 22 0,5 ‐ 1,0 139957,95 5

d = 900mm 4,8 100 30 < 1,5 201539,45 3,5

The use of a Trailing Suction Hopper Dredge can be considered likewise. To illustrate this with an example a relatively small TSHD with a length of 86m and a maximum draft of 5,25m is looked at. This ship can carry 1600m3 sand with a loading production of 40m3/min, a trailing speed of 2 knots and navigation speed of 10


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knots. The ship has approximately 145 effective hours per week. With this information the production capacity per week can be calculated for our design. One dredging cycle for the TSHD consists of one filling session, one dumping session and two travel paths (from the deposition basin to the dump site and back). The filling/dumping session takes 40minutes. Covering 800m to the dump site, with a navigation speed of 10 knots which corresponds to 5,14m/s, takes the ship 155,5 seconds. In total one cycle period then equals 5111 seconds. In one cycle 1600m3 is transported. The time used effectively during one week is obtained by (145/168) * 604800s which equals 522000s/week. This value divided by 5111 seconds gives 102,13 cycles per week. Finally, when multiplying this value by 1600m3 the value of the operation capacity per week becomes 163412,25m3/wk. With this operation capacity, the yearly duration of the dredging will be approximately 4,3 weeks. The results as given above give an indication for the dredging operation options for the Araranguá jetty system design. The final choice of the most suitable dredging equipment will be a trade off between costs and availability. A smaller CSD from close by will be less expensive, but will not be able to continue the operation non‐stop due to the wave conditions. Additionally, the production capacity is relatively small, which means that the contractor has to be hired for a longer period of time. Planning the dredging operation in order to avoid bad weather conditions is of vital importance. A larger CSD or TSHD will be more expensive to hire, because it has a longer way to travel before reaching the project site. However, dredging operation can be carried out non‐stop and only for a couple of weeks. Dumpsite locat ion As mentioned earlier, following from the recommendation of CPE(2009), the dredged material has to be replenished 800 meters downdrift of the jetty structure, in order to prevent the bypassed sediment returning to the jetty system during reversals. The position in the cross‐shore profile where the replenishment is placed is mostly applied at the beach (beach nourishment) or at the shore face (shore face nourishment). Cutter Suction Dredges use pipeline discharge for sediment transport to the dumpsite. The path of the pipeline follows the bottom of the navigation channel and is then guided over the northern jetty and the beach. Finally, it dumps the sand/water mixture either at the beach or at the foreshore. Trailing Suction Hopper Dredges transport the dredged sand by navigating to the dump site and dumping it by opening the hull or rainbowing it to the foreshore. Possibly, the mild beach slope at Araranguá will limit the use of an open‐hull barge. For this project the application of shoreface nourishment is considered the best solution, because the shoreface nourishment will directly bring sediment back to littoral transport system, which is the main purpose of the sand‐by passing. Cross shore transport will bring the sand back into equilibrium profile, as it is native sand this will happen in a natural way. To estimate the beach‐feeding processes proper insight in cross‐shore sediment transport processes is required. Addit ional modif icat ion options In the event that the structure as designed in this project will actually be built and turn out to fail to some extent or in certain sections, a number of adjustments are possible:

• If the hydraulic efficiency of the channel is not optimal, i.e. the channel is shoaling due to unexpected sediment impoundment; groins can be placed perpendicular to the inlet side of either the northern or southern jetty arm.

• If ebb flow currents turn out to be pulling through the deposition basin, causing dispersal of sediments, the shoreward southern jetty arm length can be increased seaward in order to divert the channel flow even more.


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• If it turns out that periods of reversals cause the navigation channel to shoal due to sand bypassing form north to south, a second weir section positioned at the shoreward end of the northern jetty arm could be an option. In addition, sediment will have to be collected in a second deposition basin and mechanically bypassed.

7.5 Summary and conclusion on jetty lay out In Figure 38 the jetty lay out is drawn. More detailed drawings can be found in Appendix N. In this chapter it will be explained what dimensions the jetty lay out has and what choices were made.

Figure 38: Jetty Lay out

Channel width and depth The channel width is determined calculating the width for navigational and morphological purposes. The width of the channel is set at 120m. The channel depth is mainly configured using the minimum depth for vessels and was used as a design parameter in the morphological approach to determine the channel width. The channel depth is set at ‐6m IBGE. Jetty length The jetty length is chosen by comparing arguments for a long/shorter breakwater. Costs, sediment intrusion, sheltering from breaking waves, longshore current, ships manoeuvrability and the interruption of the longshore transport causing erosion and sedimentation up‐ and downstream play an important role in making the decision. A modification on the previous design is done on the jetty length resulting from the implementation


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of the deposition basin in the weir section. Because of the placing of the basin waves could penetrate much further in the jetty channel, which affects sediment deposition and navigation negative, by extension this wave penetration is reduced. The length of the NE breakwater is extended till 630m in seaward direction to a depth of 6,6m. The SW breakwater stays the same and will extend 650m into the ocean to a depth of 7,2m. Jetty orientat ion The jetty orientation depends on shipping manoeuvrability and implementing the sediment bypass system. The shipping manoeuvrability depends on angle of incidence and height of the waves, the cross currents and wind strength and angle of incidence. The implementation of the sediment bypass system in the jetty lay out depends on the width of the breaker zone and on the size and the positioning of the deposition basin. Size and position of the deposition basin depend on the seasonal net sediment transport rates from south to north, annual dredging operation cycle possibilities and available equipment. Dredging operations should start when the basin is completely filled which is in the month August. The width of the breaker zone determines the length of the offset and weir section. The inland breakwater is necessary to guide the river flow in the right direction. The offset at the shoreline is placed 250m to the SW and has a length of 150m from the coast. The weir section starts at the seaward end of the offset, is placed under an angle and has a length of 300m. For details see the drawings in Appendix N and figure above.


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8. Structural design The structural design of the jetty system is in fact the detailed design of the final dimensions of the jetty. All elements will be calculated and optimized and drawings will be made for clarification. The objective is to have a design that is ready to be built. The final result will be a jetty design that meets all requirements as derived from the requirement analysis. In this chapter first the cross sectional design is done for 9 different cross‐sections in the structure. Besides the dimensions in this calculations also the mass of armourstone, concrete units will be determined and the dimensions of the toe structure that will give required stability.

After this overtopping and settlement calculations are done, this will largely determine the total height of the structure. Also more attention is given on the way of construction and the quantities of material used.

8.1 Cross sectional design In literature (USACE, 2006, CERC, 1984, CIRIA, 2007, Van der Meer, 1995) it is stressed that design formulae for jetties are suitable for conceptual design purposes only. It will always be necessary to confirm the results of the conceptual design with physical model tests. For the cross sectional design all design formulas and rules of thumb which are presented in this chapter have been extracted from the above literature, unless indicated otherwise. The design formulas are considered general knowledge, which can be found in any of the above literature. This section covers the structural design of the cross sections of the jetty, which is divided in several parts. First the parameters of the cross section are described, which mostly depend on the mass and size of the armourstone. The calculation methods for mass and size are therefore explained in the next subsection. After that special aspect are covered: toe structure, bedding layer, tetrapods and the roundhead. Finally the results are presented. Cross sect ional dimensions Figure 39 and Figure 40 present standard cross sections for rubble mound structures with wave attack from one side, which is the case in this design.

Figure 39: Rubble mound cross section (CIRIA, 2007:793)


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Figure 40: Recommend cross section zero‐moderate overtopping (U.S. Army corps of engineers, 2006:VI‐5‐

115) Below the design issues for the parameters in Figure 39 are explained together separately. Sources used for these parameters are the Rock Manual (CIRIA, 2007) and the Coastal Engineering Manual (USACE, 2006). All units, except for the slope angle are measured in meters. Crest freeboard, Rc Rc follows from overtopping requirements. The crest freeboard should be such that it allows for post construction settlement and future sea level rise. The determining of the crest freeboard is described in next paragraph 7.2 overtopping calculations. Crest width, B and core width Bcore The crest width should be minimal 3 to 4 armour stone diameters and it also depends on Bcore. The core width should be wide enough to allow for trucks and cranes when construction is done from land. For further information see paragraph 7.4 construction aspect.) Slope angle, α Generally the slope angle is as steep as possible, with a maximum slope of 1 : 1,5. When reducing the slope, the stability can be obtained using smaller stones, but more stones need to be used. In seismic areas and poor subsoil the slope should generally bee more gentle, but are both not the case here. Armour layer thickness, ta and underlayer thickness tu For an armourlayer consisting of two layers of armourstone or concrete armourstone, the layer thickness is 2 kt Dn50, where kt is the layer coefficient and Dn50 is the mean diameter of the armourstone. For rough quarry stone which is randomly placed in two layers, kt has a value of 1,00. For tetrapods, kt = 1,02. Double layers are applied to protect inner layers in case of washing away of individual stones and hence to provide stability. The underlayer also consists of 2 layers of stone, with a smaller dimension and relations are similar. Bottom elevation of armour layer When the water depth is more than 1,5 Hs, the armourstone should be extended downslope to a depth of Hs below minimum SWL. Otherwise it should be extended to the toe. In this design the water depth is about 1,5 Hs, therefore it is chosen to extend the armourlayer to the toe of the structure. Seaward toe level, ht Should extend about 1‐1,5 Hs below minimum SWL. Typically for the toe the same stones are used as for the underlayer. It should be designed such that ht/h > 0,5, otherwise the structure is regarded as a berm breakwater. The diameter of the stones is reduced if the ratio ht/h increases, and reduced when stability is not ensured, which will be explained later in the section on the toe structure.


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Leeward berm or shoulder height, hl It was chosen to extend the armourlayer on the leeside to 0,5 Hs below minimum SWL. Below this level the underlayer stone size is adopted, with a toe structure that is identical to the one on the seaside. The choice to adopt the same toe structure is a safety aspect, to assure stability of the toe in case of high river discharge. Toe width, Bt The toe width should at least allow for 3 stones to be placed. If severe scour is expected, the toe can be widened. In this case the allowable damage level for the toe can be increased. In this design it is however chosen to only vary the size of the stones at the toe, and not the dimensions. Shoulder width, Ss, Sl Ss is mainly determined by placing tolerances and should not be less than 2 m. When severe scour is expected, the shoulder has to control erosion. In this case Ss should not be less than 6 m or Hs. In this design, severe scour is not expected and 2 m is considered to be sufficient. This was concluded from preceding research (CPE, 2009)For the leeward side it holds that: Sl = 0,5 * tu. Mass of the armourstone From the above it can be concluded that most parameters depend on stone size and weight. Therefore this section covers the size and weight calculation of the stones. For the design of the jetty some assumptions have been made in chapter 4 on boundary conditions. Material availability and cost considerations have limited material options to granite rock and tetrapods. The rock is assumed to be rough angular and randomly placed. Random placement is also assumed for the tetrapods. The underlayer stone mass is 1/15 to 1/10 (see also Figure 40) of the stone mass of the armourlayer, which gives a relatively large stone size in the underlayer. A large stone size increases the interlocking with the armour layer, especially when concrete units are used. Also, the permeability parameter of the structure and therefore the stability are increased with larger underlayer dimension. Furthermore heavy stones in the underlayer increases stability during construction. Therefore the upper limit of 1/10th of the armour stone weight is used. The grading of armour and first underlayer should be narrow, to avoid stones being washed away because they are not heavy enough. Narrow grading means: d85/d15 < 1,5. Grading of the quarry run can be wide (1,5 < d85/d15 < 2,5) to very wide (d85/d15 >2,5). The nominal diameter of the armourstone rock is given by:

13

5050n

MDρ

⎛ ⎞= ⎜ ⎟

⎝ ⎠ (18)

M50 is the median mass of the armour units and ρ is the mass density of the rock, in this case granite with ρr = 2650 kg/m3, and for the concrete ρc = 2400 kg/m

3. The stability formula, also known as the Hudson formula, results in the required mass of the armour stone is:

3

50 3 cotr

D

HMK

ρα

=Δ

(19)


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Where KD is the stability factor, which can be derived from table 1 of the US design guidelines (CERC, 1984, retrieved from, USACE, 2004b). KD depends on the location of the cross section, trunk or head, on the slope and on the type of wave, breaking or non breaking. If the wave is breaking or non breaking depends on the dept at the toe of the structure. Furthermore H is the wave height and Δ is the relative mass density. Δ is given by:

1r

w

ρρ

Δ = − (20)

It must be noted that the Hudson formula is very easy in application, but it has some disadvantages:

• It uses regular waves only

• The wave period and storm duration are not taken into account

• There is no description of the damage level

• It is applicable to non‐overtopped and permeable structures only In general the Hudson formula is viewed as being somewhat outdated, therefore it was chosen to use the more recent Van der Meer equations (see below) for design purposes. The Van der Meer equations are known to be more precise and distinguish deep‐ and shallow water conditions, where deep water conditions are defined as h > 3Hs,toe and Hs,toe follows from wave modeling with SwanOne (TUDelft). Furthermore different equations are used for plunging and surging waves, which depend on the structure slope and wave steepness, see equation (21) and (22). For deep water the Van der Meer equations are: Plunging waves:

0,20,18 0.5

50

s dpl m

n

H Sc PD N

ξ −⎛ ⎞= ⎜ ⎟Δ ⎝ ⎠ (21)

Surging waves:

0,2

0,13

50

cot Ps ds m

n

H Sc PD N

α ξ− ⎛ ⎞= ⎜ ⎟Δ ⎝ ⎠ (22)

And the surf similarity parameter ξm is given by:

2

2tan / sm

m

HgTπξ α

⎛ ⎞= ⎜ ⎟

⎝ ⎠ (23)

Where:

• N Number of waves in a storm

• Sd damage parameter (rock manual 569)


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• Hs is the significant wave height, which can also be a design wave (m)

• α slope angle (degrees)

• P permeability of the structure

• g gravitational acceleration (m/s2)

• cpl plunging coefficient = 6,2

• cs surging coefficient = 1,0 To determine whether surging or plunging waves hit the structure ξm should be compared to ξcr:

10,5

0,31 tanP

plcr

s

cP

cξ α

+⎡ ⎤= ⎢ ⎥

⎣ ⎦ (24) Where for ξm < ξcr waves are plunging and for ξm ³ ξcr waves are surging. In the above equations, some of the parameters have to be determined or estimated. Storm duration The number of waves is dependent on the storm duration and the mean period Tm. These values are estimated by analyzing the wave data. For a threshold value of HT ³ 6 m, there are 6 storms in the record (see appendix.G), with an average duration of 8 hours. Therefore, for the design storms an average duration of 8 hours is assumed, together with the peak period, Tp, of the wave record depending on the direction of the waves. The design storm, which is the input for the SwanOne simulation, is a storm of 1/50 years and 1/475 years from directions ENE and SSW. Damage parameter The damage parameter Sd has been described by Van der Meer (1995) and is 2 for initial damage and 3‐5 for intermediate damage for a slope of 1:1,5. A slope of 1:2 has damage parameters of 2 and 4‐6 respectively. For a 1/50 year storm, no damage should occur and the damage parameter for this storm is 2. For a 1/475 year storm, the breakwater may be damaged, but should not fail, so a damage parameter of 5 is chosen. Permeability parameter For the permeability parameter a value of 0,5 is assumed, which is estimated according to a buildup of an armourlayer, underlayer and a core of quarry run. In shallow water, the Van der Meer equations change a little. Research has shown that the equations produce better results when H2% is used instead of Hs, where H2% is approximated by H2% =1,4 Hs. The coefficients cpl and cs change to 8,7 and 1,4 respectively. Equations (21) and (22) change to: Plunging waves:

0,20,18 0.5

50

8,7s dm

n

H SPD N

ξ −⎛ ⎞= ⋅ ⎜ ⎟Δ ⎝ ⎠ (25)

Surging waves: 0,2

0,13

50

1,4 tan Ps dm

n

H SPD N

α ξ− ⎛ ⎞= ⋅ ⎜ ⎟Δ ⎝ ⎠ (26)


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For structural safety, one should work with partial safety factors for both load and material. In the calculation for the design conditions, a probabilistic calculation is made for the wave conditions. Because of this no partial safety factor for the load is required as an acceptable safety margin has already been included. For the granite as well as the concrete a material factor for the strength of γm = 1,1 is used, as described in the chapter on boundary conditions. The material factor is multiplied by the stability factor, Ns, to incorporate the safety factor throughout the design:

50

1 ss

m n

HNDγ

=Δ (27)

Toe structure According to Van der Meer (1995), the relation of the stability parameter to ht is:

0,15

50 50

0,24 1,6s tod

n n

H h ND D

⎛ ⎞= ⋅ +⎜ ⎟Δ Δ⎝ ⎠

(28)

(28) has a range of application of:

