AISC SEISMIC PROVISIONS FOR STRUCTURAL STEEL BUILDINGS

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PB262-FM PB262/Seismic.cls December 18, 2002 15:40 Char Count=

ANSI/AISC 341-02An American National Standard

Seismic Provisions forStructural Steel Buildings

May 21, 2002

Supersedes the Seismic Provisionsfor Structural Steel Buildings

dated April 15, 1997including Supplements No. 1 and 2

and all previous versions

Approved by theAISC Committee on Specifications andissued by the AISC Board of Directors

AMERICAN INSTITUTE OF STEEL CONSTRUCTION, INC.One East Wacker Drive, Suite 3100

Chicago, Illinois 60601-2000

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Copyright C© 2002

by

American Institute of Steel Construction, Inc.

All rights reserved. This book or any part thereof must not be reproduced in anyform without the written permission of the publisher.

The AISC logo is a registered trademark of AISC and is used under license.

The information presented in this publication has been prepared in accordance withrecognized engineering principles and is for general information only. While it isbelieved to be accurate, this information should not be used or relied upon for anyspecific application without competent professional examination and verification ofits accuracy, suitability, and applicability by a licensed engineer, architect or otherprofessional. The publication of the material contained herein is not intended as arepresentation or warranty on the part of the American Institute of Steel Construc-tion, Inc. or of any other person named herein, that this information is suitable for anygeneral or particular use or of freedom from infringement of any patent or patents.Anyone making use of this information assumes all liability arising from such use.

Caution must be exercised when relying upon other specifications and codes devel-oped by other bodies and incorporated by reference herein since such material maybe modified or amended from time to time subsequent to the printing of this edition.The American Institute of Steel Construction, Inc. bears no responsibility for suchmaterial other than to refer to it and incorporate it by reference at the time of theinitial publication of this edition.

Seismic Provisions for Structural Steel Buildings, May 21, 2002American Institute of Steel Construction

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DEDICATION

Professor Egor Popov

This edition of the AISC Seismic Provisions is dedicated to the memory of Professor EgorPopov. Professor Popov was a Professor for over 50 years at the University of Californiaat Berkeley, and a long time member of the AISC Committee on Specifications. ProfessorPopov focused a major portion of his career improving the understanding and seismic per-formance of steel structures. He was instrumental in the development of seismic designprovisions for steel structures for over thirty years, and initiated the activity of AISC inthis regard in the late 1980’s. As Chair of TC113 (the predecessor of TC9), he led thepublication of the first two editions of the AISC Seismic Provisions. Until the time of hisdeath at the age of 88 early in 2001, Professor Popov remained a very active member ofTC9 in the role of Vice Chair. His contributions to the development of these provisionsand understanding of the seismic performance of steel buildings is unequaled, and willlong be remembered and appreciated by AISC, the steel industry and the structural engi-neering profession. It is entirely fitting that these provisions be dedicated to the memoryof Professor Egor Popov.


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PREFACE

(This Preface is not a part of ANSI/AISC 341–02, Seismic Provisions for Structural SteelBuildings, but is included for information purposes only.)

The AISC Load and Resistance Factor Design (LRFD) Specification for Structural SteelBuildings is intended to cover the common design criteria in routine office practice. Ac-cordingly, it is not feasible for it to also cover the many special and unique problemsencountered within the full range of structural design practice. This document, the AISCSeismic Provisions for Structural Steel Buildings (hereafter referred to as Seismic Pro-visions) is a separate consensus document that addresses one such topic: the design andconstruction of structural steel and composite structural steel/reinforced concrete buildingsystems for seismic demands.

These Provisions are presented in three parts: Part I is intended for the design and con-struction of structural steel buildings, using LRFD; Part II is intended for the designand construction of composite structural steel/reinforced concrete buildings; Part III isan allowable stress design alternative to the LRFD provisions for structural steel build-ings in Part I. In addition, three appendices, a list of Symbols, a Glossary, and a non-mandatory Commentary with background information are provided. The first letter(s)of words or terms that appear in the glossary are generally capitalized throughout theseProvisions.

The previous edition of the AISC Seismic Provisions for Structural Steel Buildings, pub-lished on April 15, 1997, incorporated many of the early advances achieved as part of theFEMA/SAC program and other investigations and developments related to the seismicdesign of steel buildings. Recognizing that rapid and significant changes in the knowl-edge base were occurring for the seismic design of steel buildings, especially MomentFrames, the AISC Specifications Committee committed to generating frequent supple-ments to the Seismic Provisions. This commitment was intended to keep the provi-sions as current as possible. The first such supplement was completed and publishedon February 15, 1999, Supplement No. 1 to the 1997 AISC Seismic Provisions. Supple-ment Number 2 to the 1997 AISC Seismic Provisions was published on November 10,2000.

This edition of the AISC Seismic Provisions incorporates Supplements No. 1 (February15, 1999) and No. 2 (November 10, 2000) to the 1997 Seismic Provisions. This versionalso includes Errata to Sections 8.4 and 9.9. Additional revisions resulted from con-sidering new information generated by the FEMA/SAC project, which culminated latein 2000, and other sources. These provisions were also modified to be consistent withthe ASCE 7-02 document, Minimum Design Loads for Buildings and Other Structures.This allows these provisions to be incorporated by reference into both the 2003 IBC and2002 NFPA 5000 building codes that use ASCE 7-02 as their basis for design loadings.Because the scope of changes that have been made to these provisions since 1997 is solarge, they are being republished in their entirety. A major update to the commentaryto these provisions is also provided. Specific changes to these provisions include the


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vi PREFACE

following:

A clarification to the glossary to verify that chord and collector/drag elementsin floor diaphragms are considered to be part of the Seismic Load ResistingSystem.

Additional requirements for the toughness of filler metals to be used in complete-joint-penetration groove welds in intermediate and Special Moment Frame sys-tems.

A revision to clarify member slenderness ratio requirements and better coordi-nate with the LRFD provisions.

Increasing the Moment Frame column splice requirements to reflect the FEMA/SAC recommendations.

Requiring that splices of columns that are not part of the Moment Frames developa minimum shear force.

Clarifying Column Base design demands for various systems. Adding a section on the use of H-pile members. Clarifying lateral bracing requirements of Moment Frame beams, including the

provision of a required stiffness to be consistent with Section 3 of LRFD. Increasing SMF web Connection design requirements to be consistent with the

FEMA/SAC recommendations. Adding a new appendix (Appendix P) that defines procedures to be used in the

pre-qualification of moment Connections. Incorporating FEMA/SAC recommendations for weld access holes in OMF

systems. Incorporating FEMA/SAC recommendations for the removal of weld back-

ing and run-off tabs in OMF systems, including grinding surfaces to adequatesmoothness.

Dual units format. Values and equations are given in both U.S. customaryand metric units. The metric conversions (given in parentheses following theU.S. units) are based on IEEE/ASTM SI 10, Standard for Use of the Inter-national System of Units (SI): The Modern Metric System. The equations arenon-dimensionalized where possible by factoring out material constants, suchas E .

The AISC Committee on Specifications, Task Committee 9—Seismic Provisions is re-sponsible for the ongoing development of these Provisions. The AISC Committee onSpecifications gives final approval of the document through an ANSI accredited ballot-ing process, and has enhanced these Provisions through careful scrutiny, discussion, andsuggestions for improvement. AISC further acknowledges the significant contributions ofseveral groups to the completion of this document: the Building Seismic Safety Council(BSSC), the SAC Joint Venture, the Federal Emergency Management Agency (FEMA),the National Science Foundation (NSF), and the Structural Engineers Association ofCalifornia (SEAOC).


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The reader is cautioned that professional judgment must be exercised when data or rec-ommendations in these provisions are applied, as described more fully in the disclaimernotice preceding the Preface.

By, the members of AISC Committee on Specifications, Task Committee 9 – SeismicDesign:

James O. Malley, Chair James R. HarrisMark Saunders, Vice-Chairman Patrick M. HassettRoy Becker Roberto T. LeonGregory G. Deierlein Robert LyonsRichard M. Drake Harry W. MartinMichael D. Engelhardt Clarkson W. PinkhamRoger E. Ferch Rafael SabelliTimothy P. Fraser Thomas A. SabolSubhash Goel Kurt D. SwenssonJohn L. Gross Nabih F. G. Youssef

Cynthia J. Lanz, Secretary

Approved by the AISC Committee on Specifications,

Stanley D. Lindsey, Chairman John L. GrossRoger E. Ferch, Vice-Chairman James R. HarrisHansraj G. Ashar Tony C. HazelWilliam F. Baker Mark V. HollandJohn M. Barsom Lawrence A. KloiberWillam D. Bast Roberto T. LeonReidar Bjorhovde James O. MalleyRoger L. Brockenbrough Richard W. MarshallWai-Fah Chen Harry W. MartinGregory G. Deierlein David L. McKenzieDuane S. Ellifritt Duane K. MillerBruce R. Ellingwood Thomas M. MurrayShu-Jin Fang R. Shankar NairSteven J. Fenves Jack E. PetersenJames M. Fisher Douglas A. Rees-EvansJohn W. Fisher Donald R. ShermanTimothy P. Fraser W. Lee ShoemakerTheodore V. Galambos William A. ThorntonLouis F. Geschwindner Raymond H. R. TideLawrence G. Griffis Joseph A. Yura

Cynthia J. Lanz, Secretary


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TABLE OF CONTENTS

SYMBOLS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . xix

PART I STRUCTURAL STEEL BUILDINGS

GLOSSARY . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1

1. SCOPE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5

2. REFERENCED SPECIFICATIONS, CODES, AND STANDARDS . . . . 5

3. GENERAL SEISMIC DESIGN REQUIREMENTS . . . . . . . . . . . . . . 7

4. LOADS, LOAD COMBINATIONS, AND NOMINAL STRENGTHS . . . 74.1. Loads and Load Combinations . . . . . . . . . . . . . . . . . . . . . . . . 74.2. Nominal Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7

5. STORY DRIFT . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 7

6. MATERIALS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 76.1. Material Specifications . . . . . . . . . . . . . . . . . . . . . . . . . . . . 76.2. Material Properties for Determination of Required Strength . . . . . . . 86.3. Notch-Toughness Requirements . . . . . . . . . . . . . . . . . . . . . . . 8

7. CONNECTIONS, JOINTS, AND FASTENERS . . . . . . . . . . . . . . . . 97.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 97.2. Bolted Joints . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 97.3. Welded Joints . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 97.4. Other Connections . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 10

8. MEMBERS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 108.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 108.2. Local Buckling . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 108.3. Column Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 128.4. Column Splices . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 128.5. Column Bases . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 138.6. H-Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 13

9. SPECIAL MOMENT FRAMES (SMF) . . . . . . . . . . . . . . . . . . . . . 139.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 139.2. Beam-to-Column Joints and Connections . . . . . . . . . . . . . . . . . . 149.3. Panel Zone of Beam-to-Column Connections . . . . . . . . . . . . . . . . 149.4. Beam and Column Limitations . . . . . . . . . . . . . . . . . . . . . . . . 169.5. Continuity Plates . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 169.6. Column-Beam Moment Ratio . . . . . . . . . . . . . . . . . . . . . . . . . 169.7. Beam-to-Column Connection Restraint . . . . . . . . . . . . . . . . . . . 179.8. Lateral Bracing of Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . 189.9. Column Splices . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 18


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10. INTERMEDIATE MOMENT FRAMES (IMF) . . . . . . . . . . . . . . . . 1810.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1810.2. Beam-to-Column Joints and Connections . . . . . . . . . . . . . . . . . . 1910.3. Panel Zone of Beam-to-Column Connections . . . . . . . . . . . . . . . . 1910.4. Beam and Column Limitations . . . . . . . . . . . . . . . . . . . . . . . . 2010.5. Continuity Plates . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2010.6. Column-Beam Moment Ratio . . . . . . . . . . . . . . . . . . . . . . . . . 2010.7. Beam-to-Column Connection Restraint . . . . . . . . . . . . . . . . . . . 2010.8. Lateral Bracing of Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . 2010.9. Column Splices . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 20

11. ORDINARY MOMENT FRAMES (OMF) . . . . . . . . . . . . . . . . . . . 2011.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2011.2. Beam-to-Column Joints and Connections . . . . . . . . . . . . . . . . . . 2011.3. Panel Zone of Beam-to-Column Connections . . . . . . . . . . . . . . . . 2111.4. Beam and Column Limitations . . . . . . . . . . . . . . . . . . . . . . . . 2111.5. Continuity Plates . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2211.6. Column-Beam Moment Ratio . . . . . . . . . . . . . . . . . . . . . . . . . 2311.7. Beam-to-Column Connection Restraint . . . . . . . . . . . . . . . . . . . 2311.8. Lateral Bracing of Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . 2311.9. Column Splices . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 23

12. SPECIAL TRUSS MOMENT FRAMES (STMF) . . . . . . . . . . . . . . . 2312.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2312.2. Special Segment . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2312.3. Nominal Strength of Special Segment Members . . . . . . . . . . . . . . 2412.4. Nominal Strength of Non-Special Segment Members . . . . . . . . . . . 2412.5. Compactness . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2512.6. Lateral Bracing . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 25

13. SPECIAL CONCENTRICALLY BRACED FRAMES (SCBF) . . . . . . . 2513.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2513.2. Bracing Members . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2513.3. Bracing Connections . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2613.4. Special Bracing Configuration Requirements . . . . . . . . . . . . . . . . 2713.5. Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 27

14. ORDINARY CONCENTRICALLY BRACED FRAMES (OCBF) . . . . . 2814.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2814.2. Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 28

15. ECCENTRICALLY BRACED FRAMES (EBF) . . . . . . . . . . . . . . . . 2815.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2815.2. Links . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2815.3. Link Stiffeners . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 2915.4. Link-to-Column Connections . . . . . . . . . . . . . . . . . . . . . . . . . 3015.5. Lateral Bracing of Link . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3115.6. Diagonal Brace and Beam Outside of Link . . . . . . . . . . . . . . . . . 3115.7. Beam-to-Column Connections . . . . . . . . . . . . . . . . . . . . . . . . 3215.8. Required Column Strength . . . . . . . . . . . . . . . . . . . . . . . . . . 32

16. QUALITY ASSURANCE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 32


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APPENDIX P. PREQUALIFICATION OF BEAM-TO-COLUMN ANDLINK-TO-COLUMN CONNECTIONS

P1. SCOPE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 33

P2. GENERAL REQUIREMENTS . . . . . . . . . . . . . . . . . . . . . . . . . . 33P2.1. Basis for Prequalification . . . . . . . . . . . . . . . . . . . . . . . . . . . 33P2.2. Authority for Prequalification . . . . . . . . . . . . . . . . . . . . . . . . . 33

P3. TESTING REQUIREMENTS . . . . . . . . . . . . . . . . . . . . . . . . . . . 33

P4. PREQUALIFICATION VARIABLES . . . . . . . . . . . . . . . . . . . . . . 34

P5. DESIGN PROCEDURE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 35

P6. PREQUALIFICATION RECORD . . . . . . . . . . . . . . . . . . . . . . . . 35

APPENDIX S. QUALIFYING CYCLIC TESTS OF BEAM-TO-COLUMNAND LINK-TO-COLUMN CONNECTIONS

S1. SCOPE AND PURPOSE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 37

S2. SYMBOLS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 37

S3. DEFINITIONS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 37

S4. TEST SUBASSEMBLAGE REQUIREMENTS . . . . . . . . . . . . . . . . 38

S5. ESSENTIAL TEST VARIABLES . . . . . . . . . . . . . . . . . . . . . . . . 38S5.1. Sources of Inelastic Rotation . . . . . . . . . . . . . . . . . . . . . . . . . 38S5.2. Size of Members . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 39S5.3. Connection Details . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 39S5.4. Continuity Plates . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 39S5.5. Material Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 39S5.6. Welds . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 39S5.7. Bolts . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 40

S6. LOADING HISTORY . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 40S6.1. General Requirements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 40S6.2. Loading Sequence for Beam-to-Column Moment Connections . . . . . . 41S6.3. Loading Sequence for Link-to-Column Connections . . . . . . . . . . . . 41

S7. INSTRUMENTATION . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 41

S8. MATERIALS TESTING REQUIREMENTS . . . . . . . . . . . . . . . . . . 42S8.1. Tension Testing Requirements . . . . . . . . . . . . . . . . . . . . . . . . 42S8.2. Methods of Tension Testing . . . . . . . . . . . . . . . . . . . . . . . . . . 42

S9. TEST REPORTING REQUIREMENTS . . . . . . . . . . . . . . . . . . . . 42

S10. ACCEPTANCE CRITERIA . . . . . . . . . . . . . . . . . . . . . . . . . . . . 43

APPENDIX X. WELD METAL/WELDING PROCEDURESPECIFICATION TOUGHNESS VERIFICATION TEST

X1. SCOPE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 44

X2. TEST CONDITIONS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 44


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X3. TEST SPECIMENS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 45

X4. ACCEPTANCE CRITERIA . . . . . . . . . . . . . . . . . . . . . . . . . . . . 45

PART II COMPOSITE STRUCTURAL STEEL AND REINFORCEDCONCRETE BUILDINGS

GLOSSARY . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 47

1. SCOPE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 49


3. SEISMIC DESIGN CATEGORIES . . . . . . . . . . . . . . . . . . . . . . . 49

4. LOADS, LOAD COMBINATIONS, AND NOMINAL STRENGTHS . . . 49

5. MATERIALS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 495.1. Structural Steel . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 495.2. Concrete and Steel Reinforcement . . . . . . . . . . . . . . . . . . . . . . 50

6. COMPOSITE MEMBERS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 506.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 506.2. Composite Floor and Roof Slabs . . . . . . . . . . . . . . . . . . . . . . . 506.3. Composite Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 516.4. Reinforced-Concrete-Encased Composite Columns . . . . . . . . . . . . 516.5. Concrete-Filled Composite Columns . . . . . . . . . . . . . . . . . . . . . 55

7. COMPOSITE CONNECTIONS . . . . . . . . . . . . . . . . . . . . . . . . . 557.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 557.2. General Requirements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 567.3. Nominal Strength of Connections . . . . . . . . . . . . . . . . . . . . . . 56

8. COMPOSITE PARTIALLY RESTRAINED (PR) MOMENT FRAMES(C-PRMF) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 578.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 578.2. Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 588.3. Composite Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 588.4. Partially Restrained (PR) Moment Connections . . . . . . . . . . . . . . . 58

9. COMPOSITE SPECIAL MOMENT FRAMES (C-SMF) . . . . . . . . . . 589.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 589.2. Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 589.3. Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 589.4. Moment Connections . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 589.5. Column-Beam Moment Ratio . . . . . . . . . . . . . . . . . . . . . . . . . 59

10. COMPOSITE INTERMEDIATE MOMENT FRAMES (C-IMF) . . . . . . 5910.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5910.2. Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5910.3. Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 5910.4. Moment Connections . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 59

11. COMPOSITE ORDINARY MOMENT FRAMES (C-OMF) . . . . . . . . 6011.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6011.2. Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 60


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11.3. Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6011.4. Moment Connections . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 60

12. COMPOSITE ORDINARY BRACED FRAMES (C-OBF) . . . . . . . . . . 6012.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6012.2. Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6012.3. Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6112.4. Braces . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6112.5. Connections . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 61

13. COMPOSITE CONCENTRICALLY BRACED FRAMES (C-CBF) . . . . 6113.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6113.2. Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6113.3. Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6113.4. Braces . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6113.5. Bracing Connections . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 62

14. COMPOSITE ECCENTRICALLY BRACED FRAMES (C-EBF) . . . . . 6214.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6214.2. Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6214.3. Links . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6214.4. Braces . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6214.5. Connections . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 62

15. ORDINARY REINFORCED CONCRETE SHEAR WALLSCOMPOSITE WITH STRUCTURAL STEEL ELEMENTS(C-ORCW) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6315.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6315.2. Boundary Members . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6315.3. Coupling Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 63

16. SPECIAL REINFORCED CONCRETE SHEAR WALLSCOMPOSITE WITH STRUCTURAL STEEL ELEMENTS(C-SRCW) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6416.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6416.2. Boundary Members . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6416.3. Coupling Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 64

17. COMPOSITE STEEL PLATE SHEAR WALLS (C-SPW) . . . . . . . . . . 6517.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6517.2. Wall Elements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6517.3. Boundary Members . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 65

PART III ALLOWABLE STRESS DESIGN (ASD) ALTERNATIVE

1. SCOPE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 67


4. LOADS, LOAD COMBINATIONS, AND NOMINAL STRENGTHS . . . 674.2. Nominal Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 684.3. Design Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 69

7. CONNECTIONS, JOINTS, AND FASTENERS . . . . . . . . . . . . . . . . 697.2. Bolted Joints . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 69


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9. SPECIAL MOMENT FRAMES . . . . . . . . . . . . . . . . . . . . . . . . . 699.3. Panel Zone of Beam-to-Column Connections . . . . . . . . . . . . . . . 699.7. Beam-to-Column Connection Restraint . . . . . . . . . . . . . . . . . . 70

12. SPECIAL TRUSS MOMENT FRAMES . . . . . . . . . . . . . . . . . . . . 7012.4. Nominal Strength of Non-Special Segment Members . . . . . . . . . . 7012.6. Lateral Bracing . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 70

13. SPECIAL CONCENTRICALLY BRACED FRAMES (SCBF) . . . . . . . 71

14. ORDINARY CONCENTRICALLY BRACED FRAMES (OCBF) . . . . . 7114.2. Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 71

COMMENTARY . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 73

PART I STRUCTURAL STEEL BUILDINGS

C1. SCOPE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 75

C2. REFERENCED SPECIFICATIONS, CODES, AND STANDARDS . . . . 76

C3. GENERAL SEISMIC DESIGN REQUIREMENTS . . . . . . . . . . . . . . 76

C4. LOADS, LOAD COMBINATIONS, AND NOMINAL STRENGTH . . . . 77

C5. STORY DRIFT . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 79

C6. MATERIALS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 82C6.1. Material Specifications . . . . . . . . . . . . . . . . . . . . . . . . . . . . 82C6.2. Material Properties for Determination of Required Strength . . . . . . . 82C6.3. Notch-Toughness Requirements . . . . . . . . . . . . . . . . . . . . . . 83

C7. CONNECTIONS, JOINTS, AND FASTENERS . . . . . . . . . . . . . . . . 85C7.2. Bolted Joints . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 85C7.3. Welded Joints . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 86C7.4. Other Connections . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 87

C8. MEMBERS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 87C8.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 87C8.2. Local Buckling . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 88C8.3. Column Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 88C8.4. Column Splices . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 89C8.5. Column Bases . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 91C8.6. H-Piles . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 94

C9. SPECIAL MOMENT FRAMES (SMF) . . . . . . . . . . . . . . . . . . . . . 96C9.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 97C9.2. Beam-to-Column Joints and Connections . . . . . . . . . . . . . . . . . 98C9.3. Panel Zone of Beam-to-Column Connections . . . . . . . . . . . . . . . 99C9.4. Beam and Column Limitations . . . . . . . . . . . . . . . . . . . . . . 102C9.5. Continuity Plates . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 102C9.6. Column-Beam Moment Ratio . . . . . . . . . . . . . . . . . . . . . . . 104C9.7. Beam-to-Column Connection Restraint . . . . . . . . . . . . . . . . . 105C9.8. Lateral Bracing of Beams . . . . . . . . . . . . . . . . . . . . . . . . . 106

C10. INTERMEDIATE MOMENT FRAMES (IMF) . . . . . . . . . . . . . . . 107C10.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 107C10.2. Beam-to-Column Joints and Connections . . . . . . . . . . . . . . . . 107


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C11. ORDINARY MOMENT FRAMES (OMF) . . . . . . . . . . . . . . . . . . 108C11.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 108C11.2. Beam-to-Column Joints and Connections . . . . . . . . . . . . . . . . 108C11.5. Continuity Plates . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 110

C12. SPECIAL TRUSS MOMENT FRAMES (STMF) . . . . . . . . . . . . . . 110C12.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 110C12.2. Special Segment . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 111C12.3. Nominal Strength of Special Segment Members . . . . . . . . . . . . 112C12.4. Nominal Strength of Non-Special Segment Members . . . . . . . . . 113C12.5. Compactness . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 113C12.6. Lateral Bracing . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 113

C13. SPECIAL CONCENTRICALLY BRACED FRAMES (SCBF) . . . . . . 113C13.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 113C13.2. Bracing Members . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 117C13.3. Bracing Connections . . . . . . . . . . . . . . . . . . . . . . . . . . . . 118C13.4. Special Bracing Configuration Special Requirements . . . . . . . . . 120C13.5. Columns . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 120

C14. ORDINARY CONCENTRICALLY BRACED FRAMES (OCBF) . . . . 121C14.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 121C14.2. Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 121

C15. ECCENTRICALLY BRACED FRAMES (EBF) . . . . . . . . . . . . . . . 123C15.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 123C15.2. Links . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 125C15.3. Link Stiffeners . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 127C15.4. Link-to-Column Connections . . . . . . . . . . . . . . . . . . . . . . . 128C15.5. Lateral Bracing of the Link . . . . . . . . . . . . . . . . . . . . . . . . 129C15.6. Diagonal Brace and Beam Outside of Links . . . . . . . . . . . . . . . 129C15.7. Beam-to-Column Connections . . . . . . . . . . . . . . . . . . . . . . 132C15.8. Required Column Strength . . . . . . . . . . . . . . . . . . . . . . . . 132

C16. QUALITY ASSURANCE . . . . . . . . . . . . . . . . . . . . . . . . . . . . 134

APPENDIX P. PREQUALIFICATION OF BEAM-TO-COLUMN ANDLINK-TO-COLUMN CONNECTIONS

CP1. SCOPE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 136CP2.1. Basis for Prequalification . . . . . . . . . . . . . . . . . . . . . . . . . 137CP2.2. Authority for Prequalification . . . . . . . . . . . . . . . . . . . . . . . 137

CP3. TESTING REQUIREMENTS . . . . . . . . . . . . . . . . . . . . . . . . . . 138

CP4. PREQUALIFICATION VARIABLES . . . . . . . . . . . . . . . . . . . . . 139

CP5. DESIGN PROCEDURE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 139

CP6. PREQUALIFICATION RECORD . . . . . . . . . . . . . . . . . . . . . . . 139

APPENDIX S. QUALIFYING CYCLIC TESTS OF BEAM-TO-COLUMNAND LINK-TO-COLUMN CONNECTIONS

CS1. SCOPE AND PURPOSE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 140

CS3. DEFINITIONS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 141


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CS4. TEST SUBASSEMBLAGE REQUIREMENTS . . . . . . . . . . . . . . . 142

CS5. ESSENTIAL TEST VARIABLES . . . . . . . . . . . . . . . . . . . . . . . 142CS5.1. Sources of Inelastic Rotation . . . . . . . . . . . . . . . . . . . . . . 142CS5.2. Size of Members . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 143CS5.5. Material Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 144CS5.6. Welds . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 145

CS6. LOADING HISTORY . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 145

CS8. MATERIALS TESTING REQUIREMENTS . . . . . . . . . . . . . . . . 146

CS10. ACCEPTANCE CRITERIA . . . . . . . . . . . . . . . . . . . . . . . . . . 147

APPENDIX X. WELD METAL/WELDING PROCEDURESPECIFICATION TOUGHNESS VERIFICATION TEST . . . . . . . . . . . . 148

PART II COMPOSITE STRUCTURAL STEEL AND REINFORCEDCONCRETE BUILDINGS

C1. SCOPE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 151

C2. REFERENCED SPECIFICATIONS, CODES, AND STANDARDS . . . 152

C3. SEISMIC DESIGN CATEGORIES . . . . . . . . . . . . . . . . . . . . . . 152

C4. LOADS, LOAD COMBINATIONS, AND NOMINAL STRENGTHS . . 152

C5. MATERIALS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 152

C6. COMPOSITE MEMBERS . . . . . . . . . . . . . . . . . . . . . . . . . . . 153C6.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 153C6.2. Composite Floor and Roof Slabs . . . . . . . . . . . . . . . . . . . . . 153C6.3. Composite Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 154C6.4. Reinforced-Concrete-Encased Composite Columns . . . . . . . . . . 155C6.5. Concrete-Filled Composite Columns . . . . . . . . . . . . . . . . . . . 158

C7. COMPOSITE CONNECTIONS . . . . . . . . . . . . . . . . . . . . . . . . 158C7.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 158C7.2. General Requirements . . . . . . . . . . . . . . . . . . . . . . . . . . . 159C7.3. Nominal Strength of Connections . . . . . . . . . . . . . . . . . . . . 160

C8. COMPOSITE PARTIALLY RESTRAINED (PR) MOMENT FRAMES(C-PRMF) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 163

C9. COMPOSITE SPECIAL MOMENT FRAMES (C-SMF) . . . . . . . . . 164C9.1. Scope . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 164C9.3. Beams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 164C9.4. Moment Connections . . . . . . . . . . . . . . . . . . . . . . . . . . . 165

C10. COMPOSITE INTERMEDIATE MOMENT FRAMES (C-IMF) . . . . 167

C11. COMPOSITE ORDINARY MOMENT FRAMES (C-OMF) . . . . . . . 167

C12. COMPOSITE ORDINARY BRACED FRAMES (C-OBF) . . . . . . . . 167

C13. COMPOSITE CONCENTRICALLY BRACED FRAMES (C-CBF) . . 167

C14. COMPOSITE ECCENTRICALLY BRACED FRAMES (C-EBF) . . . 169


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C15. ORDINARY REINFORCED CONCRETE SHEAR WALLSCOMPOSITE WITH STRUCTURAL STEEL ELEMENTS(C-ORCW) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 169

C16. SPECIAL REINFORCED CONCRETE SHEAR WALLSCOMPOSITE WITH STRUCTURAL STEEL ELEMENTS(C-SRCW) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 173

C17. COMPOSITE STEEL PLATE SHEAR WALLS (C-SPW) . . . . . . . . . 174

PART III ALLOWABLE STRESS DESIGN (ASD) ALTERNATIVE

C1. SCOPE . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 177C4.2. Nominal Strength . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 177

REFERENCES . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 179


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SYMBOLS

Numbers in parentheses after the definition of a symbol refer to the Section in either Part Ior II of these Provisions in which the symbol is first used.

Symbol Definition SectionA f Flange area, in.2 (mm2) . . . . . . . . . . . . . . . . . . . . . . (I-8)Ag Gross area, in.2 (mm2) . . . . . . . . . . . . . . . . . . . . . . (I-9)As Cross-sectional area of structural steel elements in composite

members, in.2 (mm2) . . . . . . . . . . . . . . . . . . . . . . . (II-6)Ash Minimum area of tie reinforcement, in.2 (mm2) . . . . . . . . (II-6)Asp Horizontal area of the steel plate in composite shear wall,

in.2 (mm2) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . (II-17)Ast Area of Link stiffener, in.2 (mm2) . . . . . . . . . . . . . . . . (I-15)Aw Link web area, in.2 (mm2) . . . . . . . . . . . . . . . . . . . . (I-15)D Dead load due to the weight of the structural elements and

permanent features on the building, kips (N) . . . . . . . . . . (I-9)Outside diameter of round HSS, in. (mm) . . . . . . . . . . . . (Table I-8-1)

E Effect of horizontal and vertical earthquake-induced loads . . (Glossary)Es Modulus of elasticity of steel, Es = 29,000 ksi (200 000 MPa) (I-8, II-6)Es I Flexural elastic stiffness of the chord members of the special

segment, kip-in.2 (N-mm2) . . . . . . . . . . . . . . . . . . . . (I-12)Fy Specified minimum yield stress of the type of steel to be used,

ksi (MPa). As used in the LRFD Specification, “yield stress”denotes either the minimum specified yield point (for thosesteels that have a yield point) or the specified yield strength(for those steels that do not have a yield point). . . . . . . . . . (I-6)

Fyb Fy of a beam, ksi (MPa) . . . . . . . . . . . . . . . . . . . . . (I-9)Fyc Fy of a column, ksi (MPa) . . . . . . . . . . . . . . . . . . . . (I-9)Fy f Fy of column flange, ksi (MPa) . . . . . . . . . . . . . . . . . (I-8)Fyh Specified minimum yield strength of transverse reinforcement,

ksi (MPa) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . (II-6)Fyw Fy of the Panel Zone steel, ksi (MPa) . . . . . . . . . . . . . .Fu Specified minimum tensile strength, ksi (MPa) . . . . . . . . . (I-7)H Height of story, which may be taken as the distance between the

centerline of floor framing at each of the levels above andbelow, or the distance between the top of floor slabs at each ofthe levels above and below, in. (mm) . . . . . . . . . . . . . . (I-8)

K Effective length factor for prismatic member . . . . . . . . . . (I-13)L Live load due to occupancy and moveable equipment,

kips (kN) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . (I-9)Span length of the truss, in. (mm) . . . . . . . . . . . . . . . . (I-12)


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xx SYMBOLS

L p Limiting laterally unbraced length for full plastic flexuralstrength, uniform moment case, in. (mm) . . . . . . . . . . . . (I-12)

Ls Length of the special segment, in. (mm) . . . . . . . . . . . . (I-12)Mn Nominal flexural strength, kip-in. (N-mm) . . . . . . . . . . . (I-11)Mnc Nominal flexural strength of the chord member of the special

segment, kip-in. (N-mm) . . . . . . . . . . . . . . . . . . . . . (I-12)Mp Nominal plastic flexural strength, kip-in. (N-mm) . . . . . . . (I-9)Mpa Nominal plastic flexural strength modified by axial load, kip-in.

(N-mm) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . (I-15)Mpc Nominal plastic flexural strength of the column, kip-in.

(N-mm) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . (I-8)Mv Additional moment due to shear amplification from the location

of the plastic hinge to the column centerline, kip-in.(N-mm) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . (I-9)

Mu Required flexural strength of a member or Joint, kip-in.(N-mm) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . (I-9)

Pn Nominal axial strength of a column, kips (N) . . . . . . . . . . (I-8)Nominal axial strength of a Composite Column, kips (N) . . . (II-6)

Pnc Nominal axial compressive strength of diagonal members of thespecial segment, kips (N) . . . . . . . . . . . . . . . . . . . . . (I-12)

Pnt Nominal axial tensile strength of diagonal members of thespecial segment, kips (N) . . . . . . . . . . . . . . . . . . . . . (I-12)

Po Nominal axial strength of a Composite Column at zeroeccentricity, kips (N). . . . . . . . . . . . . . . . . . . . . . . . (II-6)

Pu Required axial strength of a column or a Link, kips (N) . . . . (I-8)Required axial strength of a Composite Column, kips (N) . . . (II-9)

Puc Required axial strength of a column in compression, kips (N) . (I-9)Py Nominal axial yield strength of a member, which is equal to

Fy Ag , kips (N) . . . . . . . . . . . . . . . . . . . . . . . . . . . (I-9)Qb Maximum unbalanced vertical load effect applied to a beam by

the braces, kips (N) . . . . . . . . . . . . . . . . . . . . . . . . (I-13)Rn Nominal Strength . . . . . . . . . . . . . . . . . . . . . . . . .Ru Required strength . . . . . . . . . . . . . . . . . . . . . . . . . (I-9)Rv Panel Zone nominal shear strength . . . . . . . . . . . . . . . . (I-9)Ry Ratio of the Expected Yield Strength to the minimum specified

yield strength Fy . . . . . . . . . . . . . . . . . . . . . . . . . . (I-6)S Snow load, kips (N) . . . . . . . . . . . . . . . . . . . . . . . . (I-9)Vn Nominal shear strength of a member, kips (N) . . . . . . . . . (I-15)Vns Nominal shear strength of the steel plate in composite plate

shear walls, kips (N) . . . . . . . . . . . . . . . . . . . . . . . . (II-17)Vp Nominal shear strength of an active Link, kips (N) . . . . . . . (I-15)Vpa Nominal shear strength of an active Link modified by the axial

load magnitude, kips (N) . . . . . . . . . . . . . . . . . . . . . (I-15)Vu Required shear strength of a member, kips (N) . . . . . . . . . (I-9)Ycon Distance from top of steel beam to top of concrete slab or

encasement, in. (mm) . . . . . . . . . . . . . . . . . . . . . . . (II-6)Z Plastic section modulus of a member, in.3 (mm3) . . . . . . . . (I-9)Zb Plastic section modulus of the beam, in.3 (mm3) . . . . . . . . (I-9)Zc Plastic section modulus of the column, in.3 (mm3) . . . . . . . (I-9)a Angle that diagonal members make with the horizontal . . . . (I-12)


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SYMBOLS xxi

b Width of compression element as defined in LRFDSpecification Section B5.1, in. (mm) . . . . . . . . . . . . . . . (Table I-8-1)

bcf Width of column flange, in. (mm) . . . . . . . . . . . . . . . . (I-9)b f Flange width, in. (mm) . . . . . . . . . . . . . . . . . . . . . . (I-9)bw Width of the concrete cross-section minus the width of the

structural shape measured perpendicular to the direction ofshear, in. (mm) . . . . . . . . . . . . . . . . . . . . . . . . . . . (II-6)

d Nominal fastener diameter, in. (mm) . . . . . . . . . . . . . . . (I-7)db Overall beam depth, in. (mm). . . . . . . . . . . . . . . . . . . (I-9)dc Overall column depth, in. (mm) . . . . . . . . . . . . . . . . . (I-9)dz Overall Panel Zone depth between Continuity Plates, in. (mm) (I-9)e EBF Link length, in. (mm) . . . . . . . . . . . . . . . . . . . . (I-15)f ′c Specified compressive strength of concrete, ksi (MPa) . . . . . (II-6)

hcc Cross-sectional dimension of the confined core region inComposite Columns measured center-to-center of thetransverse reinforcement, in. (mm) . . . . . . . . . . . . . . . . (II-6)

l Unbraced length between stitches of built-up bracing members,in. (mm). . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . (I-13)Unbraced length of compression or bracing member, in. (mm) (I-13)

r Governing radius of gyration, in. (mm) . . . . . . . . . . . . . (I-9)ry Radius of gyration about y axis, in. (mm) . . . . . . . . . . . . (I-9)s Spacing of transverse reinforcement measured along the

longitudinal axis of the structural composite member, in. (mm) (II-6)t Thickness of connected part, in. (mm) . . . . . . . . . . . . . . (I-7)

Thickness of element, in. (mm). . . . . . . . . . . . . . . . . . (Table I-8-1)Thickness of column web or doubler plate, in. (mm) . . . . . . (I-9)

tb f Thickness of beam flange, in. (mm) . . . . . . . . . . . . . . . (I-9)tc f Thickness of column flange, in. (mm) . . . . . . . . . . . . . . (I-9)t f Thickness of flange, in. (mm) . . . . . . . . . . . . . . . . . . . (I-9)tp Thickness of Panel Zone including doubler plates, in. (mm) . . (I-9)tw Thickness of web, in. (mm) . . . . . . . . . . . . . . . . . . . . (Table I-8-1)wz Width of Panel Zone between column flanges, in. (mm) . . . . (I-9)zb Minimum plastic section modulus at the Reduced Beam

Section, in.3 (mm3) . . . . . . . . . . . . . . . . . . . . . . . . (I-9)M*pc Moment at beam and column centerline determined by

projecting the sum of the nominal column plastic momentstrength, reduced by the axial stress Puc/Ag , from the top andbottom of the beam moment connection . . . . . . . . . . . . . (I-9)

M*pb Moment at the intersection of the beam and column centerlinesdetermined by projecting the beam maximum developedmoments from the column face. Maximum developed momentsshall be determined from test results . . . . . . . . . . . . . . . (I-9)

o Horizontal seismic overstrength factor . . . . . . . . . . . . . . (I-4) Deformation quantity used to control loading of the Test

Specimen . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . (S6)y Value of deformation quantity at first significant yield of Test

Specimen . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . (S6) ′ Ratio of required axial force Pu to required shear strength Vu of

a Link . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . (I-15) Slenderness parameter. . . . . . . . . . . . . . . . . . . . . . . (I-13)


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xxii SYMBOLS

ps Limiting slenderness parameter for compact element. . . . . . (Table I-8-1) Resistance Factor . . . . . . . . . . . . . . . . . . . . . . . . . (I-8)c Resistance Factor for compression . . . . . . . . . . . . . . . . (I-13)v Resistance Factor for shear strength of Panel Zone of

beam-to-column connections . . . . . . . . . . . . . . . . . . . (I-9)Resistance Factor for the shear strength of a Composite Column (II-6)

Link Rotation Angle . . . . . . . . . . . . . . . . . . . . . . . . (S2)


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PB262-01 PB262/Seismic.cls January 2, 2003 16:16 Char Count=

1PART I. STRUCTURAL STEEL BUILDINGS

GLOSSARY

The first letter(s) of words or terms that appear in this glossary are generally capitalizedthroughout these Provisions.

Applicable Building Code. The building code under which the building is designed. Inthe absence of an Applicable Building Code, the loads and load combinations shall bethose stipulated in ASCE 7.

Amplified Seismic Load. The horizontal component of earthquake load E multiplied byo, where E and the horizontal component of E are defined in the Applicable BuildingCode.

Authority Having Jurisdiction. The organization, political subdivision, office or individualcharged with the responsibility of administering and enforcing the provisions of thisstandard.

Beam. A structural member that primarily functions to carry loads transverse to its longi-tudinal axis; usually a horizontal member in a seismic frame system.

Braced Frame. A vertical truss system of concentric or eccentric type that resists lateralforces on the Structural System.

Column Base. The assemblage of plates, connectors, bolts, and rods at the base of a columnused to transmit forces between the steel superstructure and the foundation.

Connection. A combination of joints used to transmit forces between two or more mem-bers. Connections are categorized by the type and amount of force transferred (moment,shear, end reaction).

Continuity Plates. Column stiffeners at the top and bottom of the Panel Zone; also knownas transverse stiffeners.

Design Earthquake. The earthquake represented by the Design Response Spectrum asspecified in the Applicable Building Code.

Design Story Drift. The amplified story drift (drift under the Design Earthquake, includingthe effects of inelastic action), determined as specified in the Applicable BuildingCode.

Design Strength. Resistance (force, moment, stress, as appropriate) provided by elementor connection; the product of the Nominal Strength and the Resistance Factor.