50

0,4 / 0,93 / 25

t

t n

h hh D< <

< <

Furthermore (28) applies to “standard” toe sizes of 2‐3 stones height and 3‐5 stones wide (Van der Meer, 1995:286). The factor Nod in equation (28) is a measure for the damage level of the toe structure. For Nod =0,5 the toe starts to damage and when Nod =2 some flattening of the toe has occurred, but the functions remains, thus the damage is acceptable. The stability coefficient, Ns=Hs/ΔDn50, for the toe should not be larger than the value for the armour stone. In order to take the current into account that cause scour at the toe of the structure, the Coastal Engineering Manual (USACE, 2006) prescribes the equation of Smith for toe stability under combined waves and currents:

( )50

ss c

n sc

H U uN aD gh

⎛ ⎞⎛ ⎞ += = ⎜ ⎟⎜ ⎟ ⎜ ⎟Δ⎝ ⎠ ⎝ ⎠ (29)

Where:

; 51,0 26,42

b

s

hgHTu aL h

⎛ ⎞= = −⎜ ⎟

⎝ ⎠ (30)

And:

• u maximum wave orbital velocity in shallow water (m/s)

• U current magnitude (m/s)

• g gravitational acceleration (m/s2)

• hs total water depth (m)

• hb water depth over toe berm (=ht) (m)

• H breaking wave height (m)


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• T wave period (s)

• L local wavelength (m) Equation (29) should be reconsider when (Ns)c > Ns, because the toe is unstable. A larger stoned diameter should then be chosen to find new values for Ns and hb. The maximum current is derived in the boundary conditions to be 1,0 m/s. The structure on the leeward side is exposed to minor waves with high currents in special events of high discharge of the River Araranguá. Therefore it was chosen to design the toe structure on the leeward side in the same fashion as the seaward side. The designs of Figure 39 and Figure 40 are combined in a solution for this jetty system. Bedding layer A bedding layer has the following objectives:

• Prevention of subsoil to penetrate into the rubble mound structure; this can cause settlement

• Distribute structure weight evenly to avoid differential settlement

• Prevent erosion below and close to outer stone layer (undermining)

• Allow water pressure from subsoil to penetrate into structure; bedding layer must have a certain permeability

To perform all these functions, beddings are constructed from granular layers with progressively larger grain size. This is a labour intensive process, which can be replaced by the use of geotextiles. The use of geotextiles can potentially lead to large savings to a project, however, an extensive cost calculation should verify this. Geotextiles perform five basic geotechnical functions (CIRIA, 2007):

• Separation

• Filtration

• Transmission

• Reinforcement

• Protection Mostly used in hydraulic engineering are the functions of filtration and separation. Geotextiles are often placed on beach material, such as fine sand, to prevent the transmission of the sand into the structure, while allowing water to flow through the geotextile. Summarized geotextiles have high permeability with fine filtration and are resistant to the loads of the quarry stone of the structure, which makes them ideal for the use in bedding layers. For the cross sections in this design, the layout of Figure 40 is adopted, where a bedding layer of core material is extended below the toe to provide an evenly distributed pressure on the subsoil. Below that on the seabed a geotextile is laid down. For the fine sand subsoil a filtration of 0,3 mm is needed and a permeability of 1*10‐5 is needed (GEOfabricsLTD, 2009). Further design of the geotextile is beyond the scope of this project. The thickness of the bedding layer is at least 3 kt Dn50 and should not be less than than 0,3 meter to spread the load of the structure over the seabed. Furthermore, as explained above, the bed protection should extend at least 2 m in front of the toe for protection against scour.


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Roundhead design No specific design rules are available for the design of a roundhead. However, the roundhead is the part of the jetty where highest loads are experienced. Therefore, the head of a breakwater is usually reinforced by one of the following measures (Verhagen et al., 2009):

• using larger size armour units

• reducing the slope

• increasing the density of the armour units Of the above measures, the slope reduction is not considered to be a good idea, because it increases the volume of material and interlocking when using tetrapods is generally better when the slope is steeper. This could be a serious option when using armourstone, where a threshold value must be taken into account, but this is not the case here. Using larger units or increasing the volume both relates to an increase in mass. Of these measures an increase in size seems the best option, especially since the armour units are relatively not too large. Increase in size poses a second problem, namely that there are no guidelines available by how much it has to be increased. The present state of research is related to the Hudson formula, as presented above. The Hudson formula relates the mass to a stability number as:

{ }31 , , , cotrD

M f HK

ρ α= ⋅ Δ

The above relation means that a linear relationship exists between a decrease of KD and an increase of mass, assuming all other parameters remain the same. The SPM(CERC, 1984) give tables for KD where the coefficient varies for trunk and head, which for the same situation and considering a slope of 1:1,5 decreases by a factor 1,33 – 1,4. Because no use is made of the Hudson formula, since it is considered outdated, a conservative increase of mass is taken into account for the tetrapods of the head of the breakwater of 1,5. Using this value, the dimensions of the tetrapods are calculated. The radius is another important parameter of the roundhead. The radius follows as a function of the significant wave height, Hs. As with the size of the armour units, there is no design formula for this parameter. A widely adopted relationship is that the radius related to the significant wave height as: R=3 Hs. The radius, R for both roundheads is thus 3*4,2 = 12,6 m. Figure 41 shows where the damage on a roundhead is critical and what this means for the layout of the roundhead. It will be taken into account that at an angle of about 135 degrees with respect to the direction of wave propagation the damage is critical and the roundhead should extend to that side.


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Figure 41: critical area for damage on a roundhead (USACE, 2006: 2243)

Concrete Units When armourstone becomes too heavy, i.e. more than the threshold value of 2,7 tons (Dn50 = 1,0m), concrete units are applied. In this case tetrapods will be used, because of availability (see chapter on trade off). When applying tetrapods, the recommended slope is 1,5 and most design formulas are based on this slope angle. Concrete armour units, and also tetrapods, are mostly made of conventional unreinforced concrete for cost effectiveness. In this case concrete C28/35 is assumed, with a density ρc of 2400 kg/m

3 and a tensile strength of 2 MPa. Furthermore the units are placed in 2 layers, to ensure some interlocking and therefore increase stability of the structure. It must be noted that no design formulas for shallow water are available for tetrapod structural design. Formulas of Van der Meer (1988) and De Jong (1996), which are discussed in the Coastal Engineering Manual (USACE, 2006), the Rock Manual (CIRIA, 2007) and Van der Meer (1995), give stability formula for non depth limited wave conditions, on a slope of 1:1,5. Because of the lack of equations for shallow water and the fact that in this design the tetrapods will be applied in relatively deep water, these relations will be used for design purposes. It is however noted that model testing is required. The equations for stability of the tetrapods are: For surging waves:

0,50,23,75 0,85s od

omn

H N sD N

−⎛ ⎞⎛ ⎞

= +⎜ ⎟⎜ ⎟⎜ ⎟Δ ⎝ ⎠⎝ ⎠ (30)

And for plunging waves:

0,50,28,6 3,94s odom

n

H N sD N

⎛ ⎞⎛ ⎞= +⎜ ⎟⎜ ⎟⎜ ⎟Δ ⎝ ⎠⎝ ⎠ (31)

Where the wave steepness, som, is:


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2

2 som

m

HsgTπ

= (32)

The mass of the unit is related to the volume, V, nominal diameter, Dn and characteristic armour unit length, D, as:

3 3n s

c

MV D k Dρ

= = = (33)

For values of Nod= 0,2‐0,5 corresponds to start of damage, 1 means intermediate damage and a value between 1 and 5 means failure. These are related to damage levels of respectively 0‐5%, 10% and >20%. Furthermore a relation between Sd and Nod exists:

( )1od v dN G n S= − (34)

Where G=1 for concrete units. For tetrapods, nv=1 and above it was described that Sd has a value of 2 for start of damage, which would give Nod=1. This value however, corresponds to intermediate damage. Therefore a value of 0,5 is assumed in this design representing the start of damage. For the dimensions of the tetrapods Figure 42 shows technical drawings. The dimensions are all related to the characteristic armour unit length, D, from equation (33), which is H in Figure 42.

Figure 42: tetrapod dimensions (Verhagen et al., 2009:305)


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Breakage The CEM states: “Slender units are the most vulnerable to cracking and breaking because the limited cross‐sectional areas give rise to relatively large tensile stresses. Many failures of breakwaters armored with Tetrapods and Dolosse were caused by breakage of the units before the hydraulic stability of the unbroken units was exceeded. Much of the damage could have been avoided if design diagrams for concrete armor unit structural integrity had been available during design” (U.S. Army corps of engineers, 2006:VI‐2‐21) For the tetrapods design breakage is taken into account, since the units are sensitive to breaking. For granite armourunits it was assumed that breakage is covered by the introduction of the material safety factor. Burchart (1999) described the formula for breakage of slender concrete units in a practical engineering form as:

31 20

CC CsB C M H S=

(35)

Where for tetrapods:

• B relative breakage

• M armour unit mass (ton)

• S characteristic concrete tensile strength (MPa)

• C0 0,25

• C1 ‐0,79

• C2 ‐2,73

• C3 3,84 It must be noted that equation (35) has a range of validity for the mass of 2,5 £ M £ 50 tons. Neither the CEM, nor Burchart (1999) states which values the breakage should have for a safe design. Therefore the relative breakage, B, is coupled to the Nod value where 0‐5% means start of damage, 10% means intermediate damage and above 10% is intolerable. Then when Nod should be between 0,2 and 0,5, B should be lower than 5% and for intermediate damage B < 10%. Weir sect ion Because at the weir section the structure is overtopped to allow for sediment bypassing, design formulas change. The crest freeboard at the weir section is 0, in other words, the crest of the structure is located at SWL. Research by Van der Meer has shown that a fairly simple relation exists between the reduction factor for the diameter of the armourstone, Dn50, and the crest freeboard, Rc. Figure 43 shows a design graph in which it is shown that for a wave steepness up to 4%, which is always the case, the reduction factor for Dn50 for a crest at SWL is 0,8. Therefore, the weir section will be designed according to design equations used for the rest of the structure taking into account reduction factors of 0,8 for Dn50 and related to that 0,51 for M50.


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Figure 43: Design graph for low crested structures above SWL (Van der Meer, 1995:293)

Transit ions Transitions in rubble mound structures are required when (CIRIA, 2007):

• Orientation of the jetty changes relatively rapidly

• Between different types of armour

• Between different sizes of armour Transitions are needed to avoid a change at the external profile of the structure, which decreases stability. Also at bends, the stability can be affected when the bend faces outward and similar effects as at the roundhead take place. A typical place for a transition is behind the round head. The transition is between concrete armour units and the armourstone behind the round head. The concrete armour units should then be placed first in a transition line of 45° across the side slopes (see Figure 44). Then the armourstones should be placed next to or on top of these concrete armour units. Transition between different sizes of armourstone should be carried out the same way. There the more stable units should be placed first (see Figure 44). Transitions of orientation changes (with or without different armourstone sizes) can be constructed by laying the straight section to the standard placing pattern. This layer terminated at an angle of 45° across the side slope when the same armourstone size is used and ‐45° when a bigger armourstone is used (see Figure 39). Then in the area in between the armour the stones are laid from toe to crest ensuring a good interlock. When using different sizes lighter armourstone is put on top of the heavier armourstone.


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Figure 44: typical transitions in concrete armour unit breakwaters (CIRIA, 2007: figure 6.22)

Transition of the underlayer also requires extra attention. This because the differences of the underlayer can be significantly different due to different armour and underlayer thicknesses and the external profile is kept constant. Calculat ions and results Taking into account all of the above, the stone sizes for the jetty structure have been calculated. The results of stone mass and nominal diameter are presented in Table 16. From the diameters and masses different parameter for the required cross section and foundation can be calculated.

Figure 45shows the layout of the jetty, where critical points at which the cross section is calculated are indicated with numbers 1‐10. Also, a transformed coordinate system is used to describe the location of the different points.


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Figure 45: top view of jetty structure with cross sections The following coordinate transformation is used to come from global to local coordinates and vice versa:

0

0

cos sin 'sin cos '

XX xY Y y

α αα α

⎡ ⎤⎡ ⎤ ⎡ ⎤ ⎡ ⎤= + ⋅⎢ ⎥⎢ ⎥ ⎢ ⎥ ⎢ ⎥−⎣ ⎦ ⎣ ⎦ ⎣ ⎦⎣ ⎦

(36)

Where α is the angle that the shoreline makes w.r.t. the vertical (north) direction. In this case α = 45o. X and Y are the global coordinates of the South American Datum 1969. In Table 13 every point of Figure 45 has two entries, which represent the values for the design storms Hss,1/475 and Hss,1/50. As explained in the previous section mass of the armourstone these different storms have different safety factors, which account for ‘no damage’ and ‘intermediate damage’. Furthermore point 8 has four entries, due to the fact that for the roundhead and for the trunk, different waves are governing. It was assumed, as stated in the boundary conditions, that for the roundhead waves from the SSW are governing and for the trunk, the waves from ENE are governing. Furthermore, to calculate the values of the stone masses and sizes, the standard ratio’s are adopted as used in Figure 40. The standard ratios are:

• Underlayer Mu50 = Ma50/10

• Toe Mu50 = Ma50/10

• Core Mc50 = Ma50/200 ‐ Ma50/4000


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Only when criteria for toe stability are not met, the toe stone size is increased or decreased, until equations (toes) are within the operational limits. The stability of the cross sections consisting of armourstone are calculated with a slope of 1:2. For tetrapods this slope is 1:1,5, as explained in the section on Tetrapods (concrete units).


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Table 13: results of stone size calculation

point x' (m) y' (m) h (m) Hs (m) Tm (s) Sd Nod Nod,toe Ma50 (t)

B (%) Mt50 (kg) Dn50 (m) Dtoe50

(m) Du50 (m)

Stone Tetra Stone Tetra Stone Tetra

2 150 ‐250 1,9 1,8 7,6 5 2,0 0,75 244 0,6 0,3 0,3

2 150 ‐250 1,9 1,8 7,3 2 0,5 1,21,5 164 0,8 0,3 0,4

3 370 ‐50 4,2 2,9 8,1 5 1,0 2,0 2,35 2,53 9,62 230 490 1,0 1,0 0,6 0,5

3 370 ‐50 4,2 2,8 7,8 2 0,5 0,5 3,51,5,6 6,33 52 3501 270 1,1 1,3 0,5 0,6

4 420 0 5,1 3,3 8,3 1,0 2,0 3,0 5,8 10001 1,0 0,7 0,5

4 420 0 5,1 3,2 8,0 0,5 0,5 4,31 4,82 500 1,2 0,6 0,5

5 650 0 7,2 4,2 8,8 1,0 2,0 5,6 1,8 20001 1,3 0,9 0,6

5 650 0 7,2 4,1 8,4 0,5 0,5 6,21 1,8 1200 1,3 0,8 0,6

7 150 150 1,8 2,0 7,0 5 2,0 0,8 174 0,7 0,3 0,3

7 150 150 1,8 1,7 6,9 2 0,5 1,01 124 0,7 0,3 0,3

8 630 150 6,6 4,0 8,7 1,0 2,0 4,9 2,3 16501 1,2 0,9 0,6

8 630 150 6,6 3,9 8,3 0,5 0,5 5,41 2,3 1100 1,3 0,7 0,6

8 630 150 6,6 3,2 7,6 5 1,0 2,0 2,7 2,53 7,2 1100 1700 1,0 1,0 0,9 0,5

8 630 150 6,6 2,7 6,9 2 0,5 0,5 2,71 7,53 4,82 13003 900 1,0 1,4 0,8 0,7

9 390 150 4,5 2,8 7,8 5 2,0 2,0 300 0,9 0,5 0,4

9 390 150 4,5 2,5 7,2 2 0,5 2,51 3501 1,0 0,5 0,5

10 260 ‐150 2,9 2,3 7,8 5 2,0 1,35 130 0,8 0,4 0,4

10 260 ‐150 2,9 2,3 7,6 2 0,5 2,21,5 1501 0,9 0,4 0,4

point Corresponding to Figure 45 Nod,toe damage parameter of the toe structure Du50 Mean nominal diameter of underlayer stone x' X coordinate in coordinate system of Figure 45 Ma50 Mean armour unit mass Notesy' Y coordinate in coordinate system of Figure 45 Stone Armour units granite 1 Governing value for specified pointh Water dept from SWL Tetra Armour units concrete tetrapods 2 Breakage is decisiveHs Design wave height B relative breakage of tetrapods 3 Outside range of validity breakage formula Tm Mean period Mt50 Mean toe stone mass 4 Outside range of validity toe stability formula Sd damage parameter armourstone Dn50 Mean nominal diameter of armourstone 5 Value is halved for submerged part (weir Nod damage parameter tetrapods Dtoe50 Mean nominal diameter of toe stone 6 Larger than treshold value

To incorporate the results of the calculation into the design of the jetty, some discussion of the results is needed. Due to economical and practical considerations, stone sizes cannot be varied to an unlimited extent. Therefore the size ranges of armour stones (and therefore underlayer and toe stones) are limited to a few categories according to the nominal diameter Dn50. As can be seen the largest stone needed is one with Dn50 = 1,0 m and M50 = 2,7 ton. When filling in the results of Table 13 in Figure 45, an overview is created of different stone sizes in different places of the jetty structure. Taking this into account and looking at the results in Table 13, an appropriate division of size ranges would be:

• Dn50 < 0,4 m gives Dn50 = 0,4 m

• 0,4 m < Dn50 £ 0,6 m gives Dn50 = 0,6 m

• 0,6 m < Dn50 £ 0,8 m gives Dn50 = 0,8 m

• 0,8 m < Dn50 £ 1,0 m gives Dn50 = 1,0 m As indicated above, the stone sizes will always be rounded of to the higher standard dimension, because of safety reasons. Stone sizes smaller than the calculated value, could cause instability of the structure. For the tetrapods it can be seen that throughout the cross section of the south arm (point 3‐5) tetrapods are needed, because the armour stone is over the limits of the threshold value. Furthermore, at point 3 the calculations give an unexpected high value for the required mass of the tetrapod. This is because the range of validity of the breakage formula has been crossed. Nevertheless, for safety reasons also this part of the jetty is equipped with the same tetrapods. Because the head needs armourstone of 1,5 times the weight of the trunk, the following 2 diameters are used for tetrapods:

• Dn = 1,4 m trunk

• Dn = 1,6 m head Table 14 shows the dimensions for different tetrapods that have been used (see Figure 42) for nominal diameters of 1,4 and 1,6 meter. Also appendix L shows technical drawings of the tetrapods.

Table 14: Tetrapod dimensions in meters

Dn D A B C D E F G H I J K L

1,40 2,14 0,65 0,32 1,02 1,01 0,50 1,38 0,46 2,14 1,30 0,65 2,33 2,57

1,60 2,45 0,74 0,37 1,17 1,15 0,57 1,58 0,53 2,45 1,48 0,74 2,67 2,94

Both equation (28) and (29) lose their validity when the water becomes too shallow, because the toe height becomes very high. Calculations have shown that at a distance from the coast of 150 m, where for both arms of the jetty a transition is needed, values for the toe stone size that follow from calculation are very small. Therefore from these points shoreward the armour layer is extended to the bedding layer, with no separate toe construction. Variations of the cross section leads to economic savings, but, as stated above, cannot be done to an unlimited extent. Therefore the final dimensions of the various armour‐, underlayer‐ and toe units are presented in Table 15. It can be seen that cross sections remain similar over a certain distance.


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Table 15: Final stone dimensions

Section Start End Armour Dn50 M50 Dt50 Du50 Dc50

1 1 2 Stone 0,8 1,4 ‐ 0,4 0,14

2 2 10 Stone 0,8 1,4 0,4 0,4 0,14

3 10 3 Stone 1,0 2,7 0,6 0,6 0,17

4 3 4 Tetra 1,4 6,6 0,8 0,8 0,23

5 4 5 Tetra 1,4 6,6 1 0,8 0,23

6 RH South Tetra 1,6 9,8 1 0,8 0,26

7 6 7 Stone 1,0 2,7 ‐ 0,4 0,17

8 7 8 Stone 1,0 2,7 0,8 0,6 0,17

9 RH North Tetra 1,6 9,8 1 0,8 0,26

Table 15 describes the following characteristics:

• Section Number of the cross section

• Start Start point of the cross section according to Figure 45

• End End point of the cross section according to Figure 45

• Armour Armour layer stone is granite or tetrapods

• RH Roundhead

It can be seen that the core material varies to a certain extent. Therefore it was chosen to divide the core material in two separate categories, where the material is rounded off to the lowest value, because core material should not be to large:

• Dc50 < 0,2 m gives Dc50 = 0,1 m

• Dc50 ³ 0,2 m gives Dc50 = 0,2 m This gives a bed height of 3 kt Dc50 (³ 0,3m) of either 0,3 or 0,6 meter. The cross section dimensions as calculated using the relations described in the section mass of the armourstone are presented in Table 16. But some of this cross section dimension will be influenced by aspects in the next paragraphs, final cross section dimensions will be shown in the structural overview, the last paragraph of this chapter.

Table 16: Cross section dimensions

Section Bmin α (1:) ta tu Bt 2,9 H ht hbed

1 2,4 2,0 1,6 0,8 4,2 1,9 1,9 0,3

2 2,4 2,0 1,6 0,8 1,2 5,1 2,1 0,3

3 3,0 2,0 2,0 1,2 1,8 1,2 3,0 0,3

4 4,3 1,5 2,9 1,6 2,4 1,6 3,5 0,6

5 4,3 1,5 2,9 1,6 3,0 2,0 7,2 5,2 0,6

6 4,9 1,5 3,3 1,6 3,0 2,0 7,2 5,2 0,6

7 3,0 2,0 2,0 0,8 1,8 1,8 0,3

8 3,0 2,0 2,0 1,2 2,4 1,6 6,6 5,0 0,3

9 4,9 1,5 3,3 1,6 3,0 2,0 6,6 4,6 0,6


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8.2 Overtopping calculation In the functional analysis criteria are set for mean overtopping discharges (liter/s/m). During normal service conditions, which will be represented by a storm with an occurrence of once a year, the maximum overtopping discharge is: 0,1 l/s/m. For extreme design conditions this maximum mean discharge is: 10 l/s/m. Overtopping discharge will determine the freeboard above HDW, from this the crest height of the breakwater will follow. To calculate this freeboard 2 methods are used; a simple empirical calculation tool provided on the overtopping manual internet site and the neural network method (CLASH) provided by Delft Hydraulics (Van Gent et al., 2007). The first method will be used to calculate the first rough crest height, in empirical models simple equations are used to describe the overtopping and structure parameters. With the neural network method the design of the breakwater can be optimized. In this tool you can adjust different lay out or design parameters; implement a berm, widening crest, increasing crest, increasing crest wall. And as the input file has no limitations it is easier to increase one or more parameters to find trends for certain measures. Empirical method For easy use of empirical equations a calculation tool is available on the website of the overtopping manual (HR Wallingford Ltd, 2008) to execute the calculations. The empirical method uses the following equation to a deterministic calculation:

1,031,0 00

0,067 exp 4,3tan

cb m

m m b fm

RqHg H β ν

γ ξξ γ γ γ γα −

−

⎛ ⎞= ⋅ ⋅ − ⋅⎜ ⎟⎜ ⎟⋅ ⋅ ⋅ ⋅ ⋅⋅ ⎝ ⎠

(37)

With a maximum of

300

0, 2 exp 2,3 c

m fm

RqHg H βγ γ

⎛ ⎞= ⋅ − ⋅⎜ ⎟⎜ ⎟⋅ ⋅⋅ ⎝ ⎠ (38)

Where: Rc= Freeboard ξm‐1,0= Breaker parameter γb= Influence factor berm γf= Influence factor roughness elements on the slope γβ= Influence factor of oblique wave attack γv= Influence factor wave wall A probabilistic approach will give less conservative values, to indicate the difference this approach is used too. For a probabilistic approach the following formulas are used:

1,031,0 00

0,067 exp 4,75tan

cb m

m m b fm

RqHg H β ν

γ ξξ γ γ γ γα −

−

⎛ ⎞= ⋅ ⋅ − ⋅⎜ ⎟⎜ ⎟⋅ ⋅ ⋅ ⋅ ⋅⋅ ⎝ ⎠ (39)

With a maximum of:


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300

0, 2 exp 2,6 c

m fm

RqHg H βγ γ

⎛ ⎞= ⋅ − ⋅⎜ ⎟⎜ ⎟⋅ ⋅⋅ ⎝ ⎠

(40)

Empirical calculations are done for an armoured simple slope structure, in this configuration it isn’t possible to include a toe at the bottom of the structure, oblique wave attack or a berm. For the reduction factor of roughness elements a permeable core with two layers of stone is set, giving a value of 0,4. For the slope 1:2 is used, as is in our design for rock armour stones. So in these calculations no distinction is made for cross‐sections where tetrapods or rock are used as armour layer. In Table 17 the input parameters are given for different wave conditions and cross‐sections of the breakwater, it also gives the results of the calculated freeboard.

Table 17: Input and results of empirical method

Input Results

Wave height Hm (m)

Period Tm (s)

Breakwater Overtopping allowed (l/s/m)

Rc

deterministic Rc

probabilistic

H1/475

2,3 7,8 S (mid) 10 2,2 2,0

4,2 8,8 S (head) 10 4,6 4,1

2,3 7,6 N (mid) 10 2,2 2,0

3,3 7,6 N (head) 10 3,5 3,1

H1/1

1,8 6,8 S (mid) 0,1 3,0 2,7

2,9 7,8 S (head) 0,1 5,2 4,6

1,7 7,3 N (mid) 0,1 2,8 2,5

2,0 7,5 N (head) 0,1 3,4 3,0

From the result in Table 17 it can be concluded that the overtopping conditions for the storm once a year are governing for the design crest height. In the next paragraph this will be recalculated and implemented. Neural Network (CLASH) A neural network tool performs the fitting of curves with dimensionless parameters. It is not necessary to define relations between individual parameters with those tools, but because a neural network does not use physical relations the amount of data needs to be large and the quality needs to be good. The neural network tool (CLASH) that is used depends on nearly all existing observations of overtopping, both from model tests as well as from prototype observations, more than 10.000 observations are in the database. A neural network can be trained to obtain more reliable answers, the CLASH tool indicates the reliability of the answers (Verhagen et al., 2009). For the neural network tool 14 different parameters are needed (Figure 46); some are for all configurations the same, some are different for each setting. Assumed is that no berm is used in the breakwater, therefore all parameters regarding the berm are set to zero.

Figure 46: Parameters used for the NN modeling of wave overtopping discharge ((Van Gent et al., 2007)


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Parameters that are the same for every configuration:

• B=0 (width of berm)

• αb=0 (angle of berm)

• Hb=0 (waterdepth above berm)

• Cot αd =1,5/2,0* (down slope of structure)

• Cot αu = 1,5/2,0* (upper slope of the structure)

• γf = 0,38/0,4* (influence parameter elements)

• Rc = Ac** (Freeboard and crest of armour layer)

*slash: first number is valid for tetrapod armour layer, second for rock armour layer. **Means that freeboard is set equal to crest of the armour layer, which indicates that no crown wall is used in the design. In Table 18 the parameters that differ for each configuration or cross‐section are given, these cross‐sections correspond to Figure 45. Not all cross‐sections have to be calculated for overtopping, because for the weir section the crest height is already determined to be at MSL. The cross sections of the roundheads also don’t need to be calculated, because extra overtopping can be allowed at these points. Furthermore the Crest width is determined by construction requirements, as will be explained in section 8,5. The angle of incidence is calculated with Swan One, which takes refraction of the waves into account when approaching the coast. This angle of wave attack is related to the normal of the structure. Note that for the water depth at the toe the Design High Water depth (MSL +1,54m) is used, as explained in the chapter on Design High Water. The wave heights are chosen the most severe for each cross‐section; at the deepest point of each cross‐section.

Table 18: Input parameters CLASH

Cross‐section

Wave height Hm (m)

Period Tm (s)

Water depth h (m)

Angle of incidence

β (°)

Water depth at toe Ht (m)

Width of the toe bt (m)

Freeboard Rc (m)

Crest width Gc (m)

Output qmean

(l/s/m)

H1/475

5 4,2 8,8 8,74 20 6,14 3,0 4,1 10,5 3,080

4 3,3 8,2 6,64 20 4,46 2,4 3,0 10,5 1,953

1 1,8 7,6 3,44 30 3,44 0 1,7 8,7 0,222

8 3,2 7,6 8,14 25 6,24 2,4 3,1 9,1 0,829

7 2,0 7,0 3,34 30 3,34 0 1,7 8,9 0,253

H1/1

5 2,9 7,8 8,74 20 6,14 3,0 4,6 10,5 0,056

4 1,8 6,8 6,64 20 4,44 2,4 3,0 10,5 0,012

1 1,5 6,8 3,44 30 3,44 0 1,7 8,7 0,051

8 2,0 7,5 8,14 25 6,24 2,4 2,5 9,1 0,055

7 1,5 6,8 3,34 30 3,34 0 1,7 8,9 0,048

*Cross‐section 1 and 7 don’t have a toe structure, therefore ht =h and bt=0

In Table 18 also the first results of calculations with CLASH are given, as can be seen the empirical method overestimates the freeboard and design can be optimized. To optimize the design the freeboard parameter (Rc) is decreased with steps of 10 cm. First this is done for the overtopping criteria for normal conditions (H1/1) , then for the newly obtained freeboard parameters it is checked if they still meet overtopping requirement for storm conditions (H1/475). If requirements are not met a new optimization for these conditions is done for this cross‐section. Outputs of the calculation with CLAHS and more explanation is given in Appendix J. The results for freeboard parameters are given in Table 19, these are the values used in the governing cross‐sections.


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Table 19: Final freeboard values

Cross‐section Rc

5 4,24 1,81 1,48 2,27 1,4

Remarks on overtopping calculat ions

• For overtopping criteria the mean overtopping discharge is often used to judge allowable overtopping. However this will never describe the real behavior of wave overtopping. The major part of the overtopping discharge during storm conditions is only from a small number of waves, so the amount of overtopping discharge (m3/s/m) from a single wave can be much larger (100x) than the mean overtopping discharge during the peak of the storm.

• The overtopping manual and the PIANC guidelines present critical values of the average overtopping discharge. The tables and figures are rough guidelines, because even for the same value of q, the intensity of water hitting a specific location depends on the geometry of the structure and the distance from the front of the structure. Maximum intensities might locally be up to two times larger than mean q.

• To reduce cost of the structure one can ask himself whether it is necessary to allow pedestrians to walk on the south arm of the jetty structure, since this means that the weir section needs to be crossed. When more overtopping on the south arm is allowed, less material is needed for construction. Of course, the overtopping criteria can be reconsidered for the entire structure when aiming to save material and thus costs.

8.3 Bearing capacity and settlement It has become apparent that only very little information is available about the soil characteristics at the location of construction. More information can be found in chapter 4, paragraph 3 “Geotechnical data”. In this section Table 9: basic values for calculation give values that are used in these calculations. Bearing capacity calculat ions Calculation of the bearing capacity can be done with different equations. Theory uses the Terzaghi, Prandtl, (USACE, 2004a) Brinch Hansen (Molenaar et al., 2008) and Meyerhof (USACE, 2006: 2383) to calculate the maximum bearing capacity. The Brinch Hansen and the Meyerhof methods are a further development of Prandtl’s and Terzaghi’s methods and are valid for static loading and homogenous soil conditions. The latter are preferred over Prandtl’s and Terzaghi’s theories. Meyerhof and Brinch Hansen use the same equation to determine the maximum soil bearing capacity:

max' ' ' 0,5 'c c c c q q q qp c N s i d q N s i d BN s i dγ γ γ γγ= + + (41)

Because horizontal forces can be neglected on a rubble mound breakwater (decided in collaboration with ir. H.J. Verhagen, Delft University of Technology, department of Hydraulic Engineering) and there is no cohesion, equation 41 can be shortened and the Meyerhof method can be choosen to work with since the vertical force acts centrically on the structure. Below the new shortened equation (42) and an explanation of the factors used in this equation are given.