Diagonal Bracing. Inclined structural members carrying primarily axial load that areemployed to enable a structural frame to act as a truss to resist lateral loads.

Dual System. A Dual System is a Structural System with the following features: (1) anessentially complete space frame that provides support for gravity loads; (2) resistanceto lateral load provided by moment resisting frames (SMF, IMF or OMF) that arecapable of resisting at least 25 percent of the base shear, and concrete or steel shearwalls, or steel Braced Frames (EBF, SCBF or OCBF); and, (3) each system designedto resist the total lateral load in proportion to its relative rigidity.


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2 PART I – GLOSSARY

Eccentrically Braced Frame (EBF). A diagonally Braced Frame meeting the require-ments in Section 15 that has at least one end of each bracing member connected toa beam a short distance from another beam-to-brace connection or a beam-to-columnconnection.

Expected Yield Strength. The probable yield strength of the material, equal to the minimumspecified yield strength, Fy , multiplied by Ry .

Fully Restrained (FR). Sufficient rigidity exists in the connection to maintain the anglesbetween intersecting members.

Intermediate Moment Frame (IMF). A Moment Frame system that meets the requirementsin Section 10.

Interstory Drift Angle. Interstory displacement divided by story height, radians.

Inverted-V-Braced Frame. See V-Braced Frame.

Joint. An area where two or more ends, surfaces or edges are attached. Joints are catego-rized by the type of fastener or weld used and the method of force transfer.

k-Area. An area of potentially reduced notch-toughness located in the web-to-flange filletarea. See Figure C-I-6.1.

K-Braced Frame. An OCBF in which a pair of diagonal braces located on one side of acolumn is connected to a single point within the clear column height.

Lateral Bracing Member. A member that is designed to inhibit lateral buckling or lateral-torsional buckling of primary framing members.

Link. In EBF, the segment of a beam that is located between the ends of two diagonalbraces or between the end of a diagonal brace and a column. The length of the Link isdefined as the clear distance between the ends of two diagonal braces or between thediagonal brace and the column face.

Link Intermediate Web Stiffeners. Vertical web stiffeners placed within the Link in EBF.

Link Rotation Angle. The inelastic angle between the Link and the beam outside of theLink when the total story drift is equal to the Design Story Drift.

Link Shear Design Strength. The lesser of the design shear strength of the Link developedfrom the moment or shear strength of the Link.

Load and Resistance Factor Design (LRFD). A method of proportioning structural com-ponents (members, connectors, connecting elements, and assemblages) such that noapplicable limit state is exceeded when the building is subjected to all appropriate loadcombinations.

Moment Frame. A building frame system in which seismic shear forces are resisted byshear and flexure in members and connections of the frame.

Nominal Loads. The magnitudes of the loads specified by the Applicable Building Code.

Nominal Strength. The capacity of a building or component to resist the effects of loads,as determined by computations using specified material strengths and dimensions andformulas derived from accepted principles of structural mechanics or by field tests orlaboratory tests of scaled models, allowing for modeling effects and differences betweenlaboratory and field conditions.


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PART I – GLOSSARY 3

Ordinary Concentrically Braced Frame (OCBF). A diagonally Braced Frame meeting therequirements in Section 14 in which all members of the bracing system are subjectedprimarily to axial forces.

Ordinary Moment Frame (OMF). A Moment Frame system that meets the requirementsin Section 11.

P-Delta Effect. Second-order effect of column axial loads after lateral deflection of theframe on the shears and moments in members.

Panel Zone. The web area of the beam-to-column connection delineated by the extensionof beam and column flanges through the connection.

Partially Restrained (PR). A connection with insufficient rigidity to maintain the anglesbetween connected members in original alignment after load is applied.

Prequalified Connections. Connections that comply with the requirements of Appendix P.

Reduced Beam Section. A reduction in cross section over a discrete length that promotesa zone of inelasticity in the member.

Required Strength. The load effect (force, moment, stress, or as appropriate) acting ona member or connection that is determined by structural analysis from the factoredloads using the most appropriate critical load combinations, or as specified in theseProvisions.

Resistance Factor. A factor that accounts for unavoidable deviations in the actual strengthof a member or connection from the Nominal Strength and for the manner and conse-quences of failure.

Seismic Design Category. A classification assigned to a building based upon such factorsas its occupancy and use.

Seismic Load Resisting System. The assembly of structural elements in the building thatresists seismic loads, including struts, collectors, chords, diaphragms and trusses.

Slip-Critical Joint. A bolted joint in which slip resistance on the faying surface(s) of theconnection is required.

Special Concentrically Braced Frame (SCBF). A diagonally Braced Frame meeting therequirements in Section 13 in which all members of the bracing system are subjectedprimarily to axial forces.

Special Moment Frame (SMF). A Moment Frame system that meets the requirements inSection 9.

Special Truss Moment Frame (STMF). A truss Moment Frame system that meets therequirements in Section 12.

Static Yield Strength. The strength of a structural member or connection that is deter-mined on the basis of testing that is conducted under slow monotonic loading untilfailure.

Structural System. An assemblage of load-carrying components that are joined togetherto provide interaction or interdependence.

V-Braced Frame. A concentrically Braced Frame (SCBF or OCBF) in which a pair ofdiagonal braces located either above or below a beam is connected to a single point


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4 PART I – GLOSSARY

within the clear beam span. Where the diagonal braces are below the beam, the systemis also referred to as an Inverted-V-Braced Frame.

X-Braced Frame. A concentrically braced frame (OCBF) in which a pair of diagonalbraces crosses near mid-length of the braces.

Y-Braced Frame. An Eccentrically Braced Frame (EBF) in which the stem of the Y is theLink of the EBF system.

Zipper Column. A vertical (or nearly vertical) strut connecting the brace-to-beam inter-section of an Inverted-V-Braced Frame at one level to the brace-to-beam intersection atanother level. See Figure C-I-13.3(b).


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Sect. 2.] PART I – REFERENCED SPECIFICATIONS, CODES, AND STANDARDS 5

1. SCOPE

These Provisions are intended for the design and construction of structural steelmembers and connections in the Seismic Load Resisting Systems in buildings forwhich the design forces resulting from earthquake motions have been determinedon the basis of various levels of energy dissipation in the inelastic range of response.These Provisions shall apply to buildings that are classified in the ApplicableBuilding Code as Seismic Design Category D (or equivalent) and higher or whenrequired by the Engineer of Record.

These Provisions shall be applied in conjunction with the AISC Load and Resis-tance Factor Design Specification for Structural Steel Buildings, hereinafter re-ferred to as the LRFD Specification. All members and connections in the SeismicLoad Resisting System shall have a Design Strength as required in the LRFD Spec-ification, and shall also meet all of the additional requirements in these Provisions.

Part I includes a Glossary, which is specifically applicable to this Part, and Ap-pendices P, S, and X.

2. REFERENCED SPECIFICATIONS, CODES, AND STANDARDS

The documents referenced in these Provisions shall include those listed in LRFDSpecification Section A6 with the following additions and modifications:

American Concrete Institute (ACI)Building Code Requirements for Structural Concrete, ACI 318-02

American Institute of Steel Construction (AISC)Load and Resistance Factor Design Specification for Structural Steel Buildings,

December 27, 1999Load and Resistance Factor Design Specification for the Design of Steel Hollow

Structural Sections, November 10, 2000Load and Resistance Factor Design Specification for Single-Angle Members,

November 10, 2000

American Society of Civil Engineers (ASCE)Minimum Design Loads for Buildings and Other Structures, ASCE 7-02

American Society for Testing and Materials (ASTM)

Standard Specification for General Requirements for Rolled Structural Steel Bars,Plates, Shapes, and Sheet Piling, ASTM A6/A6M-01

Standard Specification for Carbon Structural Steel, ASTM A36/A36M-00

Pipe, Steel, Black and Hot-Dipped, Zinc-Coated Welded and Seamless, ASTMA53/A53M-01

Standard Specification for Low and Intermediate Tensile Strength Carbon SteelPlates, ASTM A283/A283M-00

Standard Specification for Structural Bolts, Steel, Heat Treated, 120/105 ksi Min-imum Tensile Strength, ASTM A325-01

Standard Specification for High-Strength Bolts for Structural Steel Joints[Metric], ASTM A325M-00


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6 PART I – REFERENCED SPECIFICATIONS, CODES, AND STANDARDS [Sect. 2.

Standard Test Methods and Definitions for Mechanical Testing of Steel Products,ASTM A370-02e1

Standard Specification for Heat-Treated Steel Structural Bolts, 150 ksi MinimumTensile Strength, ASTM A490-00

Standard Specification for High-Strength Steel Bolts, Classes 10.9 and 10.9.3, forStructural Steel Joints [Metric], ASTM A490M-00

Standard Specification for Cold-Formed Welded and Seamless Carbon Steel Struc-tural Tubing in Rounds and Shapes, ASTM A500-01

Standard Specification for Hot-Formed Welded and Seamless Carbon Steel Struc-tural Tubing, ASTM A501-01

Standard Specification for High-Strength Carbon-Manganese Steel of StructuralQuality, ASTM A529/A529M-00

Standard Specification for High-Strength Low-Allow Columbium-Vanadium Struc-tural Steel, ASTM A572/A572M-00a

Standard Specification for High-Strength Low-Allow Structural Steel with 50 ksi[345 MPa] Minimum Yield Point to 4 in. [100 mm] Thick, ASTM A588/A588M-00a

Standard Specification for Hot-Formed Welded and Seamless High-Strength Low-Alloy Structural Tubing, ASTM A618-01

Standard Specification for Sampling Procedure for Impact Testing of StructuralSteel, ASTM A673/A673M-95

Standard Specification for Cold-formed Welded and Seamless High Strength, LowAlloy Structural Tubing with Improved Atmospheric Corrosion Resistance,ASTM A847-99a

Standard Specification for High-Strength Low-Allow Steel Shapes of StructuralQuality, Produced by Quenching and Self-Tempering Process (QST), ASTMA913/A913M-00a

Standard Specification for Steel for Structural Shapes for Use in Building Framing,ASTM A992/A992M-00

Standard Test Methods for Tension Testing of Metallic Materials, ASTME8-01e1

Standard Test Methods for Tension Testing of Metallic Materials, ASTME8M-01e1

Standard Specification for “Twist Off” Type Tension Control Structural Bolt/Nut/Washer Assemblies, Steel, Heat Treated, 120/105 ksi Minimum TensileStrength, ASTM F1852-00

American Welding SocietyFiller Metal Procurement Guidelines, AWS A5.01-93Specification for Carbon Steel Electrodes for Flux Cored Arc Welding, AWS

A5.20-95Specification for Low Alloy Steel Electrodes for Flux Cored Arc Welding, AWS

A5.29-98


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Sect. 6.] PART I – MATERIALS 7

TABLE I-4-1System Overstrength Factor, o

Seismic Load Resisting System Ωo

All moment-frame systems meeting Part I requirements 3

Eccentrically Braced Frames (EBF) meeting Part I requirements 21/2

All other systems meeting Part I requirements 2

Standard Methods for Mechanical Testing of Welds-U.S. Customary Units, AWSB4.0-98

Standard Methods for Mechanical Testing of Welds-Metric Units, AWS B4.0M:2000

Structural Welding Code – Steel, AWS D1.1:2002

Research Council on Structural ConnectionsSpecification for Structural Joints Using ASTM A325 or A490 Bolts, June 23, 2000

3. GENERAL SEISMIC DESIGN REQUIREMENTS

The Required Strength and other seismic provisions for Seismic Design Categories(SDCs), Seismic Use Groups or Seismic Zones and the limitations on height andirregularity shall be as specified in the Applicable Building Code.

4. LOADS, LOAD COMBINATIONS, AND NOMINAL STRENGTHS

4.1. Loads and Load Combinations

The loads and load combinations shall be as stipulated by the Applicable BuildingCode (see Glossary). Where Amplified Seismic Loads are required by these pro-visions, the horizontal earthquake load E (as defined in the Applicable BuildingCode) shall be multiplied by the overstrength factor o prescribed by the Appli-cable Building Code. In the absence of a specific definition of o, the value foro shall be as listed in Table I-4-1.

4.2. Nominal Strength

The Nominal Strength of systems, members and connections shall meet the require-ments in the LRFD Specification, except as modified throughout these Provisions.

5. STORY DRIFT

The Design Story Drift and story drift limits shall be determined as specified inthe Applicable Building Code.

6. MATERIALS

6.1. Material Specifications

Structural steel used in the Seismic Load Resisting System shall meet the require-ments in LRFD Specification Section A3.1a, except as modified in this Section.


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8 PART I – MATERIALS [Sect. 6.

TABLE I-6-1Ry Values for Different Member Types

Application Ry

Hot-rolled structural shapes and barsASTM A36/A36M 1.5ASTM A572/A572M Grade 42 (290) 1.3ASTM A992/A992M 1.1All other grades 1.1

Hollow Structural SectionsASTM A500, A501, A618 and A847 1.3

Steel PipeASTM A53/A53M 1.4

Plates 1.1All other products 1.1

For buildings over one story in height, the steel used in the Seismic Load Resist-ing Systems described in Sections 9, 10, 11, 12, 13, 14 and 15 shall meet one ofthe following ASTM Specifications: A36/A36M, A53/A53M, A500 (Grade B orC), A501, A529/A529M, A572/A572M (Grade 42 (290), 50 (345) or 55 (380)),A588/A588M, A913/A913M (Grade 50 (345) or 65 (450)), or A992/A992M. Thesteel used for Column Base plates shall meet one of the preceding ASTM specifi-cations or ASTM A283/A283M Grade D. The specified minimum yield strengthof steel to be used for members in which inelastic behavior is expected shall notexceed 50 ksi (345 MPa) unless the suitability of the material is determined bytesting or other rational criteria. This limitation does not apply to columns forwhich the only expected inelastic behavior is yielding at the Column Base.

No thermal treatment of weldment or test specimens is permitted, except thatmachined tensile test specimens may be aged at 200˚F (93˚C) to 220˚F (104˚C)for up to 48 hours, then cooled to room temperature before testing.

6.2. Material Properties for Determination of Required Strength

When required in these Provisions, the Required Strength of a connection or mem-ber shall be determined from the Expected Yield Strength Ry Fy , of the connectedmember, where Fy is the specified minimum yield strength of the grade of steelto be used. For rolled shapes and bars, Ry shall be as given in Table I-6-1. Othervalues of Ry are permitted to be used if the value of the Expected Yield Strengthis determined by testing that is conducted in accordance with the requirements forthe specified grade of steel.

When both the Required Strength and the Design Strength calculations are madefor the same member or connecting element, it is permitted to apply Ry to Fy inthe determination of the Design Strength.

6.3. Notch-toughness Requirements

When used as members in the Seismic Load Resisting System, ASTM A6/A6MGroups 3, 4, and 5 shapes with flanges 11/2 in. (38 mm) thick and thicker, and platesthat are 2-in. (50 mm) thick or thicker shall have a minimum Charpy V-Notch


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Sect. 7.] PART I – CONNECTIONS, JOINTS, AND FASTENERS 9

(CVN) toughness of 20 ft-lbf (27 J) at 70˚F (21˚C), determined as specified inLRFD Specification Section A3.1c.

7. CONNECTIONS, JOINTS, AND FASTENERS

7.1. Scope

Connections, joints, and fasteners that are part of the Seismic Load ResistingSystem shall meet the requirements in LRFD Specification Chapter J, except asmodified in this Section.

7.2. Bolted Joints

All bolts shall be pretensioned high-strength bolts. All faying surfaces shall beprepared as required for Class A or better Slip-Critical Joints. The design shearstrength of bolted joints is permitted to be calculated as that for bearing-type joints.

Bolted joints shall not be designed to share load in combination with welds on thesame faying surface.

The bearing strength of bolted joints shall be provided using either standard holesor short-slotted holes with the slot perpendicular to the line of force, unless analternative hole type is justified as part of a tested assembly; see Appendix S.

The Design Strength of bolted joints in shear and/or combined tension and shearshall be determined in accordance with LRFD Specification Sections J3.7 andJ3.10, except that the nominal bearing strength at bolt holes shall not be takengreater than 2.4dtFu .

Bolted connections for members that are a part of the Seismic Load ResistingSystem shall be configured such that a ductile limit-state either in the connectionor in the member controls the design.

7.3. Welded Joints

Welding shall be performed in accordance with a Welding Procedure Specifica-tion (WPS) as required in AWS D1.1 and approved by the Engineer of Record.The WPS variables shall be within the parameters established by the filler metalmanufacturer.

7.3a. General Requirements

All welds used in members and connections in the Seismic Load Resisting Systemshall be made with a filler metal that can produce welds that have a minimumCharpy V-Notch toughness of 20 ft-lbf (27 J) at minus 20˚F (minus 29˚C), asdetermined by AWS classification or manufacturer certification. This requirementfor notch toughness shall also apply in other cases as required in these Provisions.

7.3b. Additional Requirements in Special Moment Framesand Intermediate Moment Frames

For structures in which the steel frame is normally enclosed and maintained ata temperature of 50˚F (10˚C) or higher, the following CJP welds in Special and


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10 PART I – MEMBERS [Sect. 8.

Intermediate Moment Frames shall be made with filler metal capable of providinga minimum Charpy V-Notch toughness of 20 ft-lbf (27 J) at minus 20˚F (minus29˚C) as determined by AWS classification test methods and 40 ft-lbf (54 J) at70˚F (21˚C) as determined by Appendix X or other approved method:

(1) Welds of beam flanges to columns

(2) Groove welds of shear tabs and beam webs to columns

(3) Column splices

For structures with service temperatures lower than 50˚F (10˚C), these qualificationtemperatures shall be reduced accordingly.

7.3c. Discontinuities

For members and connections that are part of the Seismic Load Resisting System,discontinuities located within a plastic hinging zone defined below, created byerrors or by fabrication or erection operations, such as tack welds, erection aids,air-arc gouging, and flame cutting, shall be repaired as required by the Engineerof Record.

7.4. Other Connections

Welded shear studs shall not be placed on beam flanges within the zones of expectedplastic hinging. The length of a plastic hinging zone shall be defined as one-half ofthe depth of the beam on either side of the theoretical hinge point. Decking arc-spotwelds as required to secure decking shall be permitted. Decking attachments thatpenetrate the beam flanges shall not be used in the plastic hinging zone.

Welded, bolted, screwed, or shot-in attachments for perimeter edge angles, exte-rior facades, partitions, duct work, piping, or other construction shall not be placedwithin the expected zone of plastic deformations of members of the Seismic LoadResisting System. Outside the expected zone of plastic deformation area, calcula-tions, based on the expected moment, shall be made to demonstrate the adequacyof the member net section when connectors that penetrate the member are used.

Exception: Welded shear studs and other connections are permitted where theyhave been included in the connection tests used to qualify the connection.

8. MEMBERS

8.1. Scope

Members in the Seismic Load Resisting System shall meet the requirements in theLRFD Specification and those of this Section. For members that are not part of theSeismic Load Resisting System, see Section 8.4c.

8.2. Local Buckling

Where required by these Provisions, members of the Seismic Load Resisting Sys-tem shall meet the p limitation in Table B5.1 in the LRFD Specification and theps limitations of Table I-8-1.


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Sect. 8.] PART I – MEMBERS 11

TABLE I-8-1Limiting Width Thickness Ratios ps for

Compression ElementsLimiting Width-

Thickness Ratios

Description of Element

WidthThickness

Ratio ps

(seismically compact)

Flanges of I-shaped rolled, hybrid orwelded beams [a], [b], [f], [h]

b/t 0.30√

Es/Fy

Flanges of I-shaped rolled, hybrid orwelded columns [a], [c]

b/t 0.30√

Es/Fy

Flanges of channels, angles and I-shapedrolled, hybrid or welded beams andbraces [a], [d], [h]

b/t 0.30√

Es/Fy

Flanges of I-shaped rolled, hybrid orwelded columns [a], [e]

b/t 0.38√

Es/Fy

Flanges of H-pile sections b/t 0.45√

Es/Fy

Un

stif

fen

edE

lem

ents

Flat bars[g] b/t 2.5

Legs of single angle, legs of double anglemembers with separators, or flangesof tees [h]

b/t 0.30√

Es/Fy

Webs of tees [h] d/t 0.30√

Es/Fy

Webs in flexural compression in beamsin SMF, Section 9, unless notedotherwise [a]

h/tw 2.45√

Es/Fy

Other webs in flexural compression [a] h/tw 3.14√

Es/Fy

Webs in combined flexure and axialcompression [a], [b], [c], [d], [e],[f], [h]

h/tw for Pu/b Py ≤ 0.125

3.14√

EsFy

(1 − 1.54 Pu

b Py

)for Pu/b Py > 0.125

1.12√

EsFy

(2.33 − Pu

b Py

)

Sti

ffen

edE

lem

ents

Round HSS in axial and/or flexural com-pression [d], [h]

D/t 0.044 Es/Fy

Rectangular HSS in axial and/or flexuralcompression [d], [h]

b/t or h/tw 0.64√

Es/Fy

Webs of H-Pile sections h/tw 0.94√

Es/Fy

[a] For hybrid beams, use the yield strength of the flange Fyfinstead of Fy.