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max' ' 0,5 'q q qp q N s d BN s dγ γ γγ= + (42)

To calculate the maximum bearing capacity input parameters from the heaviest part of the breakwater, a profile at the end of the NE breakwater, are used:

' 35ϕ = ° 3' 10 /kN mγ = 0,3fD m=

' 3q kPa= 55B m= 455L m=

A detailed calculation of the bearing capacity is given in appendix K. The maximum bearing capacity will be:

max' 10,79p MPa=

The weight is calculated in appendix K:

2122 / 0,12p KN m MPa= =

The bearing capacity is sufficient. Because the safety factor 10,79/0,12=88 is very large more detailed calculations and calculations for other cross sections are not necessary. Calculat ion of sett lement There are a few considerations on which the calculation of the settlement in this situation is based. Settlement occurs very rapidly, there will be no primary settlement or creep and the settlement is presumed to be small. The information in this chapter is derived from the US Design Guide Lines: 06b – Settlement in sand and bearing capacity (USACE, 2004a). The Schmertmann Strain Factor Method (CPT) (43) has been chosen to calculate the initial settlement because it’s a widely used method to estimate settlement under the centre of a footing. The result could be conservative and the effect could be negated due to disturbance of the soil during the construction process. Still the Schmertmann method is the best method available in this situation. Schmertmann Strain Factor Method:

1 21

nz

ii i

IC C zE

ρ σ=

⎛ ⎞= Δ Δ⎜ ⎟⎝ ⎠

∑ (43)

In order to get a result from equation (43) some further assumptions have to be made. Interpolation of the

strain influence distribution is needed because L/B is never greater than 10. 60/cq N is set at 5, so E becomes

175 MPa. Again we calculate for the heaviest part of the structure at the end of the SW breakwater. The input parameters:

Length: 455L m=

Width: 55B m=

Effective weight: 3' 8 /s KN mγ =

Depth of embedment 0.3fD m=


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Time in years after placement of footing: 50yearst year=

Pressure at foundation 2122 /q KN m=

A detailed settlement calculated is made in Appendix K:

0.27mρ =

The settlement of 27cm calculated above is calculated for the heaviest part of the structure. Also the US Design Guide Lines (2004) state that this could be a conservative value. Because it is expected that no differential settlement will take place, the structural integrity is not being compromised. For overtopping the breakwater should be designed 27cm higher. It is advised to do this over the entire structure. When it is necessary to no know the settlement on different points on the breakwater some more detailed calculations can be done.

8.4 Rock quantities Cost and logistics are very much dependent on the quantity of rock and tetrapods that the jetty needs. Therefore an estimation on the quantity of the different rock sizes and tetrapods have been made. In Table 20 and Table 24 you can see the results. It must be noted that these results are averaged. The values in Table 21 should be multiplied with a certain factor to take losses and such in consideration.

Table 20: Tetrapod quantities

D50 # concrete

needed (ton) Tetrapod 1.4 5290 34860 Tetrapod 1.6 722 7067

6011 41928

Table 21: Rock quantities

D50 porosity V (m3) Rocks needed

(ton)

Armourlayer 0.8 0.35 10211 17588

1.0 0.35 67267 115868

Underlayer

0.4 0.4 6853 10896

0.6 0.4 23911 38018

0.8 0.4 18462 29354

Toe

0.4 0.4 668 1061

0.6 0.4 1502 2388 0.8 0.4 9290 14771

1.0 0.4 6000 9540

Core 0.1 0.42 162210 249317

306374 488803

Compared to the structures designed by INHP (INHP, 1993) there is a reduction of the use in rocks. INHP choose alternative 3 for a final design (see Table 22). The difference can be explained because the INHP design doesn’t use tetrapods and in the present design the inland part (from the shoreline inward) of the breakwater is not taken into consideration.


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Table 22: Comparison of the used amount of rock

Total amount of rock (m3)

Present design 306374 INHP Alternative 1 450111 INHP Alternative 2 326543 INHP Alternative 3 359583

8.5 Construction aspects

The actual construction of the breakwater is not a part of this project, but some construction aspects will influence the design of the jetty. Therefore a few of these aspects have been taken out to take into consideration during the project. In this chapter the focus lies on two aspects: construction equipment and construction methods. There are two ways to build a breakwater; land based and water based. Both methods require different working methods and can produce different design problems. Land based construction is preferred over water based construction because of economical reasons; dumping of bulk material and controlled placement. Land‐based equipment is in general cheaper than working with waterborne equipment, only when working with land‐based material is unpractical or impossible waterborne equipment is chosen. In our case probably land‐based equipment or a combination of land based and water based construction will be used for construction. The breakwater has to be build under exposed conditions. The construction method should ensure that damage from wave attack is minimal during the construction and that exposed under layers are covered by protecting layers as soon as possible. A common distance to work with is 25 – 50m, dependant of the working method and the degree of exposure. Land based construct ion Land based construction will probably be the best way to construct the breakwater. In this paragraph it is explained what methods and equipment could used and which design issues will occur. Construction method A typical construction method for land based construction (CIRIA, 2007) is:

1. Placing of quarry run core by dump trucks 2. Placing remainder of core by crane 3. Placing scour protection with a crane 4. Placing of under layer by crane 5. Placing of toe at seaward side (and other side) by crane 6. Placing of armour layer by crane

The placing of the quarry run core is most economic to be done with dump trucks, but to achieve a good grading and the right slope a crane or excavator has to be used. The scour protection is laid directly after the laying the core layer. Scour could occur before the scour protection is laid, therefore it could be wise to dump more material or it can be prevented by dumping stones before the core material is laid with a stone‐dumping vessel. Crest width and slope of breakwater The crest width (of the core) depends partly on equipment that is used, traffic of two trucks or a truck and a crane will determine this width. Other factors are overtopping and the fact that minimal three to four armour stones or prefab elements have to fit next to each other on the top of the core.


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For working on the breakwater we assume that only highway dump trucks are used or articulated dump trucks, because the use of off‐highway trucks on the breakwater will give very large dimensions and will probably if necessary at all only be for short time. If necessary the jetty can be used as a one‐way lane for off‐highway trucks at some moment. Governing operating width for trucks is set to 3,0 meters, by combining operating width given for these two types trucks in the rock manual (CIRIA, 2007: 1087‐1088). For cranes the dimensions of a small grab or wire‐rope crane are chosen, because the armour rock is not too heavy (see chapter 4). This operating width is then set to 4,50 meters, given in the same section of the rock manual. The governing combination that determines the total crest width of the core is a crane and a truck together; this gives a crest width of 7,50 meter. This also explains the values for the crest width in Table 18. The slope choice should also be regarded. Flat slopes and wised berms may exceed the reach of the crane, but the slope of 1:1,5 can be regarded as steep. Land‐based equipment Trucks are mostly used for direct dumping and cranes for individual placing of larger stones. Depending on the size of quarry and volume, dump trucks usually carry between 20‐50t quarry. There are two types of dump trucks, highway and off‐highway. Off‐highway trucks are suitable for heavy loads and heavy armourstones (up to 300 ton) over rough terrain. Trucks are not capable of driving over armourstone, therefore small material is used to fill in between. It should also be considered that short breakwaters are more suitable for land based operations due to logistic problems. Cranes or excavators are thus used for individual placing. Hydraulic excavators are mostly used for cyclic placing of small amounts of armourstone, because of their quick duty cycle. Wire‐rope cranes are used for heavy stones and stones that require placing at greater reach. For individual placing of stones besides these wire‐rope cranes also grab and chain slings are used. Which is used depends on safety assessment and contractor. Water based construction Constructing the weir jetty could be impossible from land because trucks and cranes can not drive or be placed on top of the weir jetty. For this reason the weir and the part of the breakwater seaside of the weir have to be constructed from the water. In this paragraph it is explained what methods and equipment could be used and which design issues will occur. Construction method A typical construction method for water based construction (CIRIA, 2007) is:

1. Placing the scour protection with side stone dumping vessel 2. Placing the quarry run with split hopper barge 3. Trimming of slopes and placing of under layer by floating hydraulic excavator or floating crane 4. Pacing of toe with side stone dumping vessel or floating crane 5. Placing of armour layer by floating crane

The crest of a breakwater constructed from the water can be determined for hydraulic stability and doesn’t need to be wide because of trucks driving on top of the breakwater. The core level should preferably be 3m below sea level due to minimal depth needed for dumping the material. When the core should be higher than 3m below sea level a combination of land based and water based construction is an option. Also the working area should be bigger than working land based for sufficient anchor spaces an maneuverability for the vessels. Water based equipment Floating cranes can be used in combination with barges to dump the core material, but must be sheltered against waves and currents and are therefore not a good choice. The core material is best placed with side


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dumping vessels or split hopper barges. These vessel have a small depth and can therefore dump material up to 3m below the sea level. Side dumping vessels are preferred over split hopper barges because of a greater accuracy. When using a split hopper barge the finishing work has to be done with a side dumping barge or a crane. For water based equipment water depth, currents and waves are important factors affecting the overall downtime during construction. Tidal elevation could also alter the downtime. It should be considered that a loading terminal should be in the near vicinity to load the dumping vessels. In contrary to land based work the length of the breakwater doesn’t form a limitation for logistics.

8.6 Structural Overview In this chapter different calculations are done which determine the total dimensions of the cross‐section of the jetties. Logically most are determined in the first paragraph on cross‐sectional dimensions, but the following chapters also influence some aspects of the cross‐section. Overtopping and settlement will determine the final crest height of the structure and the construction aspects determine the final crest width of the core. In Table 14 (section 6,1) it can be seen that the width of the crest, Bmin is smaller than the width of the core that is needed to transport the stones and construct the jetty. As described in section 8,5 the minimum crest width of the core is 7,5 m. The crest width is calculated by:

( )2( ) tan 2core u aB B t t α= + + ⋅ (44)

And for the roundhead:

2 6roundhead sB R H= = ⋅ (45)

The feeboard calculated in the section on overtopping and settlement together determine the final crest height at each cross section. Following parameters are added: Water depth in front of toe structure, relative to MSL/SWL: h Design High Water (DHW): 1,54 m Freeboard: Rc Settlement: 0,27 m The final crest height of each cross‐section is: h + Rc+1,8 m Relative to MSL this is: Rc + 1, 7m IBGE In Table 23 the total en final cross sectional dimensions are summarized. In Appendix N all drawings of these cross‐sections can be found. Al heights are measured from MSL, except hbed, which is measured from the seabed.


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Table 23: Final Cross section dimensions

Section H Rc B α (1:) ta tu ht Bt htoe hbed

1 1,9 3,2 8,7 2,0 1,6 0,8 1,9 0,3

2 2,9 0,3 8,7 2,0 1,6 0,8 2,1 1,2 0,8 0,3

3 4,2 0,3 9,1 2,0 2,0 1,2 3,0 1,8 1,2 0,3

4 5,1 3,6 10,5 1,5 2,9 1,6 3,5 2,4 1,6 0,6

5 7,2 6 10,5 1,5 2,9 1,6 5,2 3,0 2,0 0,6

6 7,2 6 25,2 1,5 3,3 1,6 5,2 3,0 2,0 0,6

7 1,8 3,2 8,9 2,0 2,0 0,8 1,8 0,3

8 6,6 4 9,1 2,0 2,0 1,2 5,0 2,4 1,6 0,3

9 6,6 4 25,2 1,5 3,3 1,6 4,6 3,0 2,0 0,6


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9. Conclusions The objective of this design project was to design a jetty structure in the river mouth of Rio Araranguá. The result will be summarized in this chapter.

9.1 Functional design The conclusions on functional design will explain what the dimensions of the jetty lay out are and what choices were made. In Figure 47 the jetty lay out is drawn. More detailed drawings can be found in Appendix N.

Figure 47: Lay out dimensions


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Figure 48: Cross section locations

Channel width and depth The channel width is determined calculating the width for navigational and morphological purposes. The width of the channel is set at 120m. The channel depth is mainly configured using the minimum depth for vessels and was used as a design parameter in the morphological approach to determine the channel width. The channel depth is set at ‐6m IBGE. Jetty length The jetty length is chosen by comparing arguments for a long/shorter breakwater. Costs, sediment intrusion, sheltering from breaking waves, longshore current, ships manoeuvrability and the interruption of the longshore transport causing erosion and sedimentation up‐ and downstream play an important role in making the decision. A modification on the previous design is done on the jetty length resulting from the implementation of the deposition basin in the weir section. Because of the placing of the basin waves could penetrate much further in the jetty channel, which affects sediment deposition and navigation negatively. By extension this wave penetration is reduced. The length of the NE breakwater is extended up to 630m in seaward direction to a depth of 6,6m. The SW breakwater stays the same and will extend 650m into the ocean to a depth of 7,2m. Jetty orientat ion The jetty orientation depends on shipping manoeuvrability and implementing the sediment bypass system. The shipping manoeuvrability depends on angle of incidence and height of the waves, the cross currents and wind strength and angle of incidence. The implementation of the sediment bypass system in the jetty lay out depends on the width of the breaker zone and on the size and the positioning of the deposition basin. Size and position of the deposition basin depend on the seasonal net sediment transport rates from south to north, annual dredging operation cycles and available equipment. Dredging operation should start when the basin is completely filled, which is in the month of August. The width of the breaker zone determines the length of the offset and weir section. The inland breakwater is necessary to guide the river flow in the right direction.


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The offset at the shoreline is placed 250m to the SW of the original position and has a length of 150m from the coast. The weir section starts at the seaward end of the offset, is placed under an angle and has a length of 300m. For details see the drawings in Appendix N and figure above.

9.2 Structural design The cross sectional parameters have been defined according to a number of different analysis. The crest freeboard is determined by overtopping requirements and 0,3 meter is added because of post construction settlement. The crest width depends on construction considerations. All other dimensions are directly related to the stone size and weight, which are dependant on the design wave heights. The final result is a statically stable rubble mound breakwater with granite as construction material. In sections where the required rock exceeded the threshold value, tetrapods have been applied. The slope angle varies from 1:1,5 for tetrapods and 1:2 for armourstone. The final dimensions of the cross sections that are designed during the project are summarized in Table 24 and Table 25 presents the stone sizes. In Appendix N all drawings of these cross‐sections can be found. All heights are measured from MSL and units are meters. Units for the weight are in tons.

Table 24: Final cross section dimensions

Section h Rc B α (1:) ta tu ht Bt htoe hbed

1 1,9 3,2 8,7 2,0 1,6 0,8 1,9 0,3

2 2,9 0,3 8,7 2,0 1,6 0,8 2,1 1,2 0,8 0,3

3 4,2 0,3 9,1 2,0 2,0 1,2 3,0 1,8 1,2 0,3

4 5,1 3,6 10,5 1,5 2,9 1,6 3,5 2,4 1,6 0,6

5 7,2 6 10,5 1,5 2,9 1,6 5,2 3,0 2,0 0,6

6 7,2 6 25,2 1,5 3,3 1,6 5,2 3,0 2,0 0,6

7 1,8 3,2 8,9 2,0 2,0 0,8 1,8 0,3

8 6,6 4 9,1 2,0 2,0 1,2 5,0 2,4 1,6 0,3

9 6,6 4 25,2 1,5 3,3 1,6 4,6 3,0 2,0 0,6

Table 25: Stone dimensions

Section Armour Ma50 Dn50 Mu50 Du50 Mt50 Dt50 Dc50

1 Stone 1,4 0,8 0,2 0,4 0,1

2 Stone 1,4 0,8 0,2 0,4 0,2 0,4 0,1

3 Stone 2,7 1,0 0,6 0,6 0,6 0,6 0,1

4 Tetra 6,6 1,4 1,4 0,8 1,4 0,8 0,2

5 Tetra 6,6 1,4 1,4 0,8 2,7 1 0,2

6 Tetra 9,8 1,6 1,4 0,8 2,7 1 0,2

7 Stone 2,7 1,0 0,2 0,4 0,1

8 Stone 2,7 1,0 0,6 0,6 1,4 0,8 0,1

9 Tetra 9,8 1,6 1,4 0,8 2,7 1 0,2


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10. Recommendations The parts of the jetty design that will need more attention will be summed up in this chapter. Some recommendations came forth from the lack of time, lack of data, lack of literature and/or lack of knowledge. These recommendations are summed up in the same order the subjects were put in the report. Therefore the recommendations are grouped by “boundary conditions”, “functional design” and “structural design”. Because of its importance an extra groups was inserted: “costs”. There are some main themes returning in the recommendations:

• 3D or physical modelling has not been done, but is necessary for multiple aspects of the jetty design.

• Some data sets are not available and should be acquired

• Costs have not been calculated but is only reasoned with on a qualitative basis. It is very important that this is done and used as a design parameter in all the decisions that are made.

• Further and more detailed research has to be done in order to obtain a structurally safe jetty system.

10.1 Boundary conditions Recommendations made on the boundary conditions are given below. Soil data should be acquired No information about the soil was available. Therefore assumptions were made that the soil consists of close packed fine sand only, i.e. no layers. At least a Cone Penetration Test (CPT), but probably also other geotechnical research at the construction site will be necessary to determine the properties of the soil, soil materials and layers. By analyzing the results of the CPT the best construction method for the foundation can be determined. 3d modelling should be done for a large variety of input parameters and purposes Wave propagation should be modelled in a 3D program with a large variety of currents, wave heights, periods, direction and water levels in order to predict the wave height in the navigation channel and on the structure. 1D modelling programs as SwanOne (TUDelft) don’t take diffraction and such into account. 1D modelling programs also give a less precise approximation of the wave characteristics because 1D projection is made from a 3D situation. A variety of waves have to be modelled because design methods require different wave parameters. 3D modelling for sediment bypassing, shoaling, erosion, currents and scour will give a better output then when design equations are used. To determine the best lay out different alternatives should be modelled with numerical model simulations. More research should be done on the location of the jetty. Modelling done by CPE(2009) showed that erosion will occur in the channel adjacent to the shoreline north of the present position of the designed breakwater. This could be a more natural position for the jetty system because of the morphological indications. Further research will have to prove this. More detailed wind blown sediment transport data and calculations should be acquired/made. Dune migration by wind will bring sediment into the channel and sediment basin. Calculations should be done to verify the total amount of this transport. If the amount is of any important value, this will effect sedimentation rate of the channel and therefore the amount of dredging that has to be done. It could also cause dune degradation.