[b] Required for beams in SMF, Section 9.[c] Required for columns in SMF, Section 9, unless the ratios

from Equation 9-3 are greater than 2.0 where it ispermitted to use p in LRFD Specification Table B5.1.

[d] Required for beams and braces in SCBF, Section 13.

[e] It is permitted to use p in LRFDSpecification Table B5.1 for columns inSTMF, Section 12 and EBF, Section 15.

[f] Required for Link in EBF, Section 15.[g] Diagonal web members within the

special segment of STMF, Section 12.[h] Chord members of STMF, Section 12.


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12 PART I – MEMBERS [Sect. 8.

8.3. Column Strength

When Pu /Pn is greater than 0.4 without consideration of the Amplified SeismicLoad, the following requirements shall be met:

(1) The required axial compressive and tensile strength, considered in the absenceof any applied moment, shall be determined using the load combinations stipu-lated by the Applicable Building Code including the Amplified Seismic Load.

(2) The Required Strengths need not exceed either of the following:

(a) The maximum load transferred to the column considering 1.1Ry timesthe nominal strengths of the connecting beam or brace elements of thebuilding.

(b) The limit as determined from the resistance of the foundation to overturn-ing uplift.

8.4. Column Splices

8.4a. General

The Required Strength of column splices shall equal the Required Strength of thecolumns, including that determined from Section 8.3.

The centerline of column splices made with fillet welds or partial-joint-penetrationgroove welds shall be located 4 ft. (1.2 m) or more away from the beam-to-columnconnections. When the column clear height between beam-to-column connectionsis less than 8 ft. (2.4 m), splices shall be at half the clear height.

Welded column splices that are subject to a calculated net tensile stress determinedusing the load combinations stipulated by the Applicable Building Code includingthe Amplified Seismic Load, shall be made using filler metal with Charpy V-Notch toughness as required in Section 7.3a and shall meet both of the followingrequirements:

(1) The Design Strength of partial-joint-penetration groove welded joints shall beat least equal to 200 percent of the Required Strength.

(2) The Design Strength for each flange shall be at least 0.5 times Ry Fy A f ,where Ry Fy is the Expected Yield Strength of the column material and A f isthe flange area of the smaller column connected.

Beveled transitions are not required when changes in thickness and width of flangesand webs occur in column splices where partial-joint-penetration groove weldedjoints are permitted.

8.4b. Column Web Splices

Column web splices shall be either bolted or welded, or welded to one columnand bolted to the other. In Moment Frames using bolted splices to develop theRequired Strength, plates or channels shall be used on both sides of the columnweb.


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Sect. 9.] PART I – SPECIAL MOMENT FRAMES (SMF) 13

8.4c. Columns Not Part of the Seismic Load Resisting System

In moment frame buildings, splices of columns that are not a part of the SeismicLoad Resisting System shall satisfy the following:

(1) They shall be located 4 ft. (1.2 m) or more away from the beam-to-column con-nections. When the column clear height between beam-to-column connectionsis less than 8 ft. (2.4 m), splices shall be at half the clear height.

(2) The column splices shall have sufficient design shear strength with respect toboth orthogonal axes of the column to resist a shear force equal to Mpc/H ,where Mpc is the nominal plastic flexural strength of the column for the di-rection in question, and H is the story height.

8.5. Column Bases

The connection of the structure frame elements to the Column Base and the connec-tion of the Column Base to the foundations shall be adequate to transmit the forcesfor which the frame elements were required to be designed. Design of concreteelements at the Column Base, including anchor rod embedment and reinforcementsteel, shall be in accordance with ACI 318. The seismic loads to be transferred tothe foundation soil interface shall be as required by the Applicable Building Code.

8.6. H-Piles

8.6a. Design of H-Piles

Design of H-piles shall comply with the provisions of the AISC LRFD Speci-fication regarding design of members subjected to combined loads. The width-thickness ratios of member elements shall meet the ps limitations of Table I-8-1.

8.6b. Batter H-Piles

If batter (sloped) and vertical piles are used in a pile group, the vertical piles shallbe designed to support combined effects of the dead and live loads without theparticipation of batter piles.

8.6c. Tension in H-Piles

Tension in the pile shall be transferred to the pile cap by mechanical means suchas shear keys, rebars or studs welded to the embedded portion of pile. A length ofpile below the bottom of the pile cap equal to at least the overall depth of the pilecross section shall be free of attachments and welds.

9. SPECIAL MOMENT FRAMES (SMF)

9.1. Scope

Special Moment Frames (SMF) are expected to withstand significant inelasticdeformations when subjected to the forces resulting from the motions of the DesignEarthquake. SMF shall meet the requirements in this Section.


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14 PART I – SPECIAL MOMENT FRAMES (SMF) [Sect. 9.

9.2. Beam-to-Column Joints and Connections

9.2a. Requirements

All beam-to-column joints and connections used in the Seismic Load ResistingSystem shall satisfy the following three requirements:

(1) The connection must be capable of sustaining an Interstory Drift Angle of atleast 0.04 radians.

(2) The required flexural strength of the connection, determined at the columnface, must equal at least 80 percent of the nominal plastic moment of theconnected beam at an Interstory Drift Angle of 0.04 radians.

(3) The required shear strength Vu of the connection shall be determined using theload combination 1.2D + 0.5L + 0.2S plus the shear resulting from the appli-cation of a moment of 2[1.1Ry Fy Z/distance between plastic hinge locations].Alternatively, a lesser value of Vu is permitted if justified by analysis.

Connections that accommodate the required Interstory Drift Angle within theconnection elements and provide the required flexural and shear strengths notedabove are permitted, provided it can be demonstrated by analysis that the additionaldrift due to connection deformation can be accommodated by the building. Suchanalysis shall include effects of overall frame stability including second ordereffects.

9.2b. Conformance Demonstration

All beam-to-column joints and connections used in the Seismic Load ResistingSystem shall be demonstrated to satisfy the requirements of Section 9.2a by oneof the following:

(a) Use a connection Prequalified for SMF in accordance with Appendix P.

(b) Provide qualifying cyclic test results in accordance with Appendix S. Resultsof at least two cyclic connection tests shall be provided and are permitted tobe based on one of the following:

(i) Tests reported in research literature or documented tests performed forother projects that are demonstrated to represent project conditions, withinthe limits specified in Appendix S.

(ii) Tests that are conducted specifically for the project and are representa-tive of project member sizes, material strengths, connection configura-tions, and matching connection processes, within the limits specified inAppendix S.

9.3. Panel Zone of Beam-to-Column Connections (beam webparallel to column web)

9.3a. Shear Strength

The required thickness of the panel zone shall be determined in accordance withthe method used in proportioning the panel zone of the tested connection. As aminimum, the required shear strength Ru of the panel zone shall be determined from


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the summation of the moments at the column faces as determined by projectingthe expected moments at the plastic hinge points to the column faces. The designshear strength v Rv of the panel zone shall be determined using v = 1.0.

(a) When Pu ≤ 0.75Py ,

Rv = 0.6Fydctp

[1 + 3bcf t2

c f

dbdctp

](9-1)

where

tp = total thickness of Panel Zone including doubler plate(s), in. (mm)dc = overall column depth, in. (mm)

bcf = width of the column flange, in. (mm)tc f = thickness of the column flange, in. (mm)db = overall beam depth, in. (mm)Fy = specified minimum yield strength of the Panel Zone steel, ksi (MPa)

(b) When Pu > 0.75Py , Rv shall be calculated using LRFD Specification EquationK1-12.

9.3b. Panel Zone Thickness

The individual thicknesses t of column webs and doubler plates, if used, shallconform to the following requirement:

t ≥ (dz + wz)/90 (9-2)

where

t = thickness of column web or doubler plate, in. (mm)dz = Panel Zone depth between Continuity Plates, in. (mm)wz = Panel Zone width between column flanges, in. (mm)

Alternatively, when local buckling of the column web and doubler plate is pre-vented with plug welds between them, the total Panel Zone thickness shall satisfyEquation 9-2.

9.3c. Panel Zone Doubler Plates

Doubler plates shall be welded to the column flanges using either a complete-joint-penetration groove-welded or fillet-welded joint that develops the designshear strength of the full doubler plate thickness. When doubler plates are placedagainst the column web, they shall be welded across the top and bottom edges todevelop the proportion of the total force that is transmitted to the doubler plate.When doubler plates are placed away from the column web, they shall be placedsymmetrically in pairs and welded to Continuity Plates to develop the proportionof the total force that is transmitted to the doubler plate.


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16 PART I – SPECIAL MOMENT FRAMES (SMF) [Sect. 9.

9.4. Beam and Column Limitations

Abrupt changes in beam flange area are not permitted in plastic hinge regions. Thedrilling of flange holes or trimming of beam flange width is permitted if testingdemonstrates that the resulting configuration can develop stable plastic hingesthat meet the requirements in Section 9.2b. Where employed, the Reduced BeamSection shall meet the required strength as specified in Section 9.2a(2).

Beams and columns shall satisfy the width-thickness limitations given in TableI-8-1.

9.5. Continuity Plates

Continuity Plates shall be provided to match the tested connection.

9.6. Column-Beam Moment Ratio

The following relationship shall be satisfied at beam-to-column connections:M∗

pc

M∗pb

> 1.0 (9-3)

where

M*pc = the sum of the moments in the column above and below the jointat the intersection of the beam and column centerlines. M*pcis determined by summing the projections of the nominal flexu-ral strengths of the column (including haunches where used) aboveand below the joint to the beam centerline with a reduction forthe axial force in the column. It is permitted to take M*pc =Zc(Fyc − Puc/Ag). When the centerlines of opposing beams in thesame joint do not coincide, the mid-line between centerlines shallbe used.

M*pb = the sum of the moment(s) in the beam(s) at the intersection of thebeam and column centerlines. M*pb is determined by summingthe projections of the expected beam flexural strength(s) at the plas-tic hinge location(s) to the column centerline. It is permitted totake M*pb = (1.1Ry Fyb Zb + Mv ), where Mv is the additionalmoment due to shear amplification from the location of the plastichinge to the column centerline. Alternatively, it is permitted to de-termine M*pb from test results as required in Section 9.2b orby analysis based upon the tests. When connections with ReducedBeam Sections are used, it is permitted to take M*pb =(1.1Ry Fybzb + Mv ).

Ag = gross area of column, in.2(mm2)Fyc = specified minimum yield strength of column, ksi (MPa)Puc = required column axial compressive strength, kips (a positive

number) (N)Zb = plastic section modulus of the beam, in.3 (mm3)Zc = plastic section modulus of the column, in.3 (mm3)zb = minimum plastic section modulus at the Reduced Beam Section, in.3

(mm3)


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Exception: When columns conform to the requirements in Section 9.4, this re-quirement does not apply in the following two cases:

(a) Columns with Puc < 0.3Fyc Ag for all load combinations other than thosedetermined using the Amplified Seismic Load that meet either of the followingrequirements:

(i) Columns used in a one-story building or the top story of a multistorybuilding.

(ii) Columns where: (1) the sum of the design shear strengths of all exemptedcolumns in the story is less than 20 percent of the required story shearstrength; and (2) the sum of the design shear strengths of all exemptedcolumns on each column line within that story is less than 33 percentof the required story shear strength on that column line. For the purposeof this exception, a column line is defined as a single line of columns orparallel lines of columns located within 10 percent of the plan dimensionperpendicular to the line of columns.

(b) Columns in any story that have a ratio of design shear strength to requiredshear strength that is 50 percent greater than the story above.

9.7. Beam-to-Column Connection Restraint

9.7a. Restrained Connections

Column flanges at beam-to-column connections require lateral bracing only at thelevel of the top flanges of the beams when a column is shown to remain elasticoutside of the Panel Zone. It shall be permitted to assume that the column remainselastic when the ratio calculated using Equation 9-3 is greater than 2.

When a column cannot be shown to remain elastic outside of the Panel Zone, thefollowing requirements shall apply:

(1) The column flanges shall be laterally supported at the levels of both the topand bottom beam flanges.

(2) Each column-flange lateral bracing shall be designed for a Required Strengththat is equal to 2 percent of the nominal beam flange strength (Fyb f tb f ).

(3) Column flanges shall be laterally supported, either directly or indirectly, bymeans of the column web or by the flanges of perpendicular beams.

9.7b. Unrestrained Connections

A column containing a beam-to-column connection with no lateral bracing trans-verse to the seismic frame at the connection shall be designed using the distancebetween adjacent lateral braces as the column height for buckling transverse to theseismic frame and shall conform to LRFD Specification Chapter H, except that:

(1) The required column strength shall be determined from the LRFD Specifica-tion, except that E shall be taken as the lesser of:

(a) The Amplified Seismic Load.

(b) 125 percent of the frame Design Strength based upon either the beamdesign flexural strength or Panel Zone design shear strength.


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18 PART I – INTERMEDIATE MOMENT FRAMES (IMF) [Sect. 10.

(2) The slenderness L/r for the column shall not exceed 60.

(3) The column required flexural strength transverse to the seismic frame shallinclude that moment caused by the application of the beam flange force spec-ified in Section 9.7a(2) in addition to the second-order moment due to theresulting column flange displacement.

9.8. Lateral Bracing of Beams

Both flanges of beams shall be laterally braced directly or indirectly. The un-braced length between lateral braces shall not exceed 0.086ry Es /Fy . The RequiredStrength of lateral bracing shall be at least 2 percent of the beam flange NominalStrength, Fyb f t f .

In addition, lateral braces shall be placed near concentrated forces, changes incross-section and other locations where analysis indicates that a plastic hinge willform during inelastic deformations of the SMF. Where the design is based uponassemblies tested in accordance with Appendix S, the placement of lateral bracingfor the beams shall be consistent with that used in the tests. The Required Strengthof lateral bracing provided adjacent to plastic hinges shall be at least 6 percentof the expected Nominal Strength of the beam flange computed as Ry Fyb f t f .The required stiffness of all lateral bracing shall be determined in accordancewith Equation C3-8 or C3-10, as applicable, of the LRFD Specification. In theseequations, Mu shall be computed as RyZFy .

9.9. Column Splices

Column splices shall comply with the requirements in Sections 8.4 and 7.3b. Inaddition, column splices in Special Moment Frames shall be located as describedin Section 8.4a, and shall have a required flexural strength that is at least equal toRy times the design flexural strength of the smaller column. Where groove weldsare used to make the splice, they shall be complete-joint-penetration groove welds.Weld tabs shall be removed. Steel backing need not be removed unless required bythe Engineer of Record. The required shear strength of column web splices shallbe at least equal to 2Mpc/H .

Exception: The Required Strength of the column splice considering appropriatestress concentration factors or fracture mechanics stress intensity factors need notexceed that determined by inelastic analyses.

10. INTERMEDIATE MOMENT FRAMES (IMF)

10.1. Scope

Intermediate Moment Frames (IMF) are expected to withstand limited inelasticdeformations in their members and connections when subjected to the forces result-ing from the motions of the Design Earthquake. IMF shall meet the requirementsin this Section.


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Sect. 10.] PART I – INTERMEDIATE MOMENT FRAMES (IMF) 19


10.2a. Requirements

All beam-to-column joints and connections used in the Seismic Load ResistingSystem shall satisfy the following three requirements:

(1) The connection must be capable of sustaining an Interstory Drift Angle of atleast 0.02 radians.

(2) The flexural strength of the connection, determined at the column face, mustequal at least 80 percent of the nominal plastic moment of the connected beamat an Interstory Drift Angle of 0.02 radians.

(3) The required shear strength Vu of the connection shall be determined usingthe load combination 1.2D + 0.5L + 0.2S plus the shear resulting from theapplication of 2[1.1Ry Fy Z / distance between plastic hinge segments]. Alter-natively, a lesser value of Vu is permitted if justified by analysis. The requiredshear strength need not exceed the shear resulting from the application of LoadCombinations using the Amplified Seismic Load.

Connections that accommodate the required Interstory Drift Angle within theconnection elements and provide the required flexural and shear strengths notedabove are permitted, provided it can be demonstrated by analysis that the additionaldrift due to connection deformation can be accommodated by the building. Suchanalysis shall include effects of overall frame stability including second ordereffects.

10.2b. Conformance Demonstration

All beam-to-column joints and connections used in the Seismic Load ResistingSystem shall be demonstrated to satisfy the requirements of Section 10.2a by oneof the following:

(a) Use a connection prequalified for IMF in accordance with Appendix P.

(b) Provide qualifying cyclic test results in accordance with Appendix S. Resultsof at least two non-identical cyclic connection tests shall be provided and arepermitted to be based on one of the following:


(ii) Tests that are conducted specifically for the project and are representativeof project member sizes, material strengths, connection configurations,and matching connection processes, within the limits specified in Ap-pendix S.


No additional requirements beyond the AISC LRFD Specification.


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20 PART I – ORDINARY MOMENT FRAMES (OMF) [Sect. 11.




Continuity Plates shall be provided to be consistent with the tested connection.







10.9. Column Splices

Column splices shall comply with the requirements in Sections 8.4 and 7.3b.

11. ORDINARY MOMENT FRAMES (OMF)

11.1. Scope

Ordinary Moment Frames (OMF) are expected to withstand minimal inelastic de-formations in their members and connections when subjected to the forces resultingfrom the motions of the Design Earthquake. OMF shall meet the requirements inthis Section.


Beam-to-column connections shall be made with welds and/or high-strength bolts.Connections are permitted to be FR or PR moment connections as follows:

(1) FR moment connections that are part of the Seismic Load Resisting Systemshall be designed for a required flexural strength Mu that is at least equal to1.1Ry Mp of the beam or girder or the maximum moment that can be deliveredby the system, whichever is less.

(a) Wheresteelbackingisusedinconnectionswithcomplete-joint-penetration(CJP) flange welds, steel backing and tabs shall be removed except thattop-flange backing attached to the column by a continuous fillet weld onthe edge below the CJP groove weld need not be removed. Removal ofsteel backing and tabs shall be as follows:

(i) Following the removal of backing, the root pass shall be backgougedto sound weld metal and backwelded with a reinforcing fillet. Thereinforcing fillet shall have a minimum leg size of 5/16-in. (8 mm).


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Sect. 11.] PART I – ORDINARY MOMENT FRAMES (OMF) 21

(ii) Weld tab removal shall extend to within 1/8 in. (3 mm) of the basemetal surface except at Continuity Plates where removal to within1/4 in. (6 mm) of the plate edge is acceptable. Edges of the weldtab shall be finished to a surface roughness value of 500 micro-in.(13 micrometers) or better. Grinding to a flush condition is not re-quired. Gouges and notches are not permitted. The transitional slopeof any area where gouges and notches have been removed shall notexceed 1:5. Material removed by grinding that extends more than1/16 in. (2 mm) below the surface of the base metal shall be filledwith weld metal. The contour of the weld at the ends shall provide asmooth transition, free of notches and sharp corners.

(b) Where weld access holes are provided, they shall be as shown inFigure 11-1. The weld access hole shall be ground smooth to a surfaceroughness value not to exceed 500 micro in. (13 micrometers), and shallbe free of notches and gouges. Notches and gouges shall be repaired asrequired by the Engineer of Record.

(c) Double-sided partial-joint-penetration groove welds and double-sided fil-let welds that resist tensile forces in connections shall be designed to resista required force of 1.1Ry Fy Ag of the connected element or part. Single-sided partial-joint-penetration groove welds and single-sided fillet weldsshall not be used to resist tensile forces in the connections.

(2) PR moment connections are permitted when the following requirements aremet:

(a) Such connections shall provide for the Design Strength as specified inSection 11.2a(1) above.

(b) The nominal flexural strength of the connection, Mn , shall be no less than50 percent of Mp of the connected beam or column, whichever is less.

(c) The stiffness and strength of the PR moment connections shall be consid-ered in the design, including the effect on overall frame stability.

For FR moment connections, the required shear strength Vu of a beam-to-columnconnection shall be determined using the load combination 1.2D + 0.5L + 0.2Splus the shear resulting from the application of a moment of 2[1.1Ry Fy Z / distancebetween plastic hinge segments]. Alternatively, a lesser value of Vu is permitted ifjustified by analysis. For PR moment connections, Vu shall be determined from theload combination above plus the shear resulting from the maximum end momentthat the PR moment connections are capable of resisting.






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22 PART I – ORDINARY MOMENT FRAMES (OMF) [Sect. 11.

Notes: 1. Bevel as required by AWS D1.1 for selected groove weld procedure.2. Larger of tbf or 1/2 in. (13 mm) (plus 1/2 tbf, or minus 1/4 tbf)3. 3/4 tbf to tbf, 3/4 in. (19 mm) minimum (± 1/4 in.) (± 6 mm)4. 3/8 in. (10 mm) minimum radius (plus not limited, minus 0)5. 3 tbf (± 1/2 in.) (±13 mm)

Tolerances shall not accumulate to the extent that the angle of the access hole cutto the flange surface exceeds 25.

Fig. 11-1. Weld access hole detail (from FEMA 350, “Recommended Seismic Design Criteria forNew Steel Moment-Frame Buildings”).


When FR moment connections are made by means of welds of beam flangesor beam-flange connection plates directly to column flanges, Continuity Platesshall be provided to transmit beam flange forces to the column web or webs.Plates shall have a thickness greater than or equal to that of the beam flange orbeam-flange connection plate. The welded joints of the Continuity Plates to thecolumn flanges shall be made with either complete-joint-penetration groove welds,two-sided partial-joint-penetration groove welds combined with reinforcing filletwelds, or two-sided fillet welds. The Required Strength of these joints shall not


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Sect. 12.] PART I – SPECIAL TRUSS MOMENT FRAMES (STMF) 23

be less than the Design Strength of the contact area of the plate with the columnflange. The Required Strength of the welded joints of the Continuity Plates to thecolumn web shall be the least of the following:

(a) The sum of the Design Strengths at the connections of the continuity plate tothe column flanges.

(b) The design shear strength of the contact area of the plate with the column web.

(c) The weld Design Strength that develops the design shear strength of the columnPanel Zone.

(d) The actual force transmitted by the stiffener.







11.9. Column Splices

Column splices shall comply with the requirements in Section 8.4.

12. SPECIAL TRUSS MOMENT FRAMES (STMF)

12.1. Scope

Special Truss Moment Frames (STMF) are expected to withstand significant in-elastic deformation within a specially designed segment of the truss when subjectedto the forces from the motions of the Design Earthquake. STMF shall be limitedto span lengths between columns not to exceed 65 ft (20 m) and overall depthnot to exceed 6 ft (1.8 m). The columns and truss segments outside of the specialsegments shall be designed to remain elastic under the forces that can be generatedby the fully yielded and strain-hardened special segment. STMF shall meet therequirements in this Section.

12.2. Special Segment

Each horizontal truss that is part of the Seismic Load Resisting System shall havea special segment that is located between the quarter points of the span of the truss.The length of the special segment shall be between 0.1 and 0.5 times the truss spanlength. The length-to-depth ratio of any panel in the special segment shall neitherexceed 1.5 nor be less than 0.67.

Panels within a special segment shall either be all Vierendeel panels or allX-braced panels; neither a combination thereof nor the use of other truss diagonal


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24 PART I – SPECIAL TRUSS MOMENT FRAMES (STMF) [Sect. 12.

configurations is permitted. Where diagonal members are used in the special seg-ment, they shall be arranged in an X pattern separated by vertical members. Suchdiagonal members shall be interconnected at points where they cross. The inter-connection shall have a Design Strength adequate to resist a force that is at leastequal to 0.25 times the nominal tensile strength of the diagonal member. Boltedconnections shall not be used for web members within the special segment.

Splicing of chord members is not permitted within the special segment, nor withinone-half the panel length from the ends of the special segment. Axial forces due tofactored dead plus live loads in diagonal web members within the special segmentshall not exceed 0.03Fy Ag .

12.3. Nominal Strength of Special Segment Members

In the fully yielded state, the special segment shall develop the required verticalshear strength through the design flexural strength of the chord members and thedesign axial tensile and compressive strengths of the diagonal web members, whenprovided. The top and bottom chord members in the special segment shall be madeof identical sections and shall provide at least 25 percent of the required verticalshear strength in the fully yielded state. The required axial strength in the chordmembers shall not exceed 0.45 times Fy Ag , where = 0.9. Diagonal membersin any panel of the special segment shall be made of identical sections. The endconnection of diagonal web members in the special segment shall have a DesignStrength that is at least equal to the expected nominal axial tensile strength of theweb member, Ry Fy Ag .

12.4. Nominal Strength of Non-special Segment Members

Members and connections of STMF, except those in the special segment defined inSection 12.2, shall have a Design Strength to resist the effects of load combinationsstipulated by the Applicable Building Code, replacing the earthquake load termE with the lateral loads necessary to develop the expected vertical nominal shearstrength in the special segment Vne given as:

Vne = 3.75Ry Mnc

Ls+ 0.075Es I

(L − Ls)

L3s

+ Ry (Pnt + 0.3Pnc) sin (12-1)

where

Ry = yield stress modification factor, see Section 6.2Mnc = nominal flexural strength of the chord member of the special segment,

kip-in. (N-mm)Es I = flexural elastic stiffness of the chord members of the special segment,

kip-in.2 (N-mm2)L = span length of the truss, in. (mm)

Ls = length of the special segment, in. (mm)Pnt = nominal axial tension strength of diagonal members of the special

segment, kips (N)


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Sect. 13.] PART I – SPECIAL CONCENTRICALLY BRACED FRAMES (SCBF) 25

Pnc = nominal axial compression strength of diagonal members of the specialsegment, kips (N)

= angle of diagonal members with the horizontal

12.5. Compactness

The width-thickness ratio of chord members shall not exceed the limiting psvaluesfrom Table I-8-1. Diagonal web members within the special segment shall be madeof flat bars.

12.6. Lateral Bracing

The top and bottom chords of the trusses shall be laterally braced at the ends ofspecial segment, and at intervals not to exceed L p according to LRFD Specifi-cation Section F1, along the entire length of the truss. The Required Strength ofeach lateral brace at the ends of and within the special segment shall be at least5 percent of the nominal axial compressive strength Pnc of the special segmentchord member. Lateral braces outside of the special segment shall have a RequiredStrength at least 2.5 percent of the nominal compressive strength Pnc of the largestadjoining chord member.

13. SPECIAL CONCENTRICALLY BRACED FRAMES (SCBF)

13.1. Scope

Special Concentrically Braced Frames (SCBF) are expected to withstand signif-icant inelastic deformations when subjected to the forces resulting from the mo-tions of the Design Earthquake. SCBF have increased ductility over OCBF (seeSection 14) due to lesser strength degradation when compression braces buckle.SCBF shall meet the requirements in this Section.

13.2. Bracing Members

13.2a. Slenderness

Bracing members shall have Kl/r ≤ 5.87√

Es/Fy .

13.2b. Required Compressive Strength

The Required Strength of a bracing member in axial compression shall not exceedc Pn .

13.2c. Lateral Force Distribution

Along any line of bracing, braces shall be deployed in alternate directions such that,for either direction of force parallel to the bracing, at least 30 percent but no morethan 70 percent of the total horizontal force is resisted by tension braces, unlessthe Nominal Strength Pn of each brace in compression is larger than the RequiredStrength Pu resulting from the application of load combinations stipulated by theApplicable Building Code including the Amplified Seismic Load. For the purposesof this provision, a line of bracing is defined as a single line or parallel lines whose


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26 PART I – SPECIAL CONCENTRICALLY BRACED FRAMES (SCBF) [Sect. 13.

plan offset is 10 percent or less of the building dimension perpendicular to the lineof bracing.

13.2d. Width-thickness Ratios

Width-thickness ratios of stiffened and unstiffened compression elements of bracesshall meet the compactness requirements in LRFD Specification Table B5.1 (i.e., < p) and the following requirements:

(1) The width-thickness ratio of angle legs shall comply with ps in Table I-8-1.

(2) I-shaped members and channels shall comply with ps in Table I-8-1.

(3) Round HSS shall have an outside diameter to wall thickness ratio conformingto Table I-8-1 unless the round HSS wall is stiffened.

(4) Rectangular HSS shall have a flat width to wall thickness ratio conforming toTable I-8-1 unless the rectangular HSS walls are stiffened.

13.2e. Built-up Members

The spacing of stitches shall be such that the slenderness ratio l/r of individualelements between the stitches does not exceed 0.4 times the governing slendernessratio of the built-up member.

The total design shear strength of the stitches shall be at least equal to the designtensile strength of each element. The spacing of stitches shall be uniform and notless than two stitches shall be used. Bolted stitches shall not be located within themiddle one-fourth of the clear brace length.

Exception: Where it can be shown that braces will buckle without causing shearin the stitches, the spacing of the stitches shall be such that the slenderness ratiol/r of the individual elements between the stitches does not exceed 0.75 times thegoverning slenderness ratio of the built-up member.

13.3. Bracing Connections

13.3a. Required Strength

The Required Strength of bracing connections (including beam-to-column con-nections if part of the bracing system) shall be the lesser of the following:

(a) The nominal axial tensile strength of the bracing member, determined asRy Fy Ag .

(b) The maximum force, indicated by analysis that can be transferred to the braceby the system.

13.3b. Tensile Strength

The design tensile strength of bracing members and their connections, based uponthe limit states of tension rupture on the effective net section and block shearrupture strength, as specified in LRFD Specification Section J4, shall be at leastequal to the Required Strength of the brace as determined in Section 13.3a.


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Sect. 13.] PART I – SPECIAL CONCENTRICALLY BRACED FRAMES (SCBF) 27

13.3c. Flexural Strength

In the direction that the brace will buckle, the required flexural strength of theconnection shall be equal to 1.1Ry Mp of the brace about the critical buckling axis.

Exception: Brace connections that meet the requirements in Section 13.3b, canaccommodate the inelastic rotations associated with brace post-buckling deforma-tions, and have a Design Strength that is at least equal to the nominal compressivestrength Fcr Ag of the brace are permitted.

13.3d. Gusset Plates

The design of gusset plates shall include consideration of buckling.

13.4. Special Bracing Configuration Requirements

13.4a. V-Type and Inverted-V-Type Bracing

V-type and inverted-V-type Braced Frames shall meet the following requirements:

(1) A beam that is intersected by braces shall be continuous between columns.

(2) A beam that is intersected by braces shall be designed to support the effectsof all tributary dead and live loads from load combinations stipulated by theApplicable Building Code, assuming that bracing is not present.

(3) A beam that is intersected by braces shall be designed to resist the effects ofload combinations stipulated by the Applicable Building Code, except that aload Qb shall be substituted for the term E . Qb is the maximum unbalancedvertical load effect applied to the beam by the braces. This load effect shall becalculated using a minimum of Ry Py for the brace in tension and a maximumof 0.3 times c Pn for the brace in compression.

(4) The top and bottom flanges of the beam at the point of intersection of bracesshall be designed to support a lateral force that is equal to 2 percent of thenominal beam flange strength Fyb f tb f .

Exception: Limitations 2 and 3 need not apply to penthouses, one-story buildings,nor the top story of buildings.

13.4b. K-Type Bracing

K-type Braced Frames are not permitted for SCBF.

13.5. Columns

Columns in SCBF shall meet the following requirements:

Width-thickness ratios of stiffened and unstiffened compression elements ofcolumns shall meet the requirements for bracing members in Section 13.2d.

In addition to meeting the requirements in Section 8.4, column splices in SCBFshall be designed to develop at least the nominal shear strength of the smaller


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28 PART I – ECCENTRICALLY BRACED FRAMES (EBF) [Sect. 15.

connected member and 50 percent of the nominal flexural strength of the smallerconnected section. Splices shall be located in the middle one-third of the columnclear height.

14. ORDINARY CONCENTRICALLY BRACED FRAMES (OCBF)

14.1. Scope

Ordinary Concentrically Braced Frames (OCBF) are expected to withstand limitedinelastic deformations in their members and connections when subjected to theforces resulting from the motions of the Design Earthquake. OCBF shall meet therequirements in this Section.

14.2. Strength

The Required Strength of the members and connections, other than brace connec-tions, in OCBFs shall be determined using the load combinations stipulated by theApplicable Building Code, including the Amplified Seismic Load. The RequiredStrength of brace connections is the expected tensile strength of the brace, deter-mined as Ry Fy Ag . Braces with Kl/r greater than 4.23

√Es/Fy shall not be used in

V or inverted-V configurations.

15. ECCENTRICALLY BRACED FRAMES (EBF)

15.1. Scope

Eccentrically Braced Frames (EBFs) are expected to withstand significant inelasticdeformations in the Links when subjected to the forces resulting from the motionsof the Design Earthquake. The diagonal braces, the columns, and the beam seg-ments outside of the Links shall be designed to remain essentially elastic under themaximum forces that can be generated by the fully yielded and strain-hardenedLinks, except where permitted in this Section. In buildings exceeding five storiesin height, the upper story of an EBF system is permitted to be designed as an OCBFor an SCBF and still be considered to be part of an EBF system for the purposesof determining system factors in the Applicable Building Code. EBF shall meetthe requirements in this Section.

15.2. Links

Links shall comply with the width-thickness ratios in Table I-8-1.

The specified minimum yield stress of steel used for Links shall not exceed 50 ksi(345 MPa).

The web of a Link shall be single thickness without doubler-plate reinforcementand without web penetrations.

Except as limited below, the required shear strength of the Link Vu shall not exceedthe design shear strength of the Link Vn ,


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Sect. 15.] PART I – ECCENTRICALLY BRACED FRAMES (EBF) 29

where:

= 0.9Vn = Nominal shear strength of the Link, equal to the lesser of Vp or 2Mp/e,

kips (N)Vp = 0.6Fy Aw , kips (N)

e = Link length, in. (mm)Aw = (db-2t f )tw

If the required axial strength Pu in a Link is equal to or less than 0.15Py , where Pyis equal to Fy Ag , the effect of axial force on the Link design shear strength neednot be considered.

If the required axial strength Pu in a Link exceeds 0.15Py , the following additionalrequirements shall be met:

(1) The Link design shear strength shall be the lesser of Vpa or 2Mpa /e, where:

= 0.9

Vpa = Vp

√1 − (

Pu/Py)2

(15-1)

Mpa = 1.18Mp[1 − (

Pu/Py)]

(15-2)

(2) The length of the Link shall not exceed:

[1.15 − 0.5 ′(Aw/Ag)]1.6Mp/Vp when ′(Aw/Ag) ≥ 0.3, (15-3)

nor

1.6Mp/Vp when ′(Aw/Ag) < 0.3, (15-4)

where:

Aw = (db − 2t f )tw ′ = Pu /Vu

The Link Rotation Angle is the inelastic angle between the Link and the beamoutside of the Link when the total story drift is equal to the Design Story Drift, .The Link Rotation Angle shall not exceed the following values:

(a) 0.08 radians for Links of length 1.6Mp/Vp or less.

(b) 0.02 radians for Links of length 2.6Mp/Vp or greater.

(c) The value determined by linear interpolation between the above values forLinks of length between 1.6Mp/Vp and 2.6Mp/Vp.

15.3. Link Stiffeners

Full-depth web stiffeners shall be provided on both sides of the Link web at thediagonal brace ends of the Link. These stiffeners shall have a combined widthnot less than (bf − 2tw ) and a thickness not less than 0.75tw nor 3/8 in. (10 mm),whichever is larger, where bf and tw are the Link flange width and Link webthickness, respectively.


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30 PART I – ECCENTRICALLY BRACED FRAMES (EBF) [Sect. 15.

Links shall be provided with intermediate web stiffeners as follows:

(a) Links of lengths 1.6Mp/Vp or less shall be provided with intermediate webstiffeners spaced at intervals not exceeding (30tw − d/5) for a Link RotationAngle of 0.08 radians or (52tw − d/5) for Link Rotation Angles of 0.02 ra-dians or less. Linear interpolation shall be used for values between 0.08 and0.02 radians.

(b) Links of length greater than 2.6Mp/Vp and less than 5Mp/Vp shall be providedwith intermediate web stiffeners placed at a distance of 1.5 times bf from eachend of the Link.

(c) Links of length between 1.6Mp/Vp and 2.6Mp/Vp shall be provided withintermediate web stiffeners meeting the requirements of 1 and 2 above.

(d) Intermediate web stiffeners are not required in Links of lengths greater than5Mp/Vp.

(e) Intermediate Link web stiffeners shall be full depth. For Links that are less than25 in. (635 mm) in depth, stiffeners are required on only one side of the Linkweb. The thickness of one-sided stiffeners shall not be less than tw or 3/8 in.(10 mm), whichever is larger, and the width shall be not less than (bf /2)-tw .For Links that are 25 in. (635 mm) in depth or greater, similar intermediatestiffeners are required on both sides of the web.

The Required Strength of fillet welds connecting a Link stiffener to the Link webis Ast Fy , where Ast is the area of the stiffener. The Required Strength of filletwelds fastening the stiffener to the flanges is Ast Fy /4.

15.4. Link-to-Column Connections

Link-to-column connections must be capable of sustaining the maximum LinkRotation Angle based on the length of the Link, as specified in Section 15.2. Thestrength of the connection, measured at the column face, must equal at least thenominal shear strength of the Link, Vn , as specified in Section 15.2 at the maximumLink Rotation Angle.

Link-to-column connections shall be demonstrated to satisfy the above require-ments by one of the following:

(a) Use a connection Prequalified for EBF in accordance with Appendix P.

(b) Provide qualifying cyclic test results in accordance with Appendix S. Resultsof at least two cyclic connection tests shall be provided and are permitted tobe based on one of the following:


(ii) Tests that are conducted specifically for the project and are representa-tive of project member sizes, material strengths, connection configura-tions, and matching connection processes, within the limits specified inAppendix S.


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Sect. 15.] PART I – ECCENTRICALLY BRACED FRAMES (EBF) 31

Exception: Where reinforcement at the beam-to-column connection at the Link endprecludes yielding of the beam over the reinforced length, the Link is permitted tobe the beam segment from the end of the reinforcement to the brace connection.Where such Links are used and the Link length does not exceed 1.6Mp/Vp, cyclictesting of the reinforced connection is not required if the Design Strength of the re-inforced section and the connection equals or exceeds the Required Strength calcu-lated based upon the strain-hardened Link as described in Section 15.6. Full depthstiffeners as required in Section 15.3 shall be placed at the Link-to-reinforcementinterface.

15.5. Lateral Bracing of Link

Lateral bracing shall be provided at both the top and bottom Link flanges at theends of the Link. The Required Strength of end lateral bracing of Links is 6 percentof the expected Nominal Strength of the Link flange computed as Ry Fyb f t f .

15.6. Diagonal Brace and Beam Outside of Link

The required combined axial and flexural strength of the diagonal brace shall bethe axial forces and moments generated by the expected nominal shear strength ofthe Link Ry Vn increased by 125 percent to account for strain-hardening, whereVn is as defined in Section 15.2. The Design Strengths of the diagonal brace,as determined in LRFD Specification Chapter H (including Appendix H3), shallexceed the Required Strengths as defined above.

The design of the beam outside the Link shall meet the following requirements:

(1) The Required Strength of the beam outside of the Link shall be the forcesgenerated by at least 1.1 times the expected nominal shear strength of the LinkRy Vn , where Vn is as defined in Section 15.2. For determining the DesignStrength of this portion of the beam, it is permitted to multiply the DesignStrengths determined from the LRFD Specification by Ry .

(2) The beam shall be provided with lateral bracing where analysis indicates thatsupport is necessary to maintain the stability of the beam. Lateral bracing shallbe provided at both the top and bottom flanges of the beam and each shall havea Required Strength of at least 2 percent of the beam flange Nominal Strengthcomputed as Fyb f t f .

At the connection between the diagonal brace and the beam at the Link end of thebrace, the intersection of the brace and beam centerlines shall be at the end of theLink or in the Link.

The Required Strength of the diagonal brace-to-beam connection at the Link endof the brace shall be at least the expected Nominal Strength of the brace as givenin Section 15.6. No part of this connection shall extend over the Link length. If thebrace resists a portion of the Link end moment, the connection shall be designedas an FR moment connection.

The width-thickness ratio of the brace shall satisfy p in LRFD SpecificationTable B5.1.


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32 PART I – QUALITY ASSURANCE [Sect. 16.

15.7. Beam-to-Column Connections

Beam-to-column connections away from Links are permitted to be designed aspinned in the plane of the web. The connection shall have a Required Strength toresist rotation about the longitudinal axis of the beam based upon two equal andopposite forces of at least 2 percent of the beam flange Nominal Strength computedas Fyb f t f acting laterally on the beam flanges.

15.8. Required Column Strength

In addition to the requirements in Section 8, the Required Strength of columns shallbe determined from load combinations as stipulated by the Applicable BuildingCode, except that the moments and axial loads introduced into the column at theconnection of a Link or brace shall not be less than those generated by the expectedNominal Strength of the Link multiplied by 1.1 to account for strain-hardening.The expected Nominal Strength of the Link is Ry Vn , where Vn is as defined inSection 15.2.

16. QUALITY ASSURANCE

The general requirements and responsibilities for performance of a quality assur-ance plan shall be in accordance with the requirements of the Authority HavingJurisdiction and the specifications of the Engineer of Record.

The special inspections and tests necessary to establish that the construction is inconformance with these Provisions shall be included in a quality assurance plan.The contractor’s quality control program and qualifications, such as participation ina recognized quality certification program, shall be considered when establishinga quality control plan.

The minimum special inspection and testing contained in the quality assuranceplan beyond that required in LRFD Specification Section M5 shall be as follows:

(1) Visual inspection of welding shall be the primary method used to confirm thatthe procedures, materials and workmanship incorporated in construction arethose that have been specified and approved for the project. Visual inspectionsshall be conducted by qualified personnel, in accordance with a written prac-tice. Nondestructive testing of welds in conformance with AWS D1.1 shallalso be performed, but shall not serve to replace visual inspection.

(2) All complete-joint-penetration and partial-joint-penetration groove weldedjoints that are subjected to net tensile forces as part of the Seismic LoadResisting Systems in Sections 9, 10, 11, 12, 13, 14 and 15 shall be tested usingapproved nondestructive methods conforming to AWS D1.1.

Exception: The amount of nondestructive testing is permitted to be reduced ifapproved by the Engineer of Record and the Authority Having Jurisdiction.