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The environmental considerations should be studied In the chapter on boundary conditions environmental considerations are described; salt intrusion, dune migration, wildlife, vegetation and erosion. These environmental considerations should be studied in a later stage, when considered as important issues.

10.2 Functional design Recommendations made on the functional design are given below. Specific recommendations on navigation, bypass system and river equilibrium are given below. Navigat ion The design vessel characteristics should be determined with more care Smaller design vessels could result in a smaller cross section of the navigation channel and therefore reduce (dredging) costs. The downtime for shipping should be determined with more care A larger downtime for shipping can be chosen to establish a smaller depth of the river and will allow higher waves in the channel. Therefore the jetty can be designed smaller and will therefore reduce costs. Bypass system The functioning of the weir jetty system has to be simulated with a numerical model This has to be done to investigate sediment transport pathways and quantities along the structure. This will give more insight in active storage and deposition basin storage capacities, accretion, erosion and sand accumulation patterns. When doing this investigation also alternative lay out should be taken in consideration. The interaction of ebb flow currents to the weir section has to be further investigated. If strong ebb flows are redirected towards the weir, this can cause dispersal of sediment in the deposition basin. The decision for the most appropriate dredging equipment and operation cycles will have to be based on a cost calculation. Vessel supply, production rate and efficiency are the most important input parameters for cost calculation. Conceptual design of weir jetty sediment bypass systems should be done using literature from Weggel as a guideline Weggel is one of the few researchers with extensive experience and knowledge on the design aspects of weir jetties. Using this literature, some of the design choices can be verified. River equi l ibrium The calculations on river equilibrium give just an overview of what will happen to the long term river equilibrium. This can only be used as a preliminary estimation. Modelling on the river should be done to get better insight in what will happen in reality (from 1D to 3D modelling). To many assumptions have been made in this calculation:

• The bed slope for Rio Araranguá is assumptive, because no data was available. Especially the bed slope has influence in the calculations on the shortening of the river reach. Real data should be acquired

• Interaction between both shortening and narrowing is not taken into account, but should be studied.


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• The influence of the deposition basin is neglected, but should be considered.

10.3 Structural design Recommendations made on the structural design are displayed below. First general recommendations are given, later on specific recommendations on the foundation and overtopping are given. Physical modelling should be done As stated extensively in the literature, physical model testing is always required for the design of a rubble mound breakwater. The reason is that all formulas are empirical and have their own field of application. The design equations that are described in the Rock manual, CEM and other literature are for conceptual design purposes only. Therefore no guarantees on the structural stability can be given jet. Uncertainties in design equations which need physical modelling are in the fields of:

• Oblique waves

• Roundhead design

• Structural stability

• Toe stability

Scour of the beach adjacent to the deposition basin The strip of beach between the two parts of the southwest breakwater could be subject to scour. Modelling should point out if the penetrated waves over the weir or the turbulence caused by the fluvial current is necessary. Another option is to do nothing and monitor the erosion. If there is too much erosion protection can be placed afterwards. This can be justified by the low impact of the beach erosion and the relative small protection needed. Armourstone shapes have to be determined more specific For the armourstone the assumption was made that the stone is rough angular. More information about the specific shape (roundness, blockiness etc.) could enhance reliability of the design equations. Overtopping Maximum overtopping volumes have to be calculated and physical modelling has to be done Although a reliability level is given in addition to the predictions of CLASH, these reliability levels do not account for most of these influences. Therefore, the Neural Network predictions should only be used as first estimates of mean overtopping discharges. For final design also maximal overtopping volumes have to be calculated and physical modelling has to be done to determine mean overtopping discharges. Physical modelling has proved to be a reliable method to estimate this. Wind influence should be taken in to account The blowing up, change of shape and angle of incidence of the waves and the modifying of the physical form of overtopping volume can make quite some difference in the mean overtopping discharge. Foundation The maximum slope of the bottom next to the breakwater to the deposition basin has to be determined. In order to make sure that no sliding will occur a maximum value for the slope of the bottom next to the weir and breakwater should be calculated. The occurrence of differential settlement has to be checked


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Designing the foundation no differential settlement is assumed to occur. Even though rubble mound breakwaters are relatively insensitive to differential settlement (i.e.: stability is not directly affected by differential settlement), this is an uncertain assumption especially due to eccentrically loading on the foundation. The choice of geotextile as filter layer should be further investigated The use of geotextiles was assumed to be a good choice. In cooperation with a supplier, this could be researched and an optimal solution for the geotextile can be made During the design the construction methods should be considered. Better or cheaper construction methods could change the design. For example: Construction from landside or seaside could influence the jetty design and costs significantly. Further research is needed on this subject.

10.4 Costs During the design of the jetty decisions about costs were made by qualitative judgement rather than by calculating. No knowledge of cost estimation was present. Trade offs could be dealt with better when cost estimates are available. The most important choices that were made which could influence the total cost of the project are summed up below An economical choice of the jetty lay out has to be made Choices made during the functional design are very much dependant on costs. For the jetty system, it holds that the further the jetties extend seaward, the less sedimentation enters the channel. Less sedimentation means less dredging operations are needed and therefore costs are lower. However, a longer jetty means that more material will be used for construction and this increases the costs. An economical optimum should be found for this problem. An economical choice of armour layer of the breakwater should be further investigated Economical considerations were the use of tetrapods and granite rock. In general granite, when available, is always the more economic choice over concrete armourstones. However, the threshold value of the granite was very uncertain. It was assumed that due to cost considerations armourstone with a larger diameter than 1,0 m is not available and tetrapods or slope reduction would solve this problem. Tetrapods are however considered outdated and have only been used because of the nearby construction company that produces tetrapods. A study on the costs of tetrapods, larger armourstone and other concrete armour layer options should be done to be certain of the best of all options. Slope reduction was considered unfavourable. Material‐, labour‐, transportation‐ and therefore construction costs are higher when reducing the slope. However, it was beyond the scope of this project to examine these costs. For cost reduction in the final design, some requirements can be reconsidered The construction of the jetties at river Araranguá is subject to a very limited budget, because it will be funded by the Brazilian government. Therefore, with a reduction in materials in the final design, the proposal for the jetty system will be more to be likely to be built by the government. It is possible to reduce the cost of the construction, by reducing the amount of material needed. For the reasons explained above, a cost reduction is an important issue for further research on this jetty design. Some measures for cost reduction can be:

• Safety margins can be reconsidered, which may lead to a lower volume of rock required. Not many (or none) jetties in Brazil are constructed according to Dutch safety standards.


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• Also only navigation of smaller boats can be considered, which leads to a shallower channel and therefore the jetties have to be extended less far into the ocean. Fishing boats will normally be smaller than considered in the requirements.

• In the same fashion, overtopping requirements can be reconsidered, which leads to a lower construction height and less material. More overtopping can be allowed, which results in less ability to walk over the structure. Furthermore for the south arm of the jetty, the overtopping can be higher because people have to walk over the weir section to get to the south arm, which is not allowed.


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11. Appendices

• Appendix A: Fieldtrip report

• Appendix B: Project Approach

• Appendix C: Incidental loads

• Appendix D: Tide table

• Appendix E: Currents

• Appendix F: Sediment transport analysis

• Appendix G: Determining design wave height

• Appendix H: Channel cross‐sectional area versus tidal prism

• Appendix I: Overtopping outputs

• Appendix J: Bearing capacity and settlement

• Appendix K: Wave propagation

• Appendix L: Alternative lay outs

• Appendix N: AutoCad drawings cross‐sections


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Appendix A: Fieldtrip Report Will be added in the hard copy report.


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Appendix B: Project approach In this project we will use a structured approach according to the Systems Engineering (SE) theory. A very important part of this is the technical project plan, or project approach. This plan is used as a guideline for the structure throughout the entire project. The plan is used to define the purpose and the method of the systems engineering effort.

In this chapter, an introduction is given on the project and the SE approach to the project. Furthermore a description is given of the engineering approach that will be used. The SE methods that will be used in the project are presented, as well as an organizational structure. The time schedule of the project is also explained in this chapter. Project descript ion and approach In a period of eight weeks we will make a design for a jetty at the Ararangua river in the south of Brazil. A jetty that will increase the hydraulic efficiency of the river and stops the river mouth to migrate is needed to prevent the upstream hinterland from floods. More background information can be found in chapter 1. The main goal of the project is to design a jetty at Ararangua river to prevent flooding upstream. A design synthesis will be used to come to an optimal design for the jetty at given the specified requirements. Starting point is the current state of technology from which a jetty design, suited for our project environment will be designed. On beforehand, we know that more alternative designs can be a solution to our design problem. Decisions have to be taken for the method of sediment bypass and the layout and structure of the jetty. The method for design has several steps, which will be explained in more detail later on in this section. The first step is to make a detailed project plan, gather background information and derive all the requirements for the jetty. After this a functional analysis will be conducted. Before designing anything, the boundary conditions and loads on the jetty have to be estimated. A first design stage will be used to do some global dimensioning of different alternatives. A trade‐off will decide on the desired functional design of the jetty. The detailed design phase is the last phase of the project in which we will come to the final structural design of the jetty. Project stages As stated above several steps have to be taken to come to a jetty design in a structured way. In this section all the phases will be explained in more detail. Project Plan The project plan phase is the first stage of the design process. In this stage all information needed to define the scope of the project is gathered. Furthermore the approach to the project is established and all technical activities have to be identified. The technical activities will be documented in a work breakdown structure. The purpose of the project plan is to make sure that all activities needed to come to a final design are identified and managed. A timeframe is attached to every activity, in order to keep track on the development of the project. Another part of the plan phase is the requirement analysis, which is used to find all requirements on the jetty system as a whole. As an input the demands from the client, in this case CPE, can be used. Other inputs are the project constraints. The requirements following from this process state what the system has to do as well as the limitations of the system. Documentation research is a part of the plan phase as well. Background documentation on the project situation is looked up, as well as technical documentation on the design of a jetty. Also datasets are looked up, which will provide the input for our design calculations. In order to use all documentation in a structured and efficient way, the documentation is listed and ordered.


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The final output of the plan phase is a baseline report. This report covers background information, the project plan with approach strategy, the requirements, a work breakdown structure and an overview of the documentation that is to be used during the project. Functional Analysis In the functional analysis the requirements are allocated to elements of the jetty. The goal is to quantify all requirements on every element of the jetty in terms of functional and performance criteria. In this way a concrete goal is set for the calculation in the design stage. The functional analysis is used to define exactly what every element of the jetty must do and how well it has to perform. Concept design The concept design will be used to generate different alternatives for the jetty design. There are three different aspects that have to be covered in this stage. First of all the boundary conditions have to be determined, to define the inputs to our project. Related to the boundary conditions are the loads, which have to be estimated during this stage. An exact calculation cannot be made at this point, because the design of the jetty will influence the loads to some extent.

Most important of the concept design phase is the generation of different functional designs for the jetty. Different alternatives for the sediment by‐pass system and the layout of the jetty have to be generated. Outcomes for both sub‐systems will influence the channel design, for which only rough estimations will be made.

At the end of the concept design phase, the alternatives must be worked out and dimensioned in such a way that it can be compared in a trade off. The general dimensions have to be estimated and other considerations, useful for trade off, must be worked out. During the concept design stage we will have to find out which elements are crucial to conduct a trade study. Trade off The trade off is used to pick the best design for our jetty system. This design can be worked out in detail in de detailed design stage. The objective of doing a trade off is that all design decisions have been made and the detailed design phase will only be used to optimize the jetty design. The method for trade off, as well as the criteria, will be determined in a later stage of the project. This is done because they depend largely on the outcomes of the concept design phase.

The outcome of the trade off will be the most suitable type of sand bypass system, type of construction (rubble mound, caisson, etc.) and orientation of the jetty. With these outcomes, we will be able to design the jetty in a way that it will meet all requirements. Detailed design In the detailed design the final dimensions of the jetty will be determined. All elements will be calculated and optimized and drawings will be made for clarification. The objective is to have a design that is ready to be built. The final result will be a jetty design that meets all requirements as derived from the requirement analysis. Report writing The concluding stage of this project is the report writing and presentation at CPE in Florianopolis. The report has to cover the complete design process. All decisions that have been taken should be included and understandable for a reader that was not involved in the project. Furthermore it should present a clear result of the detailed design of the jetty.

Presentations will have to be given to CPE, some day in the last week of our stay in Brazil and to the sponsors in the Netherlands.


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Work breakdown structure The Work Breakdown Structure (WBS) is a means of organizing development activities needed for the jetty, based on system decompositions. During a brainstorm session the system is decomposed into subsystems and finally into elements. The purpose is to define all design and research activities that are needed to come to a final design. The output of this process is a hierarchical tree‐like structure that lists all elements and data which need to be designed or researched. The entire system, the jetty system, is decomposed into four topics:

• boundary conditions

• loads on the jetty

• structural design

• functional design These topics define the entire system of the jetty and all research and calculation needed to come to a solution to our design problem. A further decomposition is done to define all parts and elements of the system. The result is a complete and comprehensive overview of the system that is to be designed. The WBS is presented in the appendix and in Figure 49.

Figure 49: Work breakdown structure

The WBS is not only used to present an overview of all products and processes, but is also an aid in de construction of the time schedule. All activities in the WBS are represented in the planning, to make it as clear and complete as possible. Organizat ional structure The group consists of four Master students, of which three follow the Hydraulic Engineering specialization (Roderik, Maarten en Ilse) and one the Building Engineering specialization (Wim). The work is roughly divided in two parts, the technical and the process part. For both parts a separate allocation of functions is presented below in a table.


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A schematic of the technical issues that need to be resolved, a work breakdown structure (WBS) has been developed as depicted chapter 3. This WBS gives an overview of all elements of the jetty system that have to be designed or researched. In our planning (see below) every technical activity in the design process is allocated to a team member. Technical function allocation In every design stage of the project, several technical issues will be addressed. Efficiency is increased by giving every group member his/her own technical field of specialization. Throughout the concept and detailed design this person will be responsible for his/her own specialization. Not every element of the design can be subscribed to someone. However, during the process we will constantly update our planning and allocate tasks for an optimal efficiency in the design process.

Technical specialization Description Group member

Boundary conditions Hydraulic and meteorological conditions Maarten, Ilse

Other boundary conditions Wim, Roderik

Geotechnical data Wim, Maarten, Roderik

Loads Wave loads Maarten, Ilse

Current loads Wim, Roderik

Other loads Wim

Structural Design

Jetty structure design All

Foundation Design Wim, Roderik

Material types and size Wim, Roderik

Functional Design Channel design Maarten, Ilse

Jetty lay out Maarten, Ilse

Sand bypass system Roderik

Process function allocation To structure the design process each of the group members will have to manage a part of the design process. In this way, the SE approach can be used to an optimal extent. Apart from that the information and all written documents are organized in a structured way. This makes sure information is easily accessible and research will take the least possible effort. The following functions are specified:

Process function allocation Description Group member

Chairman Organize meetings and progress control Wim

Systems Engineering Manage the systems engineering process Wim

External contacts Central contact for sponsors, TU Delft, CPE and Univali

Roderik

Documentation Manage documentation and information sources Maarten

Editor Control report writing and meeting notes Ilse

Timeline and schedule The time schedule of this project is depicted in the project planning. The schedule lists all items that are present in the work breakdown structure, to make sure that no steps will be forgotten. It must be noted that the planning is never a final version. It needs to be updated at least at the beginning of every stage of the project. Furthermore it has to be controlled and adapted regularly and according to the development of the process.


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The technical tasks are allocated according to the technical specialization listed above. The process functions are not present in the time schedule. Tasks related to these functions will need some attention throughout the entire project and need therefore no special attention in the schedule.

It can be seen that the time schedule depicts every different stage of the project and allows for delays. Deadlines for different stages are indicated to assure progress throughout the project and to structure the development. The aim is to finish the final report before we leave Univali. However, it must be noted that minor changes can be made afterwards. Also, preparation and execution of presentations at the TU Delft and to our sponsors will take some time after the project.

The time planning is added in this appendix.