When welds from web doubler plates or Continuity Plates occur in the k-Areaof rolled steel columns, the k-Area adjacent to the welds shall be inspected afterfabrication, as required by the Engineer of Record, using approved nondestructivemethods conforming to AWS D1.1.


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33

APPENDIX P

PREQUALIFICATION OF BEAM-TO-COLUMNAND LINK-TO-COLUMN CONNECTIONS

PART I

P1. SCOPE

This appendix contains minimum requirements for prequalification of beam-to-column moment connections in Special Moment Frames (SMFs) and IntermediateMoment Frames (IMFs), and link-to-column connections in Eccentrically BracedFrames (EBFs). Prequalified Connections are permitted to be used, within theapplicable limits of prequalification, without the need for further qualifying cyclictests.

P2. GENERAL REQUIREMENTS

P2.1. Basis for Prequalification

Connections shall be Prequalified based on test data satisfying Section P3, sup-ported by analytical studies and design models. The combined body of evidencefor prequalification must be sufficient to assure that the connection can supply therequired Interstory Drift Angle for SMF and IMF systems, or the required LinkRotation Angle for EBFs, on a consistent and reliable basis within the specifiedlimits of prequalification. All applicable limit states for the connection that affectthe stiffness, strength and deformation capacity of the connection and the SeismicLoad Resisting System must be identified. These include fracture related limitstates, stability related limit states, and all other limit states pertinent for the con-nection under consideration. The effect of design variables listed in Section P4shall be addressed for connection prequalification.

P2.2. Authority for Prequalification

Prequalification of a connection and the associated limits of prequalification shallbe established by a Connection Prequalification Review Panel (CPRP) approvedby the Authority Having Jurisdiction.

P3. TESTING REQUIREMENTS

Data used to support connection prequalification shall be based on tests conductedin accordance with Appendix S. The CPRP shall determine the number of tests andthe variables considered by the tests for connection prequalification. The CPRPshall also provide the same information when limits are to be changed for a pre-viously prequalified connection. A sufficient number of tests shall be performed


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34 PART I – PREQUALIFICATION VARIABLES [App. P4.

on enough non-identical specimens to demonstrate that the connection has theability and reliability to undergo the required Interstory Drift Angle for SMFs andIMFs and the required Link Rotation Angle for EBFs, where the Link is adjacentto columns. For connections that are already Prequalified Connections, and thelimits of prequalification are being changed, additional non-identical specimensshall be tested prior to changing prequalification limits. The limits on membersizes for prequalification shall not exceed the limits specified in Appendix S,Section S5.2.

P4. PREQUALIFICATION VARIABLES

In order to be Prequalified, the effect of the following variables on connectionperformance shall be considered. Limits on the permissible values for each variableshall be established by the CPRP for the Prequalified Connection.

(1) Beam or Link parameters:

(a) Cross-section shape: wide flange, box, or other.

(b) Cross-section fabrication method: rolled shape, welded shape, or other.

(c) Depth.

(d) Weight per foot.

(e) Flange thickness.

(f) Material specification.

(g) Span-to-depth ratio (for SMF or IMF), or Link length (for EBF).

(h) Width thickness ratio of cross-section elements.

(i) Lateral bracing.

(j) Other parameters pertinent to the specific connection under consideration.

(2) Column parameters:

(a) Cross-section shape: wide flange, box, or other.

(b) Cross-section fabrication method: rolled shape, welded shape, or other.

(c) Column orientation with respect to beam or Link: beam or Link is con-nected to column flange, beam or Link is connected to column web, beamsor Links are connected to both the column flange and web, or other.

(d) Depth.

(e) Weight per foot.

(f) Flange thickness.

(g) Material specification.

(h) Width-thickness ratio of cross-section elements.

(i) Lateral bracing.

(j) Other parameters pertinent to the specific connection under consideration.


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App. P6.] PART I – PREQUALIFICATION RECORD 35

(3) Beam (or Link) – Column Relations:

(a) Panel zone strength.

(b) Doubler plate attachment details.

(c) Column-beam (or Link) moment ratio.

(4) Continuity Plates:

(a) Identification of conditions under which Continuity Plates are required.

(b) Thickness, width and depth.

(c) Attachment details.

(5) Welds:

(a) Weld type: CJP, PJP, fillet, or plug.

(b) Filler metal strength and toughness.

(c) Details and treatment of weld backing and weld tabs.

(d) Weld access holes: size, geometry and finish.

(e) Welding quality control and quality assurance.

(f) Other parameters pertinent to the specific connection under consideration.

(6) Bolts:

(a) Bolt diameter.

(b) Bolt Grade: ASTM A325, A490, or other.

(c) Installation requirements: pretensioned, snug tight, or other.

(d) Hole type: standard, oversize, short-slot, long-slot, or other.

(e) Hole fabrication method: drilling, punching, sub-punching and reaming,or other.

(f) Other parameters pertinent to the specific connection under consideration.

(7) Additional Connection Details: All variables pertinent to the specific connec-tion under consideration, as established by the CPRP.

P5. DESIGN PROCEDURE

A comprehensive design procedure must be available for a Prequalified Connec-tion. The design procedure must address all applicable limit states within the limitsof prequalification.

P6. PREQUALIFICATION RECORD

A Prequalified Connection shall be provided with a written prequalification recordwith the following information:

(1) General description of the Prequalified Connection and drawings that clearlyidentify key features and components of the connection.


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36 PART I – PREQUALIFICATION RECORD [App. P6.

(2) Description of the expected behavior of the connection in the elastic andinelastic ranges of behavior, intended location(s) of inelastic action, and adescription of limit states controlling the strength and deformation capacityof the connection.

(3) Listing of systems for which connection is Prequalified: SMF, IMF or EBF.

(4) Listing of limits for all prequalification variables listed in Section P4.

(5) A detailed description of the design procedure for the connection, as requiredin Section P5.

(6) A list of references of test reports, research reports and other publications thatprovided the basis for prequalification.

(7) Summary of material strengths

(8) Summary of quality control procedures.


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37

APPENDIX S

QUALIFYING CYCLIC TESTS OF BEAM-TO-COLUMNAND LINK-TO-COLUMN CONNECTIONS

PART I

S1. SCOPE AND PURPOSE

This Appendix includes requirements for qualifying cyclic tests of beam-to-columnmoment connections in Moment Frames and Link-to-column connections in Ec-centrically Braced Frames, when required in these Provisions. The purpose of thetesting described in this Appendix is to provide evidence that a beam-to-columnconnection or a Link-to-column connection satisfies the requirements for strengthand Interstory Drift Angle or Link Rotation Angle in these Provisions. Alternativetesting requirements are permitted when approved by the Engineer of Record andthe Authority Having Jurisdiction.

This Appendix provides only minimum recommendations for simplified test con-ditions. If conditions in the actual building so warrant, additional testing shallbe performed to demonstrate satisfactory and reliable performance of momentconnections during actual earthquake motions.

S2. SYMBOLS

The numbers in parentheses after the definition of a symbol refers to the Sectionnumber in which the symbol is first used.

Interstory Drift Angle (S6) Link Rotation Angle (S6)

S3. DEFINITIONS

Complete Loading Cycle. A cycle of rotation taken from zero force to zero force,including one positive and one negative peak.

Interstory Drift Angle. Interstory displacement divided by story height, radians.

Inelastic Rotation. The permanent or plastic portion of the rotation angle be-tween a beam and the column or between a Link and the column of the TestSpecimen, measured in radians. The Inelastic Rotation shall be computedbased on an analysis of Test Specimen deformations. Sources of inelasticrotation include yielding of members, yielding of connection elements andconnectors, and slip between members and connection elements. For beam-to-column moment connections in Moment Frames, inelastic rotation shallbe computed based upon the assumption that inelastic action is concentrated


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38 PART I – ESSENTIAL TEST VARIABLES [App. S5.

at a single point located at the intersection of the centerline of the beam withthe centerline of the column. For Link-to-column connections in Eccentri-cally Braced Frames, inelastic rotation shall be computed based upon theassumption that inelastic action is concentrated at a single point located atthe intersection of the centerline of the Link with the face of the column.

Prototype. The connections, member sizes, steel properties, and other design,detailing, and construction features to be used in the actual building frame.

Test Specimen. A portion of a frame used for laboratory testing, intended tomodel the Prototype.

Test Setup. The supporting fixtures, loading equipment, and lateral bracing usedto support and load the Test Specimen.

Test Subassemblage. The combination of the Test Specimen and pertinent por-tions of the Test Setup.

S4. TEST SUBASSEMBLAGE REQUIREMENTS

The Test Subassemblage shall replicate as closely as is practical the conditions thatwill occur in the Prototype during earthquake loading. The Test Subassemblageshall include the following features:

(1) The Test Specimen shall consist of at least a single column with beams orLinks attached to one or both sides of the column.

(2) Points of inflection in the test assemblage shall coincide approximately withthe anticipated points of inflection in the Prototype under earthquake loading.

(3) Lateral bracing of the Test Subassemblage is permitted near load application orreaction points as needed to provide lateral stability of the Test Subassemblage.Additional lateral bracing of the Test Subassemblage is not permitted, unlessit replicates lateral bracing to be used in the Prototype.

S5. ESSENTIAL TEST VARIABLES

The Test Specimen shall replicate as closely as is practical the pertinent design,detailing, construction features, and material properties of the Prototype. The fol-lowing variables shall be replicated in the Test Specimen.

S5.1. Sources of Inelastic Rotation

Inelastic Rotation shall be developed in the Test Specimen by inelastic action inthe same members and connection elements as anticipated in the Prototype, i.e.,in the beam or Link, in the column Panel Zone, in the column outside of the PanelZone, or within connection elements. The fraction of the total Inelastic Rotationin the Test Specimen that is developed in each member or connection elementshall be at least 75 percent of the anticipated fraction of the total Inelastic Rotationin the Prototype that is developed in the corresponding member or connectionelement.


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App. S5.] PART I – ESSENTIAL TEST VARIABLES 39

S5.2. Size of Members

The size of the beam or Link used in the Test Specimen shall be within the followinglimits:

(1) The depth of the test beam or Link shall be no less than 90 percent of the depthof the Prototype beam or Link.

(2) The weight per foot of the test beam or Link shall be no less than 75 percentof the weight per foot of the Prototype beam or Link.

The size of the column used in the Test Specimen shall properly represent theinelastic action in the column, as per the requirements in Section S5.1. In addition,the depth of the test column shall be no less than 90 percent of the depth of thePrototype column.

Extrapolation beyond the limitations stated in this Section shall be permitted sub-ject to qualified peer review and approval by the Authority Having Jurisdiction.

S5.3. Connection Details

The connection details used in the Test Specimen shall represent the Prototypeconnection details as closely as possible. The connection elements used in the TestSpecimen shall be a full-scale representation of the connection elements used inthe Prototype, for the member sizes being tested.

S5.4. Continuity Plates

The size and connection details of Continuity Plates used in the Test Specimenshall be proportioned to match the size and connection details of Continuity Platesused in the Prototype connection as closely as possible.

S5.5. Material Strength

The following additional requirements shall be satisfied for each member or con-nection element of the Test Specimen that supplies Inelastic Rotation by yielding:

(1) The yield stress shall be determined by material tests on the actual materialsused for the Test Specimen, as specified in Section S8. The use of yield stressvalues that are reported on certified mill test reports are not permitted to beused for purposes of this Section.

(2) The yield stress of the beam shall not be more than 15 percent below Ry Fy forthe grade of steel to be used for the corresponding elements of the Prototype.Columns and connection elements with a tested yield stress shall not be morethan 15 percent above or below Ry Fy for the grade of steel to be used forthe corresponding elements of the Prototype. Ry Fy shall be determined inaccordance with Section 6.2.

S5.6. Welds

Welds on the Test Specimen shall satisfy the following requirements:

(1) Welding shall be performed in strict conformance with Welding ProcedureSpecifications (WPS) as required in AWS D1.1. The WPS essential variables


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40 PART I – LOADING HISTORY [App. S6.

shall meet the requirements in AWS D1.1 and shall be within the parametersestablished by the filler-metal manufacturer.

(2) The specified minimum tensile strength of the filler metal used for the TestSpecimen shall be the same as that to be used for the corresponding Prototypewelds.

(3) The specified minimum CVN toughness of the filler metal used for the TestSpecimen shall not exceed the specified minimum CVN toughness of the fillermetal to be used for the corresponding Prototype welds.

(4) The welding positions used to make the welds on the Test Specimen shall bethe same as those to be used for the Prototype welds.

(5) Details of weld backing, weld tabs, access holes, and similar items used for theTest Specimen welds shall be the same as those to be used for the correspondingPrototype welds. Weld backing and weld tabs shall not be removed from theTest Specimen welds unless the corresponding weld backing and weld tabsare removed from the Prototype welds.

(6) Methods of inspection and nondestructive testing and standards of acceptanceused for Test Specimen welds shall be the same as those to be used for thePrototype welds.

S5.7. Bolts

The bolted portions of the Test Specimen shall replicate the bolted portions of thePrototype connection as closely as possible. Additionally, bolted portions of theTest Specimen shall satisfy the following requirements:

(1) The bolt grade (e.g., ASTM A325, ASTM A490, ASTM F1852) used in theTest Specimen shall be the same as that to be used for the Prototype.

(2) The type and orientation of bolt holes (standard, oversize, short slot, long slot,or other) used in the Test Specimen shall be the same as those to be used forthe corresponding bolt holes in the Prototype.

(3) When Inelastic Rotation is to be developed either by yielding or by slip withina bolted portion of the connection, the method used to make the bolt holes(drilling, sub-punching and reaming, or other) in the Test Specimen shall bethe same as that to be used in the corresponding bolt holes in the Prototype.

(4) Bolts in the Test Specimen shall have the same installation (pretensioned orother) and faying surface preparation (no specified slip resistance, Class A, B,or C slip resistance, or other) as that to be used for the corresponding bolts inthe Prototype.

S6. LOADING HISTORY

S6.1. General Requirements

The Test Specimen shall be subjected to cyclic loads according to the requirementsprescribed in Section S6.2 for beam-to-column moment connections in Moment


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App. S7.] PART I – INSTRUMENTATION 41

Frames, and according to the requirements prescribed in Section S6.3 for link-to-column connections in Eccentrically Braced Frames.

Loading sequences other than those specified in Sections S6.2 and S6.3 may beused when they are demonstrated to be of equivalent or greater severity.

S6.2. Loading Sequence for Beam-to-Column Moment Connections

Qualifying cyclic tests of beam-to-column moment connections in Moment Framesshall be conducted by controlling the Interstory Drift Angle, , imposed on theTest Specimen, as follows:

(1) 6 cycles at = 0.00375 rad.

(2) 6 cycles at = 0.005 rad.

(3) 6 cycles at = 0.0075 rad.

(4) 4 cycles at = 0.01 rad.

(5) 2 cycles at = 0.015 rad.




Continue loading at increments of = 0.01 radians, with two cycles of loading ateach step.

S6.3. Loading Sequence for Link-to-Column Connections

Qualifying cyclic tests of link-to-column moment connections in EccentricallyBraced Frames shall be conducted by controlling the Link Rotation Angle, ,imposed on the Test Specimen, as follows:

(1) 3 cycles at = 0.0025 rad.

(2) 3 cycles at = 0.005 rad.





Continue loading at increments of = 0.01 radians, with two cycles of loading ateach step.

S7. INSTRUMENTATION

Sufficient instrumentation shall be provided on the Test Specimen to permit mea-surement or calculation of the quantities listed in Section S9.


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42 PART I – TEST REPORTING REQUIREMENTS [App. S9.

S8. MATERIALS TESTING REQUIREMENTS

S8.1. Tension Testing Requirements

Tension testing shall be conducted on samples of steel taken from the materialadjacent to each Test Specimen. Tension-test results from certified mill test reportsshall be reported but are not permitted to be used in place of specimen testing forthe purposes of this Section. Tension-test results shall be based upon testing thatis conducted in accordance with Section S8.2. Tension testing shall be conductedand reported for the following portions of the Test Specimen:

(1) Flange(s) and web(s) of beams and columns at standard locations.

(2) Any element of the connection that supplies Inelastic Rotation by yielding.

S8.2. Methods of Tension Testing

Tension testing shall be conducted in accordance with ASTM A6/A6M, ASTMA370, and ASTM E8, with the following exceptions:

(1) The yield stress Fy that is reported from the test shall be based upon the yieldstrength definition in ASTM A370, using the offset method at 0.002 strain.

(2) The loading rate for the tension test shall replicate, as closely as practical, theloading rate to be used for the Test Specimen.

S9. TEST REPORTING REQUIREMENTS

For each Test Specimen, a written test report meeting the requirements of the Au-thority Having Jurisdiction and the requirements of this Section shall be prepared.The report shall thoroughly document all key features and results of the test. Thereport shall include the following information:

(1) A drawing or clear description of the Test Subassemblage, including keydimensions, boundary conditions at loading and reaction points, and locationof lateral braces.

(2) A drawing of the connection detail showing member sizes, grades of steel,the sizes of all connection elements, welding details including filler metal,the size and location of bolt holes, the size and grade of bolts, and all otherpertinent details of the connection.

(3) A listing of all other Essential Variables for the Test Specimen, as listed inSection S5.

(4) A listing or plot showing the applied load or displacement history of the TestSpecimen.

(5) A plot of the applied load versus the displacement of the Test Specimen. Thedisplacement reported in this plot shall be measured at or near the point ofload application. The locations on the Test Specimen where the loads anddisplacements were measured shall be clearly indicated.

(6) A plot of beam moment versus Interstory Drift Angle for beam-to-columnmoment connections; or a plot of Link shear force versus Link Rotation Anglefor link-to-column connections. For beam-to-column connections, the beam


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App. S10.] PART I – ACCEPTANCE CRITERIA 43

moment and the Interstory Drift Angle shall be computed with respect to thecenterline of the column.

(7) The Interstory Drift Angle and the total Inelastic Rotation developed by theTest Specimen. The components of the Test Specimen contributing to thetotal Inelastic Rotation due to yielding or slip shall be identified. The portionof the total Inelastic Rotation contributed by each component of the TestSpecimen shall be reported. The method used to compute Inelastic Rotationsshall be clearly shown.

(8) A chronological listing of significant test observations, including observa-tions of yielding, slip, instability, and fracture of any portion of the TestSpecimen as applicable.

(9) The controlling failure mode for the Test Specimen. If the test is terminatedprior to failure, the reason for terminating the test shall be clearly indicated.

(10) The results of the material tests specified in Section S8.

(11) The Welding Procedure Specifications (WPS) and welding inspectionreports.

Additional drawings, data, and discussion of the Test Specimen or test results arepermitted to be included in the report.

S10. ACCEPTANCE CRITERIA

The Test Specimen must satisfy the strength and Interstory Drift Angle or LinkRotation Angle requirements of these Provisions for the SMF, IMF, or EBF con-nection, as applicable. The Test Specimen must sustain the required InterstoryDrift Angle or Link Rotation Angle for at least one complete loading cycle.


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44

APPENDIX X

WELD METAL / WELDING PROCEDURESPECIFICATION TOUGHNESS VERIFICATION TEST

PART I

Preamble: This appendix provides a procedure for qualifying the weld metal toughnessand is included on an interim basis pending adoption of such a procedure by AWS or otheraccredited organization.

X1. SCOPE

This appendix provides a standard method for qualification testing of weld fillermetals required to have specified notch toughness for service in specified joints insteel Moment Frames for seismic applications.

Testing of weld metal to be used in production shall be performed by filler metalmanufacturer’s production lot, as defined in AWS A5.01, Filler Metal ProcurementGuidelines, as follows:

(1) Class C3 for SMAW electrodes,

(2) Class S2 for GMAW-S and SAW electrodes,

(3) Class T4 for FCAW and GMAW-C, or

(4) Class F2 for SAW fluxes.

Alternatively, filler metal manufacturers approved for production of products meet-ing the above requirements, under a program acceptable to the Engineer, need notconduct the mechanical A5 tests or the Weld Metal / Weld Procedure Specification(WPS) Toughness Verification Test, or require lot control for each lot, and mayrely upon the Manufacturer’s certifications that the product meets the specifiedperformance requirements.

X2. TEST CONDITIONS

Tests shall be conducted at the range of heat inputs for which the weld filler metalwill be qualified under the WPS. It is recommended that tests be conducted at theLow Heat Input Level and High Heat Input Level indicated in Table I-X-1.

Alternatively, the filler metal manufacturer or Contractor may elect to test a wideror narrower range of heat inputs and interpass temperatures. The range of heatinputs and interpass temperatures tested shall be clearly stated on the test reportsand user data sheets. Regardless of the method of selecting test heat input, the WPS,as used by the contractor, shall fall within the range of heat inputs and interpasstemperatures tested.


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App. X4.] PART I – ACCEPTANCE CRITERIA 45

TABLE I-X-1WPS Toughness Verification Test Welding

and Preheat ConditionsCooling Rate Heat Input Preheat ˚F (˚C) Interpass ˚F (˚C)

Low Heat Input Test 30 kJ/in. (1.2 kJ/mm) 70 ± 25 (21 ± 14) 200 ± 50 (93 ± 28)

High Heat Input Test 80 kJ/in. (3.1 kJ/mm) 300 ± 25 (149 ± 14) 500 ± 50 (260 ± 28)

X3. TEST SPECIMENS

Two test plates, one for each heat input level shall be used, and five Charpy V-Notch (CVN) test specimens shall be made per plate. Each plate shall be steel, ofany AISC-listed structural grade. The test plate shall be 3/4 in. (19 mm) thick witha 1/2-inch (13 mm) root opening and 45˚ included groove angle. The test plate andspecimens shall be as shown in Figure 2A in AWS A5.20-95, or as in Figure 5 inAWS A5.29-98. Except for the root pass, a minimum of two passes per layer shallbe used to fill the width.

All test specimens shall be taken from near the centerline of the weld at the mid-thickness location, in order to minimize dilution effects. CVN specimens shallbe prepared in accordance with AWS B4.0-92, Standard Methods for MechanicalTesting of Welds, Section A3. The test assembly shall be restrained during welding,or preset at approximately 5 degrees to prevent warpage in excess of 5 degrees.A welded test assembly that has warped more than 5 degrees shall be discarded.Welded test assemblies shall not be straightened.

The test assembly shall be tack welded and heated to the specified preheat tem-perature, measured by temperature indicating crayons or surface temperature ther-mometers one inch from the center of the groove at the location shown in thefigures cited above. Welding shall continue until the assembly has reached theinterpass temperature prescribed in Table I-X-1. The interpass temperature shallbe maintained for the remainder of the weld. Should it be necessary to interruptwelding, the assembly shall be allowed to cool in air. The assembly shall then beheated to the prescribed interpass temperature before welding is resumed.

X4. ACCEPTANCE CRITERIA

The lowest and highest CVN toughness values obtained from the five specimensfrom a single test plate shall be disregarded. Two of the remaining three valuesshall equal, or exceed, the specified toughness of 40 ft-lbf (54 J) energy level atthe testing temperature. One of the three may be lower, but not lower than 30 ft-lbf(41 J), and the average of the three shall not be less than the required 40 ft-lbf(54 J) energy level. All test samples shall meet the notch toughness requirementsfor the electrodes as provided in Section 7.3b.


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46


P1: MRM/SPH P2: MRM/UKS QC: MRM/UKS T1: MRM

PB262A-02 PB262/Seismic.cls January 2, 2003 11:19 Char Count=

47PART II. COMPOSITE STRUCTURAL STEELAND REINFORCED CONCRETE BUILDINGS

GLOSSARY

The following glossary terms are applicable to Part II and are in addition to those givenin the Part I Glossary.

Boundary Member. Portion along wall and diaphragm edges strengthened with structuralsteel sections and/or longitudinal steel reinforcement and transverse reinforcement.

Collector Element. Member that serves to transfer forces between floor diaphragms andthe members of the Seismic Force Resisting System.

Composite Beam. A structural steel beam that is either an unencased steel beam thatacts integrally with a concrete or composite slab using shear connectors or a fullyreinforced-concrete-encased steel beam.

Composite Brace. A reinforced-concrete-encased structural steel section (rolled or built-up) or concrete-filled steel section that is used as a brace.

Composite Column. A reinforced-concrete-encased structural steel section (rolled orbuilt-up) or concrete-filled steel section that is used as a column.

Composite Plate -Concrete Shear Wall. A wall that consists of a steel plate with reinforcedconcrete encasement on one or both sides that provides out-of-plane stiffening to preventbuckling of the steel plate.

Composite Shear Wall. A reinforced concrete wall that has unencased or reinforced-concrete-encased structural steel sections as Boundary Members.

Composite Slab. A concrete slab that is supported on and bonded to a formed steel deckand that acts as a diaphragm to transfer force to and between elements of the SeismicForce Resisting System.

Concrete-Filled Composite Column. Round or rectangular structural steel section that isfilled with concrete.

Coupling Beam. A structural steel or Composite Beam that connects adjacent reinforcedconcrete wall elements so that they act together to resist lateral forces.

Encased Composite Beam. A structural steel beam that is completely encased in rein-forced concrete that is cast integrally with the slab and for which full composite actionis provided by bond between the structural steel and reinforced concrete.

Encased Composite Column. A structural steel column (rolled or built-up) that is com-pletely encased in reinforced concrete.

Face Bearing Plates. Stiffeners that are attached to structural steel beams that are em-bedded in reinforced concrete walls or columns. The plates are located at the face ofthe reinforced concrete to provide confinement and to transfer forces to the concretethrough direct bearing.

Fully Composite Beam. A Composite Beam that has a sufficient number of shear connec-tors to develop the nominal plastic flexural strength of the composite section.




48 PART II – GLOSSARY

Load-Carrying Reinforcement. Reinforcement in composite members that is designed anddetailed to resist the required loads.

Partially Composite Beam. An unencased Composite Beam with a nominal flexuralstrength that is controlled by the strength of the shear stud connectors.

Partially Restrained Composite Connection. Partially Restrained connections as definedin the LRFD Specification that connect partially or Fully Composite Beams to steelcolumns with flexural resistance provided by a force couple achieved with steel rein-forcement in the slab and a steel seat angle or similar connection at the bottom flange.

Reinforced-Concrete-Encased Shapes. Structural steel sections that are encased in rein-forced concrete.

Restraining Bars. Steel reinforcement in composite members that is not designed to carryrequired forces, but is provided to facilitate the erection of other steel reinforcement andto provide anchorage for stirrups or ties. Generally, such reinforcement is not splicedto be continuous.




Sect. 5.] PART II – MATERIALS 49

1. SCOPE

These Provisions are intended for the design and construction of composite struc-tural steel and reinforced concrete members and connections in the Seismic LoadResisting Systems in buildings for which the design forces resulting from earth-quake motions have been determined on the basis of various levels of energydissipation in the inelastic range of response.

Provisions shall be applied in conjunction with the AISC Load and ResistanceFactor Design (LRFD) Specification for Structural Steel Buildings, hereinafterreferred to as the LRFD Specification. All members and connections in the SeismicLoad Resisting System shall have a Design Strength as required in the LRFDSpecification and shall meet the requirements in these Provisions. The applicablerequirements in Part I shall be used for the design of structural steel componentsin composite systems. Reinforced-concrete members subjected to seismic forcesshall meet the requirements in ACI 318, except as modified in these provisions.When the design is based upon elastic analysis, the stiffness properties of thecomponent members of composite systems shall reflect their condition at the onsetof significant yielding of the building.

Part II includes a Glossary, which is specifically applicable to this Part. The Part IGlossary is also applicable to Part II.


The documents referenced in these provisions shall include those listed in Part ISection 2 with the following additions and modifications:

American Society of Civil EngineersStandard for the Structural Design of Composite Slabs, ASCE 3-91

American Welding SocietyStructural Welding Code-Reinforcing Steel, AWS D1.4-98

3. SEISMIC DESIGN CATEGORIES

The Required Strength and other seismic provisions for Seismic Design Categories,Seismic Use Groups or Seismic Zones and the limitations on height and irregularityshall be as stipulated in the Applicable Building Code.


The loads and load combinations shall be as stipulated by the Applicable BuildingCode (see Glossary). Where Amplified Seismic Loads are required by these pro-visions, the horizontal earthquake load E (as defined in the Applicable BuildingCode) shall be multiplied by the overstrength factor o prescribed by the Appli-cable Building Code. In the absence of a specific definition of o, the value foro shall be as listed in Table II-4-1.

5. MATERIALS

5.1. Structural Steel

Structural steel used in composite Seismic Load Resisting Systems shall meet therequirements in LRFD Specification Section A3.1a. Structural steel used in the




50 PART II – COMPOSITE MEMBERS [Sect. 6.

TABLE II-4-1System Overstrength Factor, Ωo

Seismic Load Resisting System Ωo

All moment-frame systems meeting Part II requirements 3

All Eccentrically Braced Frames (EBF) 21/2

and wall systems meeting Part II requirements

All other systems meeting Part II requirements 2

composite Seismic Force Resisting Systems described in Sections 8, 9, 13, 14, 16and 17 shall also meet the requirements in Part I Section 6.

5.2. Concrete and Steel Reinforcement

Concrete and steel reinforcement used in composite Seismic Load Resisting Sys-tems shall meet the requirements in ACI 318, excluding Chapters 21 and 22, andthe following requirements:

(1) The specified minimum compressive strength of concrete in composite mem-bers shall equal or exceed 2.5 ksi (17 MPa).

(2) For the purposes of determining the Nominal Strength of composite members,f ′

c shall not be taken as greater than 10 ksi (69 MPa) for normal-weightconcrete nor 4 ksi (28 MPa) for lightweight concrete.

Concrete and steel reinforcement used in the composite Seismic Load ResistingSystems described in Sections 8, 9, 13, 14, 16, and 17 shall also meet the require-ments in ACI 318 Chapter 21.

6. COMPOSITE MEMBERS

6.1. Scope

The design of composite members in the Seismic Load Resisting Systems de-scribed in Sections 8 through 17 shall meet the requirements in this Section andthe material requirements in Section 5.

6.2. Composite Floor and Roof Slabs

The design of composite floor and roof slabs shall meet the requirements ofASCE 3. Composite slab diaphragms shall meet the requirements in this Section.

Details shall be designed to transfer forces between the diaphragm and BoundaryMembers, Collector Elements, and elements of the horizontal framing system.

The nominal shear strength of composite diaphragms and concrete-filled steeldeck diaphragms shall be taken as the nominal shear strength of the reinforcedconcrete above the top of the steel deck ribs in accordance with ACI 318 excludingChapter 22. Alternatively, the composite diaphragm design shear strength shall bedetermined by in-plane shear tests of concrete-filled diaphragms.




Sect. 6.] PART II – COMPOSITE MEMBERS 51

6.3. Composite Beams

Composite Beams shall meet the requirements in LRFD Specification Chapter I.Composite Beams that are part of C-SMF as described in Section 9 shall also meetthe following requirements:

(1) The distance from the maximum concrete compression fiber to the plasticneutral axis shall not exceed:

Ycon + db

1 +(

1700 Fy

Es

) (6-1)

where

Ycon = distance from the top of the steel beam to the top of concrete, in. (mm)db = depth of the steel beam, in. (mm)Fy = specified minimum yield strength of the steel beam, ksi (MPa)Es = modulus of elasticity of the steel beam, ksi (MPa)

(2) Beam flanges shall meet the requirements in Part I Section 9.4, except whenfully reinforced-concrete-encased compression elements have a reinforcedconcrete cover of at least 2 in. (50 mm) and confinement is provided by hoopreinforcement in regions where plastic hinges are expected to occur underseismic deformations. Hoop reinforcement shall meet the requirements inACI 318 Section 21.3.3.

6.4. Reinforced-Concrete-Encased Composite Columns

This Section is applicable to columns that: (1) consist of reinforced-concrete-encased structural steel sections with a structural steel area that comprises at least4 percent of the total composite-column cross-section; and (2) meet the additionallimitations in LRFD Specification Section I2.1. Such columns shall meet the re-quirements in LRFD Specification Chapter I, except as modified in this Section.Additional requirements, as specified for intermediate and special seismic sys-tems in Sections 6.4b and 6.4c, shall apply as required in the descriptions of thecomposite seismic systems in Sections 8 through 17.

Columns that consist of reinforced-concrete-encased structural steel sections witha structural steel area that comprises less than 4 percent of the total composite-column cross-section shall meet the requirements for reinforced concrete columnsin ACI 318 except as modified for:

(1) The steel shape shear connectors in Section 6.4a(2).

(2) The contribution of the reinforced-concrete-encased structural steel section tothe strength of the column as provided in ACI 318.

(3) The seismic requirements for reinforced concrete columns as specified in thedescription of the composite seismic systems in Sections 8 through 17.

6.4a. Ordinary Seismic System Requirements

The following requirements for Reinforced-Concrete-Encased CompositeColumns are applicable to all composite systems:





(1) The nominal shear strength of the column shall be determined as the nominalshear strength of the structural shape plus the nominal shear strength that isprovided by the tie reinforcement in the reinforced-concrete encasement. Thenominal shear strength of the structural steel section shall be determined inaccordance with LRFD Specification Section F2. The nominal shear strengthof the tie reinforcement shall be determined in accordance with ACI 318Sections 11.5.6.2 through 11.5.6.9. In ACI 318 Sections 11.5.6.5 and 11.5.6.9,the dimension bw shall equal the width of the concrete cross-section minusthe width of the structural shape measured perpendicular to the direction ofshear. The nominal shear strength shall be multiplied by v equal to 0.75 todetermine the design shear strength.

(2) Composite Columns that are designed to share the applied loads between thestructural steel section and reinforced concrete shall have shear connectorsthat meet the following requirements:

(a) If an external member is framed directly to the structural steel sectionto transfer a vertical reaction Vu , shear connectors shall be provided totransfer the force Vu(1 − As Fy /Pn) between the structural steel section andthe reinforced concrete, where As is the area of the structural steel section,Fy is the specified minimum yield strength of the structural steel section,and Pn is the nominal compressive strength of the Composite Column.

(b) If an external member is framed directly to the reinforced concrete to trans-fer a vertical reaction Vu , shear connectors shall be provided to transfer theforce Vu As Fy /Pn between the structural steel section and the reinforcedconcrete, where As , Fy and Pn are as defined above.

(c) The maximum spacing of shear connectors shall be 16 in. (406 mm) withattachment along the outside flange faces of the embedded shape.

(3) The maximum spacing of transverse ties shall be the least of the following:

(a) one-half the least dimension of the section

(b) 16 longitudinal bar diameters

(c) 48 tie diameters

Transverse ties shall be located vertically within one-half the tie spacing abovethe top of the footing or lowest beam or slab in any story and shall be spacedas provided herein within one-half the tie spacing below the lowest beam orslab framing into the column.

Transverse bars shall have a diameter that is not less than one-fiftieth of greatestside dimension of the composite member, except that ties shall not be smallerthan No. 3 bars and need not be larger than No. 5 bars. Alternatively, weldedwire fabric of equivalent area is permitted as transverse reinforcement exceptwhen prohibited for intermediate and special systems.

(4) All Load-Carrying Reinforcement shall meet the detailing and splice require-ments in ACI 318 Sections 7.8.1 and 12.17. Load-Carrying Reinforcementshall be provided at every corner of a rectangular cross-section. The maxi-mum spacing of other load carrying or restraining longitudinal reinforcementshall be one-half of the least side dimension of the composite member.




Sect. 6.] PART II – COMPOSITE MEMBERS 53

(5) Splices and end bearing details for reinforced-concrete-encased structural steelsections shall meet the requirements in the LRFD Specification and ACI 318Section 7.8.2. If adverse behavioral effects due to the abrupt change in memberstiffness and nominal tensile strength occur when reinforced-concrete encase-ment of a structural steel section is terminated, either at a transition to a purereinforced concrete column or at the Column Base, they shall be consideredin the design.

6.4b. Intermediate Seismic System Requirements

Reinforced-Concrete-Encased Composite Columns in intermediate seismic sys-tems shall meet the following requirements in addition to those in Section 6.4a:

(1) The maximum spacing of transverse bars at the top and bottom shall be theleast of the following:

(a) one-half the least dimension of the section

(b) 8 longitudinal bar diameters

(c) 24 tie bar diameters

(d) 12 in. (305 mm)

These spacings shall be maintained over a vertical distance equal to the greatestof the following lengths, measured from each joint face and on both sides ofany section where flexural yielding is expected to occur:

(a) one-sixth the vertical clear height of the column

(b) the maximum cross-sectional dimension

(c) 18 in. (457 mm)

(2) Tie spacing over the remaining column length shall not exceed twice thespacing defined above.

(3) Welded wire fabric is not permitted as transverse reinforcement in intermediateseismic systems.

6.4c. Special Seismic System Requirements

Reinforced-concrete-encased columns for special seismic systems shall meet thefollowing requirements in addition to those in Sections 6.4a and 6.4b:

(1) The required axial strength for Reinforced-Concrete-Encased CompositeColumns and splice details shall meet the requirements in Part I Section 8.

(2) Longitudinal Load-Carrying Reinforcement shall meet the requirements inACI 318 Section 21.4.3.

(3) Transverse reinforcement shall be hoop reinforcement as defined in ACI 318Chapter 21 and shall meet the following requirements:

(a) The minimum area of tie reinforcement Ash shall meet the followingrequirement:

Ash = 0.09hccs

(1 − Fy As

Pn

) (f ′c

Fyh

)(6-2)





where

hcc = cross-sectional dimension of the confined core measuredcenter-to-center of the tie reinforcement, in. (mm)

s = spacing of transverse reinforcement measured along thelongitudinal axis of the structural member, in. (mm)

Fy = specified minimum yield strength of the structural steel core,ksi (MPa)

As = cross-sectional area of the structural core, in.2 (mm2)Pn = nominal axial compressive strength of the Composite Column

calculated in accordance with the LRFD Specification, kips (N)f ′c = specified compressive strength of concrete, ksi (MPa)

Fyh = specified minimum yield strength of the ties, ksi (MPa)

Equation 6-2 need not be satisfied if the Nominal Strength of the reinforced-concrete-encased structural steel section alone is greater than 1.0D + 0.5L .

(b) The maximum spacing of transverse reinforcement along the length of thecolumn shall be the lesser of 6 longitudinal load-carrying bar diametersand 6 in. (152 mm)

(c) When specified in Sections 6.4c(4), 6.4c(5) or 6.4c(6), the maximumspacing of transverse reinforcement shall be the lesser of one-fourththe least member dimension and 4 in. (102 mm). For this reinforce-ment, cross ties, legs of overlapping hoops, and other confining rein-forcement shall be spaced not more than 14 in. on center in the transversedirection.

(4) Reinforced-Concrete-Encased Composite Columns in Braced Frames withaxial compression forces that are larger than 0.2 times Po shall have trans-verse reinforcement as specified in Section 6.4c(3)(c) over the total ele-ment length. This requirement need not be satisfied if the Nominal Strengthof the reinforced-concrete-encased steel section alone is greater than1.0D + 0.5L .

(5) Composite Columns supporting reactions from discontinued stiff members,such as walls or Braced Frames, shall have transverse reinforcement as speci-fied in Section 6.4c(3)(c) over the full length beneath the level at which the dis-continuity occurs if the axial compression force exceeds 0.1 times Po. Trans-verse reinforcement shall extend into the discontinued member for at least thelength required to develop full yielding in the reinforced-concrete-encasedstructural steel section and longitudinal reinforcement. This requirement neednot be satisfied if the Nominal Strength of the reinforced-concrete-encasedstructural steel section alone is greater than 1.0D + 0.5L .

(6) Reinforced-Concrete-Encased Composite Columns that are used in C-SMFshall meet the following requirements:

(a) Transverse reinforcement shall meet the requirements in Section 6.4c(3)(c)at the top and bottom of the column over the region specified in Sec-tion 6.4b.

(b) The strong-column/weak-beam design requirements in Section 9.5 shallbe satisfied. Column Bases shall be detailed to sustain inelastic flexuralhinging.




Sect. 7.] PART II – COMPOSITE CONNECTIONS 55

(c) The minimum required shear strength of the column shall meet the re-quirements in ACI 318 Section 21.4.5.1.

(7) When the column terminates on a footing or mat foundation, the transversereinforcement as specified in this section shall extend into the footing or matat least 12 in. (305 mm). When the column terminates on a wall, the transversereinforcement shall extend into the wall for at least the length required to de-velop full yielding in the reinforced-concrete-encased structural steel sectionand longitudinal reinforcement.

(8) Welded wire fabric is not permitted as transverse reinforcement for specialseismic systems.

6.5. Concrete-Filled Composite Columns

This Section is applicable to columns that: (1) consist of concrete-filled steel rect-angular or circular hollow structural sections (HSS) with a structural steel areathat comprises at least 4 percent of the total composite-column cross-section; and(2) meet the additional limitations in LRFD Specification Section I2.1. Suchcolumns shall be designed to meet the requirements in LRFD SpecificationChapter I, except as modified in this Section.

The design shear strength of the Composite Column shall be the design shearstrength of the structural steel section alone.

In the special seismic systems described in Sections 9, 13 and 14, members andcolumn splices for Concrete-Filled Composite Columns shall also meet the re-quirements in Part I Section 8.

Concrete-Filled Composite Columns used in C-SMF shall meet the followingadditional requirements:

(1) The minimum required shear strength of the column shall meet the require-ments in ACI 318 Section 21.4.5.1.

(2) The strong-column/weak-beam design requirements in Section 9.5 shall bemet. Column Bases shall be designed to sustain inelastic flexural hinging.

(3) The minimum wall thickness of concrete-filled rectangular HSS shall equal

b√

Fy/ (2Es) (6-3)

for the flat width b of each face, where b is as defined in LRFD SpecificationTable B5.1.

7. COMPOSITE CONNECTIONS

7.1. Scope

This Section is applicable to connections in buildings that utilize composite or dualsteel and concrete systems wherein seismic force is transferred between structuralsteel and reinforced concrete components.

Composite connections shall be demonstrated to have Design Strength, ductilityand toughness that is comparable to that exhibited by similar structural steel or




56 PART II – COMPOSITE CONNECTIONS [Sect. 7.

reinforced concrete connections that meet the requirements in Part I and ACI318, respectively. Methods for calculating the connection strength shall meet therequirements in this Section.