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Project planning Rio Araranguá

Activiteit Week 1 31/08 ‐ Week 2 07/09 ‐ Week 3 14/09 ‐ 18/09 Week 4 21/09 ‐ Week 5 28/09 ‐ Week 6 05/10 ‐ Week 7 12/10 ‐ Week 8 19/10 ‐ M T W T F M T W T F M T W T FR M T W T F M T W T F M T W T F M T W T F M T W T F

Travel to Brasil All Intro Univali All Instruction CPE AlProject Plan A Al ‐ Systems Engineering W W ‐ Description of situation I I ‐ Requirement analysis W All ‐ Investigation of sediment R R ‐ Documentation research M M ‐ Work breakdown W ‐ Project planning W All ‐ Baseline report writing AlFunctional Analysis All ‐ Allocation of AllConcept design All ‐ Hydraulic and M, ‐ Other boundary W, ‐ Geotechnical data W, M, ‐ Wave loads estimation M, ‐ Current loads estimation W ‐ Other loads estimation W ‐ Alternative sediment R ‐ Alternative jetty layouts M, ‐ Alternative channel M, Trade off between A ‐ Preparation and W All ‐ Writing mid term report All Detailed design of best Al ‐ Wave loads calculation ‐ Current loads calculation ‐ Other loads calculation ‐ Detailed design sediment ‐ Detailed design jetty ‐ Detailed design channel ‐ Jetty structural design ‐ Foundation design ‐ Jetty material Report and presentation Al ‐ Writing chapters ‐ Conclusions and ‐ Finilizing report ‐ Preparing presentations ‐ Presentation at CPE (one

Scheduele time Deadline Extra Time Holiday I=Ilse W=Wim M=Maarten R=Roderik


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Appendix C: Incidental loads Incidental loads are the loads that have a low probability of occurrence, this are loads that have the to be regarded during the design phase. It have to be weighted whether these loads have to be taken into account during the design process, as these loads have mostly enormous impact on the dimensions of the structure. In this appendix vessel collision and seismic loads are regarded. Vessel col l is ion A vessel collision with the breakwater is an accidental load with a relatively small probability of occurrence. The river is only used by relatively small vessels and the consequence of a vessel collision will only result in failure in the outer layer of the structure. Therefore the choice has been made to discard “vessel collision” as a design condition. It must be noted that when a breakwater is hit by a vessel and is damaged by the vessel it should be repaired quickly. When the outer layer is partly destroyed and not repaired it could result in failure of the breakwater during storm conditions. Seismic loads Seismic loads a the impact on the structure in the case of seismotectonic unrest. This reveals itself in the form of an earthquake.

Figure 50: Worldwide avaraged seismic hazard (PIANC/MarCom 34, 2001)

From Figure 50 above and information given by CPE we can conclude that seismic hazard will not be a problem for our jetty structure. So this load will be neglected.


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Appendix D: Tide table

Figure 51: Harmonic constants (FEMAR, 2000) for Region Araranguá


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Appendix E: Currents Waves approaching the coast at an angle generate longshore currents (especially in the surf zone where the waves break). Additionally, the tidal range generates ebb currents and flood currents. The parameters for current velocities and directions are derived from the numerical modelling study done by CPE (CPE, 2009: 26). The maximum velocities of the currents have been determined for the two dominant wave directions, i.e. NE‐SW (wave 1) and SW‐NE (wave 2), during low tide and high tide. This results in 4 scenarios of wave and tide dependent current velocities and directions in the present state of the Araranguá river mouth and as shown in Figure 52 to Figure 55. The period at which the currents were analyzed correspond to spring tide, where level variations are greatest and the highest current velocities can be found.

Figure 52: Current velocity and direction for wave 1 (NE‐SW) at high tide (CPE, 2009: 26)

Figure 53: Current velocities and directions for wave 1 (NE‐SW) at low tide (CPE, 2009: 27)


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Figure 54: Current velocities and directions for wave 2 (SW‐NE) at high tide (CPE, 2009: 28)

Figure 55: Current velocities and direction for wave 2 (SW‐NE) at low tide (CPE, 2009: 28)

Different current patterns will occur when the jetties are present. These velocities will be interesting as load conditions, for example for armour stability. Therefore we show the most unfavourable situation, in Figure 51 & Figure 52, which causes the highest currents at the structure. This is with wave 2.


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Figure 56: Current velocities and direction for wave 2 at low tide with jetties present (CPE, 2009: 35)

Figure 57: Currents velocities and direction for wave 2 at high tide with jetties present (CPE, 2009: 35)


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Appendix F: Sediment transport analysis Simulations done by CPE are meant to simulate the response of the transport of sand along the coast when interrupted by either three of the observed alternatives for jetty systems. This was again done according to the same simulations, observing the effects of the two dominant wave cases (the southern quadrant or wave 1 and the north‐eastern quadrant or wave 2) during low tide and high tide on the system. Further attention must be given to alternative 2 from the CPE report, where the northern jetty is interrupted by the channel adjacent to the coastline and the channel has open passage to the sea in the north. This alternative presents the more likely future solution. Again, wave 1 does not have a great of influence on the system, with transport rates not more than 0,0002 m3/m/s. There is a shadow zone of a few hundred meters to the SW of the southern jetty. Wave case 2 on the other hand, shows a clear reaction to the interruptions with sediment transport rates reaching more than 0,0016 m3/m/s. Figure 58 illustrates the path of the transport along the structure and the magnitudes, for wave 2 during low tide. A quite similar pattern is found in the high tide situation, shown in Figure 59. Figure 60 shows the relative change between the transport rates in the Current Situation and when applying alternative 2 (wave 2, low tide).

Figure 58: Simulation of the net sediment transport rate for Alternative 2, wave case 2 during low tide. (CPE,

2009: 51)


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Figure 59: Simulation of the net sediment transport rate for Alternative 2, wave case 2 during high tide. (CPE,

2009: 50)

Figure 60: Simulation of the relative change of net sediment transport rates for Alternative 2 compared to

the Current Situation, for wave case 2 during low tide. (CPE, 2009: 51)

Figure 60 illustrates that there is a significant decrease in the sediment transport rates in front of the land edges of both the northern and southern jetty, in between the jetties, and in the shadow zone of the jetty system. A significant increase is found at the heads of the jetties, where the sand is driven along the structure. From the above an approximation of the required boundary conditions with regard to sediment transport can be found. The results of wave refraction and transport rate analysis in the Current Situation are:

• The amount of wave induced sediment transport o NE‐SW directed transport 0,0002 m3/m/s o SW‐NE directed transport 0,0016 m3/m/s


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o Net transport 0,0014 m3/m/s in the SW‐NE direction

• A non specified amount of wind induced sand transport along the coast (dune formation and migration), contributing to the formation and migration of the spit in NE direction

The results of wave refraction and transport rate analysis for alternative 2 are:

• Inlet induced transport from SW shoreline (entering the inlet); no sediment entering from the southern shoreline

• Inlet induced transport from NE shoreline (entering the inlet); none during low tide, up to 0,0002 m3/m/s during high tide, for wave case 2.

• Transport impounded by left jetty; none, assuming that an impermeable jetty structure was used in the numerical modelling

• Transport impounded by right jetty; none, assuming that an impermeable jetty structure was used in the numerical modelling

• NE‐SW directed transport naturally bypassed; an insignificantly small transport rate

• SW‐NE directed transport naturally bypassed; 0,0002‐0,0004 m3/m/s

• Wind induced sediment transport entering the channel; it is hard to estimate the exact amount of windblown sand that will enter the channel from the adjacent beaches and dunes. In the CPE report (2009) some attention was given to the effect of a jetty system on the future development of the spit. Due to lack of longshore transport and low discharge rates from the river, the channel parallel to the shoreline in the Current situation will be completely closed off because of landward directed winds. Using historical aerial photographs provided by (SILVA, 2009) an estimation has been made for the progression of the protected shoreline, which is 4,07 m/year, only by wind. This means that the 211 m wide channel will be closed of within 52 years, without taking other effects into account. The obtained progression rate could give an indication for the amount of windblown sand entering the navigation channel of the jetty system. However, it is expected that, compared to the annual net longshore transport rate, this share is relatively small.

The parameters found for alternative 2 apply only to this jetty design. However, it does give a very good estimation on the quantities of sediment transport along a structure and it is expected that these transport rates will not change significantly when altering the jetty design layout. For the preliminary design of the Araranguá jetties, using these values for the design of the bypass system will be adequately accurate. In a more detailed design process, further numerical modelling is needed to investigate the true consequences of the jetty system design. Another very important aspect in the design process of the bypass system is the sand deposition/siltation volume and location, i.e. erosion and sedimentation. Therefore, CPE (2009) has made erosion/sedimentation patterns for the various alternatives after a one year simulation. Figure 61 shows the erosion/sedimentation pattern for alternative 2.


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Figure 61: The erosion and sedimentation pattern after a 1 year simulation for Alternative 2. Green implies

sedimentation, red implies erosion. (CPE, 2009: 68)


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Appendix G: Determining design wave height In this chapter it will be clarified how we come to the different design wave heights used to calculate the different loads on the jetty structure. Wave data The wave data used in this project is a series of eleven years, 1998‐2008, obtained at the break of the continental shelf, close to the project site. Data was acquired by the National Oceanic and Atmospheric Administration (NOAA) using the program WavewatchIII (Tolman, 2009) and then translated by CPE. Wavewatch III simulates processes of generation / propagation of waves in deep water, based on meteorological data reanalyzed. The series contain the significant wave height (Hs), the peak period (Tp) and the predominant direction of the waves every three hours. Design l i fe t ime and probabil ity of damage As described in the functional analysis, the jetty will be designed for a lifetime of 50 years, with a probability of serious damage during its lifetime of 10%. The probability of serious damage during the lifetime of the construction is given by the Poisson distribution (Verhagen et al., 2009):

)*exp(1 LTfp −−= (1)

Where

• p probability of occurrence of the design storm one or more times in period TL

• TL lifetime of the breakwater in years

• f average frequency of the design storm per year

Using TL = 50 years and p = 10%, (1) becomes:

1 *ln(1 0.1) 1/ 47550

f per year= − − = (2)

Furthermore a 1/50 year storm is estimated for structural design purposes. Estimation of design wave height This section will cover the steps taken and analysis that has been carried out for the estimation of the design wave height needed to estimate the loads in the Ultimate Limit State (ULS). First the method is explained, followed by the results of the analysis. To conclude this chapter, the results that will be used as boundary conditions will be summarized. The method for analysis is entirely derived from the lecture notes by Verhagen (2009). Method for analysis Wave measurements are grouped according to the measured significant wave height, Hs. A probability, P, for any wave height H’s being equal or less than a certain wave height is defined as:

'( )s sP P H H= ≤ (46)

Following from (3) is the probability of exceedance, Q:

'( ) 1s sQ Q H H P= > = − (47)


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Table 26 gives an overview of the wave data.

Table 26: Wave data

Wave heigth class Hs (cm) Number of observations P Q per bin cumulative

0 25 0 0 0,000000 1,000000 25 50 8 8 0,000249 0,999751 50 75 61 69 0,002147 0,997853 75 100 717 786 0,024453 0,975547 100 125 2288 3074 0,095635 0,904365 125 150 4004 7078 0,220203 0,779797 150 175 4997 12075 0,375665 0,624335 175 200 4933 17008 0,529135 0,470865 200 225 4117 21125 0,657219 0,342781 225 250 3239 24364 0,757988 0,242012 250 275 2315 26679 0,830010 0,169990 275 300 1698 28377 0,882836 0,117164 300 325 1195 29572 0,920014 0,079986 325 350 854 30426 0,946582 0,053418 350 375 522 30948 0,962822 0,037178 375 400 374 31322 0,974458 0,025542 400 425 271 31593 0,982889 0,017111 425 450 184 31777 0,988613 0,011387 450 475 119 31896 0,992316 0,007684 475 500 84 31980 0,994929 0,005071 500 525 64 32044 0,996920 0,003080 525 550 35 32079 0,998009 0,001991 550 575 23 32102 0,998724 0,001276 575 600 24 32126 0,999471 0,000529 600 625 10 32136 0,999782 0,000218 625 650 3 32139 0,999876 0,000124 650 675 1 32140 0,999907 0,000093 675 700 3 32143 1,000000 0,000000

Total 32143 Ns 2922

To come to an estimation of a design storm, an extrapolation was made from the wave data provided by CPE. Therefore an analysis of all the storms in the available data is made. According to Peak over Threshold (PoT) method a storm is defined as a period in which the wave height is higher than a certain threshold value (Ht). The highest wave in the storm defines the magnitude of the storm. The result of the PoT analysis is a database of all storms during the period in which the measurements (or simulation) took place. It is stated that ‘by adding a threshold it is avoided that small variations in wave height during long, calm periods have a significant influence on the final result’ (Verhagen et al., 2009:269). However, the number of storms should always be high enough that a reliable analysis can be made. Exactly when the analysis becomes less reliable has not been defined in literature. Therefore it is chosen to take 100 identified storms (storms per year (Ns=9) as a minimum for a reliable analysis, resulting in different threshold values for different datasets.


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Furthermore the analysis of all waves presented in the CPE report (2009), shows that there are two main wave directions, ENE and S. In Figure 62 different wave heights have different colors: purple for the highest and grey for the lowest waves. For the two main directions this means that waves from ENE will be significantly lower than waves from S. The choice has been made to perform two separate analyses for the different directions and an analysis for all waves to be able to compare the results. Wave data for analysis of the South waves contain all data from waves from S, SSW and SW. A thorough investigation of the wave heights reveals that some of the highest storms are identified coming from SW and SSW. The ENE direction is representative for neighbouring waves, therefore this direction is assessed only as ENE.

Figure 62: Direction and magnitude of the waves (CPE, 2009: 11)

From the above different starting points for the analysis have been derived:

• For all waves threshold values of 1,5; 2,0; 2,5; 3,0; 3,5; 4,0 m are used

• For waves from direction SSW threshold values of 1,5; 2,0; 2,5; 3,0; 3,5; m are used

• For waves from direction ENE threshold values of 1,5; 2,0; 2,5; m are used When the number of storms from different directions has been identified, an extrapolation must be made to determine the design wave height. Different distributions for extrapolation are known: the exponential, the Gumbel and the Weibull distribution. The latter two distributions result in more reliable predictions. The exponential distribution is made by deriving a value Qs that is plotted against the wave height:

s sQ N Q= ⋅ (48)

The logarithmic trend line of the plot gives the design wave height for any probability of exceedence, in our case 1/475 per year. With Hss being the storm wave height, the Gumbel distribution is given by:

( ) exp exp ssss

HP H H γβ

⎡ ⎤⎛ ⎞−≤ = − −⎢ ⎥⎜ ⎟

⎝ ⎠⎣ ⎦ (49)

Which can be modified into:

ssG A H B= ⋅ + (50)

For G, A and B the following relations hold:


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1 1ln ln ; ;G BP A

β γ β⎛ ⎞= − = = − ⋅⎜ ⎟⎝ ⎠

(51), (52), (53)

The values for A and B have to be determined by linear regression of the computed values for G and Hss. From the above relations it can be derived that the design wave height can be computed as:

ln ln sss

s s

NHN Q

γ β⎛ ⎞

= − ⎜ ⎟−⎝ ⎠ (54)

In the same way as the Gumbel distribution, the Weibull distribution is derived using the following relations:

( ) exp ssss

HP H Qα

γβ

⎡ ⎤⎛ ⎞−= = −⎢ ⎥⎜ ⎟

⎢ ⎥⎝ ⎠⎣ ⎦ (55)

Which can be modified into:

ssW A H B= ⋅ + (56)

For W the following relations hold:

( )1

lnW Q α= − (57)

To compute Hss from the relations above it can be derived that the following must hold:

1

ln sss

s

QHN

α

γ β⎛ ⎞

= + −⎜ ⎟⎝ ⎠

(58)

It must be noted that there are 3 variables, α, β and γ, but linear regression gives only β and γ. The last variable, α, is determined by modifying α in relation (56) in a way that the highest correlation coefficient is obtained. Furthermore γ should have about the same value as the applied threshold value. Comparing the values of γ and the threshold value gives a measure of the reliability of the relations obtained by the Gumbel and Weibull distribution. Results of analysis Below all results will be summarized in different tables. In order to show the method of computation one example is presented. To visualize the analysis the relations for different parameters have been plotted in the figures below. For the general case in which waves from all directions are present, using a threshold value of 1,5 meter, the data obtained are shown in Table 2. For the construction of the table the formulas explained above are used. To obtain clear figures, not all values of G and W have been used. The negative value for G and the lowest values in Table 27 have been omitted. Therefore the values for β and γ that have been calculated differ from the values shown in the graph.


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Table 27: Storm data, all waves, threshold = 1,5 m

Wave heigth class Hss (cm)

Number of storms P Q Qs G W

per bin Cumulative α = 2,11

150 175 148 148 0,246256 0,753744 41,181818 ‐0,337459 0,549499 175 200 88 236 0,392679 0,607321 33,181818 0,067462 0,719109 200 225 52 288 0,479201 0,520799 28,454545 0,307022 0,816750 225 250 53 341 0,567388 0,432612 23,636364 0,567903 0,919605

250 275 36 377 0,627288 0,372712 20,363636 0,762819 0,993793 275 300 40 417 0,693844 0,306156 16,727273 1,006465 1,083190 300 325 36 453 0,753744 0,246256 13,454545 1,263359 1,173433 325 350 26 479 0,797005 0,202995 11,090909 1,483271 1,247499 350 375 23 502 0,835275 0,164725 9,000000 1,714827 1,322450 375 400 27 529 0,880200 0,119800 6,545455 2,058804 1,428396

400 425 11 540 0,898502 0,101498 5,545455 2,234685 1,480244 425 450 17 557 0,926789 0,073211 4,000000 2,576631 1,576911 450 475 13 570 0,948419 0,051581 2,818182 2,938245 1,673714 475 500 7 577 0,960067 0,039933 2,181818 3,200234 1,740703 500 525 6 583 0,970050 0,029950 1,636364 3,493058 1,812739 525 550 5 588 0,978369 0,021631 1,181818 3,822732 1,890573 550 575 3 591 0,983361 0,016639 0,909091 4,087632 1,950826 575 600 4 595 0,990017 0,009983 0,545455 4,601823 2,062571 600 625 3 598 0,995008 0,004992 0,272727 5,297482 2,204236 625 650 2 600 0,998336 0,001664 0,090909 6,397762 2,410075 650 675 0 600 0,998336 0,001664 0,090909 6,397762 2,410075

675 700 1 601 1,000000 0,000000 0,000000

Total 601 Ns 54,64

For the Gumbel distribution the following values are obtained calculating the values from Table 27:

0,013 2,81A and B= = −

Therefore:

1/ 475

1 79,04; ( 2,81) 79,04 221,970,013

54,64221,97 79,04 ln ln 10, 25154,64 475ssH meter

β γ= = = − − ⋅ =

⎛ ⎞⎜ ⎟= − ⋅ =⎜ ⎟−⎝ ⎠

In a similar manner the design storm wave height according to the Weibull distribution can be calculated:

0,004 0,006A and B= =

Therefore:

12,11

1/ 475

1 284,80; 284,80 0,006 1,740,004

14751,74 284,8 ln 8,53

54,64ssH meter

β γ= = = − ⋅ = −

⎛ ⎞⎜ ⎟= − + ⋅ − =⎜ ⎟⎝ ⎠


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Figure 63 shows a graph of the exponential distribution. It can be noted that the data points do not lie on a straight line. The correlation coefficient is 0.95, less than the ones calculated for the Gumbel (0.98, Figure 59) and Weibull (1,00, Figure 65). Both Figure 64 and Figure 65 show that the data points represent a straighter line, as would be expected from the calculations. The design storm wave height for the exponential distribution is:

1/ 475184,3 ln 538,3 10,58

475ssH meter⎛ ⎞= − ⋅ + =⎜ ⎟⎝ ⎠

Figure 63: Exponential exceedence graph

Figure 64: Gumbel exceedence graph

y = ‐84,39ln(x) + 538,32R² = 0,949

0

100

200

300

400

500

600

700

800

0,010,101,0010,00100,00

Storm height H

ss (cm)

Storm exceedence probability per year

y = 77,929x + 228,61R² = 0,9636

0

200

400

600

800

1000

1200

0 2 4 6 8 10 12

Storm height H

ss (cm)

Gumbel reduced variable


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Figure 65: Weibull exceedence graph

It is to be noted that the difference for the above example between the Gumbel and Weibull distributions is rather big. It is most likely that this is caused by the relatively low threshold value. In the overall result, one can see that this difference becomes less. Analysis for different values of Ht results in the following values for the design wave height, presented in different tables for the 3 different cases. Table 28 presents the wave height as calculated for all waves that are present in the record. Table 29 gives the design wave heights for the waves from SSW. The waves from ENE are described by the parameters in Table 30.

Table 28: Design storm wave height, all waves Ht 1,5 2,0 2,5 3 3,5 4 Ns 54,64 65,82 48,64 31,09 17,73 9,18

Hss,1/475 Exponential 10,58 9,94 9,59 9,34 9,09 8,84 Hss,1/475 Gumbel 10,25 9,67 9,33 9,10 8,87 8,63 Hss,1/475 Weibull 8,53 8,46 8,47 8,30 8,19 8,08

Table 29: Design storm wave height, direction SSW

Ht 1,5 2,0 2,5 3 3,5 Ns 61,27 50,91 35,00 23,36 15,27

Hss,1/475 Exponential 10,32 10,00 9,71 9,43 9,14 Hss,1/475 Gumbel 9,91 9,68 9,40 9,14 8,89 Hss,1/475 Weibull 8,61 8,45 8,34 8,21 8,20

Table 30:Design storm wave height, direction ENE

Ht 1,5 2,0 2,5 Ns 42,73 23,00 10,64

Hss,1/475 Exponential 6,39 6,17 6,01 Hss,1/475 Gumbel 6,23 6,00 5,84 Hss,1/475 Weibull 5,58 5,45 5,38

In order to show the differences in the relation between γ and the threshold value, Figure 66 shows a graphical representation of this relation. The solid line represents the 1:1 relation between γ and the threshold value. As can be seen, the Gumbel and Weibull distribution outperform the exponential distribution in this aspect. Furthermore it is shown that the predictions get more reliable when the threshold value is higher.

y = 284,51x ‐ 0,8466R² = 0,9929

0

200

400

600

800

1000

1200

1400

0 1 2 3 4 5

Storm height H

ss (cm)

Weibull reduced variabel


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Figure 66: relation between gamma value and threshold, all waves

Conclusion According to the above explanation, the exponential distribution produces less reliable predictions than the Gumbel and Weibull distribution. Therefore the results of the calculations done with the exponential distributions will not be taken into account in the estimation of the design wave height. Furthermore there are three design wave heights that have to be established. A conservative average value of the Gumbel and Weibull extrapolations will be used as a value for the detailed design of the jetty. It must be noted that this value will always be an estimation, for there are no exact values in wave height estimation. This means that the design wave height will be:

• All waves 8,5 m

• SSW waves 8,7 m

• ENE waves 5,7 m The three different heights are used for different purposes:

• All waves are used in general situations.

• SSW waves will be used in the structural design of the south arm of the jetty, as well as the roundheads of both arms of the jetty

• ENE waves are used in the structural design of the north arm of the jetty Wave stat ist ics For further analysis of the waves on the jetty structure, SwanOne (TUDelft) is used to calculate wave characteristics at the toe of the structure. In SwanOne a wave spectrum is calculated from the input parameters significant wave height, Hs and the peak period, Tp. Since all wave data in the record are presented as the significant wave heights and peak period of a 3 hour measurement, no conversions have to be made. For the estimation of the peak period of the design storm, the wave steepness is estimated. The steepness s0 of deep water waves is defined as:

‐100

0

100

200

300

400

500

600

0 200 400 600

Gam

ma value

Selected Threshold

lineair

Exponential

Gumbel

Weibull


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00 2

2 m

m

HsgTπ

= (59)

In (59) Hm0 in deep water is equal to Hs and the mean period Tm=Tp/1,2. Also g is the gravitational acceleration with a value of 9,81 m/s2. The peak period can thus be calculated from the steepness as:

0

21,2 sp

HTgsπ

= (60)

For the waves in the record, the average steepness in a heavy storm can be calculated, using a very high threshold value. For example whit a threshold value HT = 6m, the average significant wave height of all waves in a storm is 6,38 m, and the average Tp = 10,95 s. The average steepness is now:

0 22 6,38 0,049 4,9%

10,959,811,2

s π ⋅= = =

⎛ ⎞⋅⎜ ⎟⎝ ⎠

In a similar way the steepness with different threshold values are calculated and presented in Table 31. It must be noted that there is only 1 wave that is higher than 7 meters. However, since the trend in the steepness is that it increases with wave height, this value is considered to be reliable for the estimation of steepness of a design storm wave. When estimating the wave period of a design storm a steepness of 6% will be used for a significant wave height above 7 meters. For design storms (once a year, 1/50 year) with a wave height in the range of table 31, the value for the steepness is interpolated. From this steepness the peak period is calculated.

Table 31: wave steepness in a storm

HT (m) S0 (%)

3 4,0

3,5 4,1

4 4,3

4,5 4,4

5 4,6

5,5 4,7

6 4,9

6,5 5,5

7 6,0

For calculation of different parameters as rock weight and overtopping, different design storms have to be modelled in SwanOne. Below the boundary conditions are presented in Table 32.


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Table 32: Wave statistics

Wave Direction Hs

(m) Tp (s)

Hss,1/475 SSW 8,7 11,6Hss,1/50 SSW 7,5 10,7Hss,1/1 SSW 5,5 10,4Hss,1/475 EN

E 5,7 10,5

Hss,1/50 ENE 4,9 9,9Hss,1/1 ENE 3,6 9,0Hs All 2,95 9,0

Furthermore the model calculates the waves that originate in deep water, which is defined as:

0

12

dL

≥ (61)

Where:

2

0 2mgTLπ

= (62)

In (18) and (19) L0 is the deep water wavelength and d is the depth of the water. The wavelength of the

governing wave is:

2

0

11,69,811, 2 145,9

2L m

π

⎛ ⎞⋅⎜ ⎟⎝ ⎠= =

To run the SwanOne model 0,5L = 75m depth is needed to let deep water waves indeed start in deep water.


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Appendix H: Channel cross‐sectional area versus tidal prism For this relationship we can differentiate multiple ways to calculate the optimum channel equilibrium. When determining the optimum channel equilibrium tidal prism plays a important role. According to (D'Aquino et al., 2009) the estuarine area of river Araranguá is approximately 3,7 km2, with the mean tidal range/amplitude of 0,26 m we can calculate the tidal prism for river Araranguá. Tidal prism is defined as:

2b bP A a= ⋅ ⋅ (63)

Where, P=tidal prism Ab=bay area ab=bay amplitude P=1,924*106 m3=0,68*108 ft3

We will use this tidal prism to calculate the channel equilibrium with two methods described below. O’Briens relationship This relationship focuses on the flow through the inlet and the minimum cross‐sectional area that can be sustained by that flow. O’Brien (1931) was the first to determine this empirical relationship, Jarret (1976) continued this analysis and extended it for various coastal areas, because the parameters of this empirical relationship change with different tidal characteristics.

qcA C P= ⋅ (64)

Where in equation 64, Ac= the minimum inlet cross‐sectional area in square meters C = constant q = constant; values vary between 0,85 and 1,05 Jarret recommends to use O’Briens equation for dual‐jettied inlets, given in equation 65.

4 0,867, 489 10cA P−= ⋅ ⋅ (65)

Below in Figure 67, O'Brien's formula (in feet) is plotted, the lines show average depth expected for given jetty spacing and tidal prism. With the use of this graph the minimum distance for jettied inlets can be calculated. The data points plotted in Figure 67 describe actual field conditions for 44 inlets.


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Figure 67: Tidal prism (P) versus jetty spacing (W) (U.S. Army corps of engineers, 2006: II‐6‐49)

The minimum depth is given by shipping boundary conditions (6 m = 19,7 feet) and the tidal prism is given: 0,68*108 ft3. With this values no equilibrium can be found in this graph. Escoffier The criteria for the stability of the inlet have first been studied by (Escoffier, 1940), where he provided a criterion for determining whether an existing inlet was more likely to stay open or close depending on its position on the "Escoffier" curve (Figure 68). In this figure a hydraulic relationship between the inlet cross sectional area and the maximum velocity in the inlet channel is plotted (closure curve). Escoffier reasoned that there was an equilibrium or critical velocity Vcr where the cross‐sectional area is in equilibrium, this is the dotted line in the graph (equilibrium flow curve). Escoffier stated that flow of 1m/s will generate a stable cross‐section. Later on also more complicated equilibrium flow curves came up. The intersections in the graph are equilibrium states; where point B is unstable and D is stable.

Figure 68: Escoffier curve (d'Angremond and Pluijm‐van der Velden, 2001: 143)


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The equilibrium flow curve and the closure curve will have to be addressed for river Araranguá. First the equilibrium flow curve is modeled with the empirical relationship between cross‐sectional area and tidal prism for inlets at equilibrium and then equation 65, O’Briens relationship for dual jetties, is used. The velocity v is defined as:

ˆc

PVA Tπ

= (66)

Where, V=velocity in the channel. If we combine equation 67 and 68, we can eliminate the tidal prism P and this will lead to the expression for the equilibrium velocity as a function of cross‐sectional area:

1 1 11ˆ q q

mV C A Tπ− −

−= ⋅ (67)

With constants as in equation 21; C =7,489*10‐4, q = 0,85, this formula will become:

0,1764ˆ 0,35067mV A= (68)

A PC program is available to perform the ‘Escoffier’ analysis (Seabergh and Kraus, 1997) which uses the method described above. The tool is derived from the Coastal Inlet Management Package of CIRP of the US Army corps of Engineers. This Channel Equilibrium Area program (CEA) determines the minimum cross‐sectional channel area for a coastal inlet. For a given inlet scenario, an analytical 1‐D model is used to calculate inlet hydraulics. This information is combined with a tidal prism‐minimum channel area relationship (Equation 67) using the channel stability concept of Escoffier to determine the minimum equilibrium area. Channel Equilibrium Area (CEA) is a software program that provides a simple, quick method of calculating the parameters required to achieve equilibrium in coastal inlet water channels. Ocean tide, tidal period, channel length and width are key factors that contribute to the equilibrium state of a channel. CEA allows the user to test multiple values for each one of these factors to assess their impact on the equilibrium state. A simulation of the Araranguá river has been made with the CEA program with a variable depth input. When ranging the width from 150 to 80m, none of the flow equilibrium and closure curves intersect. So for this method also no equilibrium cross‐sectional area is can be obtained.


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Appendix I: Overtopping outputs To optimize the design of the crest height of the different jetty cross‐sections first the different options to increase parameters is investigated. The crest width is given by the equipment requirements for construction, this value already exceeds the minimal crest width needed by design rules by far. Therefore is chosen not to increase this value anymore. The implementation of a berm can be considered to reduce overtopping, but the influence is less then increasing the crest height. Because the given design without berm complies to all other requirement criteria, the crest height needed for overtopping criteria does not seem extreme high and the influence is less, no berm will be implemented to reduce overtopping. Also it assumed no crown wall will be constructed. Therefore only the variation of freeboard will be investigated, to obtain the minimal freeboard height for each cross‐section the freeboard will be increased with steps of 10 cm. First for the normal condition criteria (0,1 l/s/m) the value is updated, because this value looked governing in empirical approach. In Table 33 the output is given, in the last column the star indicates the minimal freeboard value that meets the requirements

Table 33: Optimalization freeboard for each cross section (H1/1)

Cross‐section FreeboardRc (m)

Overtopping qmean (m

3/s/m)

Starred meets

requirement 5 4,5 0,6400E‐04 4,4 0,7284E‐04 4,3 0,8566E‐04 4,2 0,9496E‐04 * 4,1 0,1087E‐03 4 2,0 0,5622E‐04 1,9 0,6717E‐04 1,8 0,8055E‐04

1,7 0,9680E‐04 * 1,6 0,1165E‐03 1 1,5 0,7937E‐04 1,4 0,9987E‐04 * 1,3 0,1260E‐03 1,2 0,1591E‐03 1,1 0,2013E‐03 8 2,2 0,9516E‐04 * 2,1 0,1155E‐03 2,0 0,1406E‐03 1,9 0,1719E‐03 1,8 0,2039E‐03 7 1,5 0,7440E‐04 1,4 0,9331E‐04 * 1,3 0,1173E‐03 1,2 0,1476E‐03 1,1 0,1861E‐03

Of course the new freeboard values calculated for normal conditions also have to meet the requirements for extreme condition (10 l/s/m). In Table 34 the new mean overtopping values at extreme wave conditions (H1/475) are showed, starred rows do not meet the requirements for this events.


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Table 34: Check overtopping for H1/475 wave conditions

Cross‐section Rc Qmean (m3/s/m)

Starred don’t meet 10 l/s/m

5 4,2 0,2749E‐02 4 1,7 0,1136E‐01 * 1 1,4 0,4319E‐03 8 2,2 0,1320E‐02 7 1,4 0,4814E‐03

Because from Table 34 it can be concluded all cross‐sections meet the overtopping requirement for extreme conditions, except cross‐section 4, so in most conditions normal conditions are governing. For cross‐section 4 the freeboard should be increased to meet the criteria for the storm conditions. This is also done with steps of 10 centimeters (see .

Table 35: Optizimalization cross‐section 4 for H1/475

Rc Qmean (m3/s/m)

Starred meet 10 l/s/m

1,7 0,1136E‐01 1,8 0,9929E‐02 * 1,9 0,8670E‐02 2,0 0,7563E‐02 2,1 0,6593E‐02

The values that will be used for freeboard can be obtained from the tables above and are given in paragraph on overtopping calculations. As stated in the paragraph ‘overtopping calculations’ the CLASH network indicates the reliability of the answer. When calculate overtopping for the obtained freeboard values for extreme en normal conditions almost all get the same indication, remark 4: For prototype, rough‐sloping structures, a correction factor is applied: q = ’. According to the user manual (Coeveld et al., 2005) of the network tool this means: ‘With this remark the user is given the possibility of choosing between the direct output of the NN model or a corrected value to account for model effects, scale effects and wind effects in prototype situations, for rough‐sloping structures ( γf < 0.9 and cot α >1).’ All this corrected values are less than the maximum overtopping that is allowed regarding the criteria, so we can say that the values for Rc are sufficient. For the overtopping at cross‐section 5 at normal conditions gets remark 5: For prototype, rough‐sloping structures, a correction factor is applied: q = ’ , Since 10‐6 < Q < 10‐5, the NN prediction is less reliable (indicative)’ In which the first sentence means the same as in remark 4 and second part mean according to the manual of CLASH (Coeveld et al., 2005): ‘The user is also warned that the dimensionless value of the mean overtopping discharge predicted by the NN before the correction is applied lies in the region 10‐6 < Q < 10‐5, and therefore the NN prediction should be considered as less reliable, or indicative’. So this single overtopping calculation is less reliable, but it will be sufficient for this design stage.


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Appendix J: Bearing capacity and settlement Calculat ion of the bearing capacity Calculation of the bearing capacity can be done with different equations. Theory uses the Terzaghi, Prandtl (U.S. Army corps of engineers, 2004: 06b Settlement in Sand & Bearing Capacity), Brinch Hansen (Molenaar et al., 2008) and Meyerhof (U.S. Army corps of engineers, 2006:2383) to calculate the maximum bearing capacity. The Brinch Hansen and the Meyerhof methods are a further development of Prandtl’s and Terzaghi’s methods and are valid for static loading and homogenous soil conditions. The lather are preferred over Prandtl’s and Terzaghi’s theories. Meyerhof and Brinch Hansen use the same equation to determine the maximum soil bearing capacity:

max' ' ' 0,5 'c c c c q q q qp c N s i d q N s i d BN s i dγ γ γ γγ= + + (69)

Whereby:

cN , qN and Nγ are factors for the bearing force.

cs , qs and sγ are factors for the shape of the foundation

ci , qi and iγ are factors for the horizontal load

cd , qd and dγ are factors for the depth of the foundation with regard to the surrounding soil

'c is the weighted cohesion

'q is the effective stress at foundation depth next to the foundation

'γ is the effective volumetric weight of the soil

B is the width of the effective foundation area

max'p is the maximum effective stress on the effective foundation area

'ϕ is the effective friction angle of the soil

The factors for the bearing force, shape of foundation, the horizontal load and the depth of the foundation are dependant of the above and the following parameters: Width (B) and length (L) of the foundation area (Meyerhof) Effective width (B’ ) and length (L’ ) of foundation area (Brinch Hansen) Horizontal (H) and vertical (F) force on foundation Depth (D) of foundation below surrounding soil surface Because horizontal forces can be neglected on a rubble mound breakwater (decided in collaboration with ir. H.J. Verhagen, Delft University of Technology, section Hydraulic Engineering) and the there is no cohesion equation 69 can be shortened and we can choose to work with the Meyerhof method since the vertical force grabs on centrically. Below the new shortened equation (equation 70) and an explanation of the factors used in this equation are given.

max' ' 0,5 'q q qp q N s d BN s dγ γ γγ= + (70)

The factors used in (70) will be explained below: Bearing capacity factors:


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( ) ( )

tan '1 sin ' e1 sin '

1 tan 1.4 '

q

q

N

N N

π ϕ

γ

ϕϕ

ϕ

⎛ ⎞+= ⎜ ⎟−⎝ ⎠

= −

Shape coefficients:

( ) ( )1 0.1 tan 45 ' 2qs s B Lγ ϕ= = + + for ' 10ϕ ≥ °

Depth coefficients:

( ) ( )1 0.1 tan 45 ' 2qd d D Bγ ϕ= = + + for ' 10ϕ ≥ °

The inclination coefficients iγ and qi are equal to 1 because there is no horizontal force.

To calculate the maximum bearing capacity input parameters from the heaviest part of the breakwater, a profile at the end of the NE breakwater, are used:

' 35ϕ = ° 3' 10 /kN mγ = 0,3fD m=

' 3q kPa= 55B m= 455L m=

This gives the following results:

33,30qN = 37,15Nγ = 1,04qs = 1,04sγ = 1, 00qd = 1, 00dγ =

The maximum bearing capacity will then be:

max

max

max

' ' 0,5 '

' 3 33,30 1,04 1,00 0,5 10 55 37,15 1,04 1,00' 10,79

q q qp q N s d BN s d

pp MPa

γ γ γγ= +

= ⋅ ⋅ ⋅ + ⋅ ⋅ ⋅ ⋅ ⋅=

Calculation of weight: Because it is expected that the bearing capacity will be sufficient a first estimation is made with the following parameters:

• Average porosity P = 0.4

• Volume per meter above sea level (DHW) V/L = 62m3/m

• Volume per meter below sea level (DHW) V/L =286m3/m The total pressure becomes:

( )62 26,5 0.6 985 /KN m⋅ =

( )286 26,5 0.6 10 0,4 5692 /KN m⋅ + ⋅ =


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( ) 21383 7164122 / 0,12

55KN m MPa

m+

= =

The bearing capacity is sufficient. Because the safety factor 10,79 0,12 88.44= is very large more detailed

calculations and calculations for other cross sections are not necessary. Calculat ion of sett lement There are a few considerations on which the calculation of the settlement in this situation is based. Settlement occurs very rapidly, there will be no primary settlement or creep and the settlement is presumed to be small. The information in this chapter is derived from the US Design Guide Lines: 06b – Settlement in sand and bearing capacity (U.S. Army corps of engineers, 2004). The Schmertmann Strain Factor Method (CPT) (equation 71) has been chosen to calculate the initial settlement because it’s a widely used method to estimate settlement under the centre of a footing. The result could be conservative and the effect could be negated due to disturbance of the soil during the construction process. Still the Schmertmann method is the best method available in this situation. Schmertmann Strain Factor Method:

1 21

nz

ii i

IC C zE

ρ σ=

⎛ ⎞= Δ Δ⎜ ⎟⎝ ⎠

∑ (71)

Whereby: ρ is the initial settlement

1C is correction factor 1: 0'1 0.5 0.5'z

σσ

⎛ ⎞− ≥⎜ ⎟Δ⎝ ⎠

2C is correction factor 2: 101 0.2log0.1

yrst⎛ ⎞− ⎜ ⎟

⎝ ⎠

σΔ is the foundation pressure increase at the bottom of the footing

E is the Young’s modulus of the sand

izΔ is the height of every sub layer

zI is the strain influence factor at mid height at each sub layer

yearst = is the time in years after placement

In order to get a result from equation 71 some further assumptions have to be made. Interpolation of the

strain influence distribution is needed because /L B is not greater than 10 everywhere. 60/cq N is set at 5, so

E becomes 175 MPa. Again we calculate for the heaviest part of the structure at the end of the SW breakwater. Some input parameters:

Length: 455L m=

Width: 55B m=


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Effective weight: 3' 8 /s KN mγ =

Depth of embedment 0.3fD m=

Time in years after placement of footing: 50yearst year=

Pressure at foundation 2122 /q KN m=

Some further calculations:

At 0z = 0,10,1 1 0,189z

LIB

⎛ ⎞= + − =⎜ ⎟⎝ ⎠

1 0,1zI = 10 0,2zI =

( ) 2' 18 10 55 440 /zp zp u KN mσ σ= − = − ⋅ =

( ) 20' ' 122 0,3 18 10 119 /z q KN mσ σΔ = − = − ⋅ − =

' 1190.5 0.1 0.5 0.1 0,55' 440

zzp

zp

I σσΔ

= + = + = at:

( ) 0,50,5 1 49,729zp

Lz I BB

⎡ ⎤⎛ ⎞= + − =⎜ ⎟⎢ ⎥⎝ ⎠⎣ ⎦ 1 0,5 27,5z B= = 10 1,0 55z B= =

0zI = at: ( ) 20,5 1 198,99zp

Lz I BB

⎡ ⎤⎛ ⎞= + − =⎜ ⎟⎢ ⎥⎝ ⎠⎣ ⎦1 2 110z B m= = 1 4 220z B= =

The interpolation is visualized in figure 69

Iz distribution

0.00

0.50

1.00

1.50

2.00

2.50

3.00

3.50

4.00

4.50

0.00 0.20 0.40 0.60

Iz

Z=(z

/B)

L/B=1

L/B=8.27

L/B=10

Figure 69: Iz distribution


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The we calculate the correction factors:

01

'1 0.5 0.99'z

C σσ

⎛ ⎞= − =⎜ ⎟Δ⎝ ⎠

2 10 10501 0.2log 1 0.2log 1.54

0.1 0.1yrst

C⎛ ⎞ ⎛ ⎞= − = − =⎜ ⎟ ⎜ ⎟

⎝ ⎠⎝ ⎠

And finally we can calculate the settlement:

( ) ( )

1 21

0,55 0,18 2 0,55 20.99 1.54 49,72 198,9 49,7 0.27175 175

nz

ii i

IC C zE

m

ρ σ

ρ

=

⎛ ⎞= Δ Δ =⎜ ⎟⎝ ⎠

⎡ ⎤−⎛ ⎞ ⎛ ⎞= ⋅ ⋅ ⋅ + ⋅ − =⎢ ⎥⎜ ⎟ ⎜ ⎟⎝ ⎠⎢ ⎥⎝ ⎠⎣ ⎦

∑

The settlement of 27cm calculated above is calculated for the heaviest part of the structure. Also the US Design Guide Lines (2004) state that this could be a conservative value. Because it is expected that no differential settlement will take place, the structural integrity is not being compromised. For overtopping the breakwater should be designed 27cm higher. We advice to do this over the whole structure. When it is necessary to no know the settlement on different points of the breakwater some extra calculations can be done or an expert can be hired.


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Appendix K: Wave propagation Validat ion To validate the data obtained from the SwanOne modeling we compare the results of CPE (2009) (obtained with Delft3D‐Wave, which is based on a 3D version of Swan) with the SwanOne model using the same input as in the CPE report. The results of this comparison are shown in Table 36.

Table 36: Comparison between wave propagation modelled in Delft3D by CPE and wave propagation

modelled in SwanOne

Hs (m) Tp (s) angle (° from north)

Measured depth (m –

IBGE)

Hs (m) CPE report

Hs (m) SwanOne

1,5 6,8 62 3,0 0,7 0,82 (+17%) 1,5 6,8 62 6,0 0,75 0,85 (+13%) 4,3 10,7 197 3,0 1,5 1,70 (+13%) 4,3 10,7 197 6,0 2,3 2,45 (+6%)

The results show that the values obtained with SwanOne are 6% ‐ 17% higher than the values obtained in the CPE report. For now this is an acceptable (maybe conservative) difference. The SwanOne model will be used to calculate the design wave height in the shallow waters. The 3D model that has been made by CPE can later on be used again with the new wave input. It is recommended to do this in the future to model the wave heights more precisely. This could result in smaller cross sections and thus a cheaper breakwater. Design waves The design rules require different wave heights. Rules for rock size and gradation require different input wave heights then rules for overtopping, therefore multiple waves have to be modeled. The waves in Table 37 will be modeled with SwanOne.

Table 37: Waves to be modelled with SwanOne

Wave Deep water wave height

Hs (m)

Peak period Tp (s)

Direction Water level

(IBGE) Wind (m/s)

Hss,1/475 8,7 11,6 SSW HAT + Storm surge

24 – SSW

Hss,1/50 7,5 10,7 SSW HAT + Storm surge

24 – SSW

Hss,1/1 5,5 10,4 SSW MSL 0 Hss,1/475 5,7 10,5 ENE HAT + Storm

surge 12 – ENE

Hss,1/50 4,9 9,9 ENE HAT + Storm surge

12 – ENE

Hss,1/1 3,6 9,0 ENE MSL 0 Hs 2,95 9,0 SE MSL 0

Other input When trying to simulate the wave propagation the bathymetry is needed. The bathymetry needs to be perpendicular to the coast. Deep water waves will have a different behavior when they interact with the bottom. This happens when d / L <0,5. This means that for a wave with T=11,2s bathymetry until a depth of


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100m is needed. The bathymetry that is needed has to be perpendicular to the shoreline. The result is a bottom profile of 74km long from 0m to 100m depth. Figure 70 shows the bottom profile.

Figure 70: Bottom profile perpendicular to coastline

Currents could have influence on the wave propagation. When entering current data in SwanOne it needs to be detailed. This data is not available in detail. Because currents only have a small influence they are not taken into account in this simulation. Wind, tides and surges also have an influence on the wave propagation. In case of extreme loads like the design storm wind, high astronomical tide and storm surge are taken into account. In case waves with a higher chance of occurrence are modeled mean sea level is used and wind and surges are not taken into account.

-100-90-80-70-60-50-40-30-20-10

0

0 10000 20000 30000 40000 50000 60000 70000

dept

h in

m

x in m

Bottomprofile

d (m)


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Appendix L: Alternative layouts In this appendix two alternative layouts are presented for the jetty design. Alternative layouts will respond differently to local circumstances at Araranguá River like wave attack, sediment transport and ebb currents. One alternative may respond better to a certain circumstance than the other. However, when alternating the jetty layout, one can expect that valuable properties will be lost when trying to obtain the other. Alternat ive 1 In Alternative 1 the northern jetty arm is moved further towards the north and the sections of the southern jetty are connected. This alternative allows for a larger area of the deposition basin and better maneuverability possibilities for dredging operation. The northern arm can be built with a shorter seaward length. Wave action in the basin will, however, still be considerable and there is a chance of channel flow through the deposition basin. Alternat ive 2 In alternative 2, the northern arm is moved further to the north as well. The southern arm is extended seaward with a weir section perpendicular to the shoreline. After the weir transition zone, the jetty is redirected towards the navigation channel to ensure hydraulic efficiency. The deposition basin will be better protected against wave attack because of the weir orientation. Again, the basin can have a larger area and is positioned at some distance from the jetties, improving maneuverability. On the other hand, the seaward section of the southern arm will have to be constructed a lot heavier and deeper to withstand wave attack. This will result in a more expensive construction. Channel flow through the basin is also possible. The drawings of the alternatives are given on the next two pages (figure 71‐72).


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Figure 71: Alternative lay out 1


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Figure 72: Alternative layout 2


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Appendix M: AutoCad drawings ACAD drawing 1: Active and Dead storage ACAD drawing 2: Basin dimensions ACAD drawing 3: Cross section 1 ACAD drawing 4: Cross section 2 ACAD drawing 5: Cross section 3 ACAD drawing 6: Cross section 4 ACAD drawing 7: Cross section 5 ACAD drawing 8: Cross section 6 ACAD drawing 9: Cross section 7 ACAD drawing 10: Cross section 8 ACAD drawing 11: Cross section 9 ACAD drawing 12: Cross section locations ACAD drawing 13: Lay out dimensions ACAD drawing 14: Roundhead North ACAD drawing 15: Round Head South ACAD drawing 16: Tetrapod plan (Dn=1,6) ACAD drawing 17: Tetrapod cross section (Dn=1,6m) ACAD drawing 18: Tetrapod plan (Dn=1,4m) ACAD drawing 19: Tetrapod cross section (Dn=1,4m)


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Figure 73: ACAD drawing: Active and Dead storage


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Figure 74: ACAD drawing: Basin dimensions


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Figure 75: ACAD drawing: Cross section 1


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Figure 84: ACAD drawing: Cross section locations


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Figure 85: ACAD drawing: Lay out dimensions


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Figure 86: ACAD drawing: Roundhead North


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Figure 87: ACAD drawing: Roundhead South


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Figure 88: ACAD drawing: Tetrapod plan (Dn=1,6m)


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Figure 89: ACAD drawing: Tetrapod cross section (Dn=1,6m)


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Figure 90: ACAD drawing: Tetrapod plan (Dn=1,4m)


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Figure 91: ACAD drawing: Tetrapod cross section (Dn=1,4m)


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Jetty design at Rio Ara rang uá