7.2. General Requirements

Connections shall have adequate deformation capacity to resist the critical Re-quired Strengths at the Design Story Drift. Additionally, connections that are re-quired for the lateral stability of the building under seismic forces shall meet therequirements in Sections 8 through 17 based upon the specific system in whichthe connection is used. When the Required Strength is based upon nominal mate-rial strengths and nominal member dimensions, the determination of the requiredconnection strength shall account for any effects that result from the increase inthe actual Nominal Strength of the connected member.

7.3. Nominal Strength of Connections

The Nominal Strength of connections in composite Structural Systems shall bedetermined on the basis of rational models that satisfy both equilibrium of internalforces and the strength limitation of component materials and elements based uponpotential limit states. Unless the connection strength is determined by analysisand testing, the models used for analysis of connections shall meet the followingrequirements:

(1) When required, force shall be transferred between structural steel and rein-forced concrete through direct bearing of headed shear studs or suitable alter-native devices, by other mechanical means, by shear friction with the necessaryclamping force provided by reinforcement normal to the plane of shear trans-fer, or by a combination of these means. Any potential bond strength betweenstructural steel and reinforced concrete shall be ignored for the purpose of theconnection force transfer mechanism.

(2) The nominal bearing and shear-friction strengths shall meet the requirementsin ACI 318 Chapters 10 and 11, except that the strength reduction (resistance)factors shall be as given in ACI 318. Unless a higher strength is substanti-ated by cyclic testing, the nominal bearing and shear-friction strengths shallbe reduced by 25 percent for the composite seismic systems described inSections 9, 13, 14, 16, and 17.

(3) The Design Strengths of structural steel components in composite connections,as determined in Part I and the LRFD Specification, shall equal or exceed theRequired Strengths. Structural steel elements that are encased in confinedreinforced concrete are permitted to be considered to be braced against out-of-plane buckling. Face Bearing Plates consisting of stiffeners between theflanges of steel beams are required when beams are embedded in reinforcedconcrete columns or walls.

(4) The nominal shear strength of reinforced-concrete-encased steel Panel Zonesin beam-to-column connections shall be calculated as the sum of the Nomi-nal Strengths of the structural steel and confined reinforced concrete shearelements as determined in Part I Section 9.3 and ACI 318 Section 21.5,




Sect. 8.] PART II – COMPOSITE PARTIALLY RESTRAINED (PR) MOMENT FRAMES (C-PRMF) 57

respectively. The strength reduction (resistance) factors for reinforced con-crete shall be as given in ACI 318.

(5) Reinforcement shall be provided to resist all tensile forces in reinforced con-crete components of the connections. Additionally, the concrete shall be con-fined with transverse reinforcement. All reinforcement shall be fully developedin tension or compression, as appropriate, beyond the point at which it is nolonger required to resist the forces. Development lengths shall be determinedin accordance with ACI 318 Chapter 12. Additionally, development lengthsfor the systems described in Sections 9, 13, 14, 16 and 17 shall meet the re-quirements in ACI 318 Section 21.5.4. Connections shall meet the followingadditional requirements:

(a) When the slab transfers horizontal diaphragm forces, the slab reinforce-ment shall be designed and anchored to carry the in-plane tensile forces atall critical sections in the slab, including connections to collector beams,columns, braces and walls.

(b) For connections between structural steel or Composite Beams and re-inforced concrete or Reinforced-Concrete-Encased Composite Columns,transverse hoop reinforcement shall be provided in the connection regionto meet the requirements in ACI 318 Section 21.5, except for the followingmodifications:

(i) Structural steel sections framing into the connections are consideredto provide confinement over a width equal to that of face bearingstiffener plates welded to the beams between the flanges.

(ii) Lap splices are permitted for perimeter ties when confinement of thesplice is provided by Face Bearing Plates or other means that preventsspalling of the concrete cover in the systems described in Sections 10,11, 12 and 15.

(c) The longitudinal bar sizes and layout in reinforced concrete and CompositeColumns shall be detailed to minimize slippage of the bars through thebeam-to-column connection due to high force transfer associated with thechange in column moments over the height of the connection.

8. COMPOSITE PARTIALLY RESTRAINED (PR)MOMENT FRAMES (C-PRMF)

8.1. Scope

This Section is applicable to frames that consist of structural steel columns andComposite Beams that are connected with Partially Restrained (PR) moment con-nections that meet the requirements in LRFD Specification Section A2. C-PRMFshall be designed so that under earthquake loading yielding occurs in the ductilecomponents of the composite PR beam-to-column moment connections. Limitedyielding is permitted at other locations, such as the Column Base connection.Connection flexibility and Composite Beam action shall be accounted for in deter-mining the dynamic characteristics, strength and drift of C-PRMF. C-PRMF shallmeet the requirements of this section.




58 PART II – COMPOSITE SPECIAL MOMENT FRAMES (C-SMF) [Sect. 9.

8.2. Columns

Structural steel columns shall meet the requirements in Part I Section 8 and theLRFD Specification. The effect of PR moment connections on stability of indivi-dual columns and the overall frame shall be considered in C-PRMF.

8.3. Composite Beams

Composite Beams shall meet the requirements in LRFD Specification Chapter I.For the purposes of analysis, the stiffness of beams shall be determined with aneffective moment of inertia of the composite section.

8.4. Partially Restrained (PR) Moment Connections

The Required Strength for the beam-to-column PR moment connections shallbe determined from the load combinations stipulated by the Applicable BuildingCode, including consideration of the effects of connection flexibility and second-order moments. In addition, composite connections shall have a Nominal Strengththat is at least equal to 50 percent of Mp, where Mp is the nominal plastic flex-ural strength of the connected structural steel beam ignoring composite action.Connections shall meet the requirements in Section 7 and shall have an inelasticrotation capacity of 0.015 radians and a total rotation capacity of 0.03 radians thatis substantiated by cyclic testing as described in Part I Section 9.2a.

9. COMPOSITE SPECIAL MOMENT FRAMES (C-SMF)

9.1. Scope

This Section is applicable to moment-resisting frames that consist of either com-posite or reinforced concrete columns and either structural steel or CompositeBeams. C-SMF shall be designed assuming that under the Design Earthquake sig-nificant inelastic deformations will occur, primarily in the beams, but with limitedinelastic deformations in the columns and/or connections. C-SMF shall meet therequirements of this section.

9.2. Columns

Composite Columns shall meet the requirements for special seismic systems inSections 6.4 or 6.5. Reinforced concrete columns shall meet the requirements inACI 318 Chapter 21, excluding Section 21.10.

9.3. Beams

Composite Beams shall meet the requirements in Section 6.3. Neither structuralsteel nor composite trusses are permitted as flexural members to resist seismicloads in C-SMF unless it is demonstrated by testing and analysis that the particularsystem provides adequate ductility and energy dissipation capacity.

9.4. Moment Connections

The Required Strength of beam-to-column moment connections shall be deter-mined from the shear and flexure associated with the nominal plastic flexural




Sect. 10.] PART II – COMPOSITE INTERMEDIATE MOMENT FRAMES (C-IMF) 59

strength of the beams framing into the connection. The nominal connection strengthshall meet the requirements in Section 7. In addition, the connections shall be ca-pable of sustaining an inelastic beam rotation of 0.03 radians. When the beamflanges are interrupted at the connection, the inelastic rotation capacity shall bedemonstrated as specified in Part I Section 9 for connections in SMF. For connec-tions to reinforced concrete columns with a beam that is continuous through thecolumn so that welded joints are not required in the flanges and the connectionis not otherwise susceptible to premature fractures, the inelastic rotation capacityshall be demonstrated by testing or other substantiating data.


The minimum flexural strength and design of reinforced concrete columns shallmeet the requirements in ACI 318 Section 21.4.2. The minimum flexural strengthand design of Composite Columns shall meet the requirements in Part I Section 9.6with the following modifications:

(1) The flexural strength of the Composite Column M∗pc shall meet the require-

ments in LRFD Specification Chapter I with consideration of the applied axialload, Pu .

(2) The force limit for the exceptions in Part I Section 9.6(a) shall be Pu < 0.1Po.

(3) Composite Columns exempted by the minimum flexural strength requirementin Part I Section 9.6 shall have transverse reinforcement that meets the re-quirements in Section 6.4c(4).

10. COMPOSITE INTERMEDIATE MOMENT FRAMES (C-IMF)

10.1. Scope

This Section is applicable to moment resisting frames that consist of either compos-ite or reinforced concrete columns and either structural steel or Composite Beams.C-IMF shall be designed assuming that under the Design Earthquake inelasticdeformation will occur primarily in the beams but with moderate inelastic defor-mation in the columns and/or connections. C-IMF shall meet the requirements ofthis section.

10.2. Columns

Composite Columns shall meet the requirements for intermediate seismic systemsin Section 6.4 or 6.5. Reinforced concrete columns shall meet the requirements inACI 318 Section 21.10.

10.3. Beams

Structural steel and Composite Beams shall meet the requirements in the LRFDSpecification.


The nominal connection strength shall meet the requirements in Section 7. TheRequired Strength of beam-to-column connections shall meet one of the following




60 PART II – COMPOSITE ORDINARY BRACED FRAMES (C-OBF) [Sect. 12.

requirements:

(a) The connection Design Strength shall meet or exceed the forces associatedwith plastic hinging of the beams adjacent to the connection.

(b) The connection Design Strength shall meet or exceed the Required Strengthgenerated by load combinations stipulated by the Applicable Building Code,including the Amplified Seismic Load.

(c) The connections shall demonstrate an inelastic rotation capacity of at least0.02 radians in cyclic tests.

11. COMPOSITE ORDINARY MOMENT FRAMES (C-OMF)

11.1. Scope

This Section is applicable to moment resisting frames that consist of either com-posite or reinforced concrete columns and structural steel or Composite Beams.C-OMF shall be designed assuming that under the Design Earthquake limited in-elastic action will occur in the beams, columns and/or connections. C-OMF shallmeet the requirements of this section.

11.2. Columns

Composite Columns shall meet the requirements for ordinary seismic systems inSection 6.4 or 6.5. Reinforced concrete columns shall meet the requirements inACI 318, excluding Chapters 21.

11.3. Beams



Connections shall be designed for the applied factored load combinations and theirDesign Strength shall meet the requirements in Section 7.

12. COMPOSITE ORDINARY BRACED FRAMES (C-OBF)

12.1. Scope

This Section is applicable to concentrically and Eccentrically Braced Frame sys-tems that consist of either composite or reinforced concrete columns, structuralsteel or Composite Beams, and structural steel or Composite Braces. C-OBF shallbe designed assuming that under the Design Earthquake limited inelastic actionwill occur in the beams, columns, braces, and/or connections. C-OBF shall meetthe requirements of this section.

12.2. Columns

Reinforced-Concrete-Encased Composite Columns shall meet the requirementsfor ordinary seismic systems in Sections 6.4. Concrete-Filled Composite Columns




Sect. 13.] PART II – COMPOSITE CONCENTRICALLY BRACED FRAMES (C-CBF) 61

shall meet the requirements in Section 6.5. Reinforced concrete columns shallmeet the requirements in ACI 318 excluding Chapter 21.

12.3. Beams


12.4. Braces

Structural steel braces shall meet the requirements in the LRFD Specification.Composite Braces shall meet the requirements for Composite Columns in Sec-tion 12.2.

12.5. Connections

Connections shall be designed for the applied load combinations stipulated by theApplicable Building Code and their Design Strength shall meet the requirementsin Section 7.

13. COMPOSITE CONCENTRICALLY BRACED FRAMES (C-CBF)

13.1. Scope

This Section is applicable to braced systems that consist of concentrically con-nected members. Minor eccentricities are permitted if they are accounted for inthe design. Columns shall be either composite structural steel or reinforced con-crete. Beams and braces shall be either structural steel or composite structuralsteel. C-CBF shall be designed so that under the loading of the Design Earthquakeinelastic action will occur primarily through tension yielding and/or buckling ofbraces. C-CBF shall meet the requirements of this section.

13.2. Columns

Structural steel columns shall meet the requirements in Part I Section 8. Compositestructural steel columns shall meet the requirements for special systems in Section6.4 or 6.5. Reinforced concrete columns shall meet the requirements for structuraltruss elements in ACI 318 Chapter 21.

13.3. Beams


13.4. Braces

Structural steel braces shall meet the requirements for SCBF in Part I Section 13.Composite Braces shall meet the requirements for Composite Columns in Sec-tion 13.2.




62 PART II – COMPOSITE ECCENTRICALLY BRACED FRAMES (C-EBF) [Sect. 14.

13.5. Bracing Connections

Bracing connections shall meet the requirements in Section 7 and Part I Section 13.

14. COMPOSITE ECCENTRICALLY BRACED FRAMES (C-EBF)

14.1. Scope

This Section is applicable to braced systems for which one end of each braceintersects a beam at an eccentricity from the intersection of the centerlines of thebeam and column or intersects a beam at an eccentricity from the intersectionof the centerlines of the beam and an adjacent brace. C-EBF shall be designedso that inelastic deformations will occur only as shear yielding in the Links. Thediagonal braces, columns, and beam segments outside of the Link shall be designedto remain essentially elastic under the maximum forces that can be generated bythe fully yielded and strain-hardened Link. Columns shall be either compositeor reinforced concrete. Braces shall be structural steel. Links shall be structuralsteel as described in this Section. The Design Strength of members shall meetthe requirements in the LRFD Specification, except as modified in this Section.C-EBF shall meet the requirements in Part I Section 15, except as modified in thisSection.

14.2. Columns

Reinforced concrete columns shall meet the requirements for structural truss ele-ments in ACI 318 Chapter 21. Composite Columns shall meet the requirementsfor special seismic systems in Sections 6.4 or 6.5. Additionally, where a Link isadjacent to a reinforced concrete column or reinforced-concrete-encased column,transverse reinforcement meeting the requirements in ACI 318 Section 21.4.4 (orSection 6.4c(6)(a) for Composite Columns) shall be provided above and below theLink connection.

All columns shall meet the requirements in Part I Section 15.8.

14.3. Links

Links shall be unencased structural steel and shall meet the requirement for EBFLinks in Part I Section 15. It is permitted to encase the portion of the beam outsideof the Link with reinforced concrete. Beams containing the Link are permitted toact compositely with the floor slab using shear connectors along all or any portionof the beam if the composite action is considered when determining the NominalStrength of the Link.

14.4. Braces

Structural steel braces shall meet the requirements for EBF in Part I Section 15.

14.5. Connections

In addition to the requirements for EBF in Part I Section 15, connections shallmeet the requirements in Section 7.




Sect. 15.] PART II – ORDINARY REINFORCED CONCRETE SHEAR WALLS COMPOSITE 63

15. ORDINARY REINFORCED CONCRETE SHEAR WALLS COMPOSITEWITH STRUCTURAL STEEL ELEMENTS (C-ORCW)

15.1. Scope

The requirements in this Section apply when reinforced concrete walls are com-posite with structural steel elements, either as infill panels, such as reinforcedconcrete walls in structural steel frames with unencased or reinforced-concrete-encased structural steel sections that act as Boundary Members, or as structuralsteel Coupling Beams that connect two adjacent reinforced concrete walls. Rein-forced concrete walls shall meet the requirements in ACI 318 excluding Chapter 21.C-ORCW shall meet the requirements of this section.

15.2. Boundary Members

When unencased structural steel sections function as Boundary Members in rein-forced concrete infill panels, the structural steel sections shall meet the require-ments in the LRFD Specification. The required axial strength of the BoundaryMember shall be determined assuming that the shear forces are carried by the re-inforced concrete wall and the entire gravity and overturning forces are carried bythe Boundary Members in conjunction with the shear wall. The reinforced concretewall shall meet the requirements in ACI 318 excluding Chapter 21.

When fully reinforced-concrete-encased structural steel sections function as Boun-dary Members in reinforced concrete infill panels, the analysis shall be based upon atransformed concrete section using elastic material properties. The wall shall meetthe requirements in ACI 318 excluding Chapter 21. When the reinforced-concrete-encased structural steel Boundary Member qualifies as a Composite Column asdefined in LRFD Specification Chapter I, it shall be designed to meet the ordinaryseismic system requirements in Section 6.4. Otherwise, it shall be designed as aComposite Column to meet the requirements in ACI 318.

Headed shear studs or welded reinforcement anchors shall be provided to transfervertical shear forces between the structural steel and reinforced concrete. Headedshear studs, if used, shall meet the requirements in LRFD Specification Chapter I.Welded reinforcement anchors, if used, shall meet the requirements in AWS D1.4.

15.3. Coupling Beams

Structural steel Coupling Beams that are used between two adjacent reinforcedconcrete walls shall meet the requirements in the LRFD Specification and thisSection:

Coupling Beams shall have an embedment length into the reinforced concretewall that is sufficient to develop the maximum possible combination of momentand shear that can be generated by the nominal bending and shear strength of theCoupling Beam. The embedment length shall be considered to begin inside thefirst layer of confining reinforcement in the wall Boundary Member. Connectionstrength for the transfer of loads between the Coupling Beam and the wall shallmeet the requirements in Section 7.

Vertical wall reinforcement with design axial strength equal to the nominal shearstrength of the Coupling Beam shall be placed over the embedment length of the




64 PART II – SPECIAL REINFORCED CONCRETE SHEAR WALLS COMPOSITE [Sect. 16.

beam with two-thirds of the steel located over the first half of the embedment length.This wall reinforcement shall extend a distance of at least one tension developmentlength above and below the flanges of the Coupling Beam. It is permitted to usevertical reinforcement placed for other purposes, such as for vertical BoundaryMembers, as part of the required vertical reinforcement.

16. SPECIAL REINFORCED CONCRETE SHEAR WALLS COMPOSITEWITH STRUCTURAL STEEL ELEMENTS (C-SRCW)

16.1. Scope

C-SRCW systems shall meet the requirements in Section 15 for C-ORCW and theshear-wall requirement in ACI 318 including Chapter 21, except as modified inthis Section.


In addition to the requirements in Section 15.2a, unencased structural steel columnsshall meet the requirements in Part I Sections 5, 6 and 8.

Walls with reinforced-concrete-encased structural steel Boundary Members shallmeet the requirements in Section 15.2 as wells as the requirements in this Section.The wall shall meet the requirements in ACI 318 including Chapter 21. Reinforced-concrete-encased structural steel Boundary Members that qualify as CompositeColumns in LRFD Specification Chapter I shall meet the special seismic systemrequirements in Section 6.4. Otherwise, such members shall be designed as com-posite compression members to meet the requirements in ACI 318 including thespecial seismic requirements for Boundary Members in Chapter 21. Transversereinforcement for confinement of the composite Boundary Member shall extend adistance of 2h into the wall where h is the overall depth of the Boundary Memberin the plane of the wall.

Headed shear studs or welded reinforcing bar anchors shall be provided as specifiedin Section 15.2. For connection to unencased structural steel sections, the NominalStrength of welded reinforcing bar anchors shall be reduced by 25 percent fromtheir Static Yield Strength.

16.3. Coupling Beams

In addition to the requirements in Section 15.3, structural steel Coupling Beamsshall meet the requirements in Part I Sections 15.2 and 15.3. When required inPart I Section 15.3, the coupling rotation shall be assumed as 0.08 radians unlessa smaller value is justified by rational analysis of the inelastic deformations thatare expected under the Design Earthquake. Face Bearing Plates shall be providedon both sides of the Coupling Beams at the face of the reinforced concrete wall.These stiffeners shall meet the detailing requirements in Part I Section 15.3.

Vertical wall reinforcement as specified in Section 15.3 shall be confined bytransverse reinforcement that meets the requirements for Boundary Members inACI 318 Section 21.7.2.




Sect. 17.] PART II – COMPOSITE STEEL PLATE SHEAR WALLS (C-SPW) 65

17. COMPOSITE STEEL PLATE SHEAR WALLS (C-SPW)

17.1. Scope

This Section is applicable to structural walls consisting of steel plates with rein-forced concrete encasement on one or both sides of the plate and structural steel orcomposite Boundary Members. C-SPW shall meet the requirements of this section.

17.2. Wall Elements

17.2a. Nominal Shear Strength

The nominal shear strength of C-SPW with a stiffened plate conforming to Sec-tion 17.2b shall be determined as:

Vns = 0.6Asp Fy (17-1)

where

Vns = nominal shear strength of the steel plate, kips (N)Asp = horizontal area of stiffened steel plate, in.2 (mm2)Fy = specified minimum yield strength of the plate, ksi (MPa)

The nominal shear strength of C-SPW with a plate that does not meet the stiffen-ing requirements in Section 17.2b shall be based upon the strength of the plate,excluding the strength of the reinforced concrete, and meet the requirements in theLRFD Specification,including the effects of buckling of the plate.

17.2b. Detailing Requirements

The steel plate shall be adequately stiffened by encasement or attachment to thereinforced concrete if it can be demonstrated with an elastic plate buckling analysisthat the composite wall can resist a nominal shear force equal to Vns . The concretethickness shall be a minimum of 4 in. (102 mm) on each side when concrete isprovided on both sides of the steel plate and 8 in. (203 mm) when concrete isprovided on one side of the steel plate. Headed shear stud connectors or othermechanical connectors shall be provided to prevent local buckling and separationof the plate and reinforced concrete. Horizontal and vertical reinforcement shallbe provided in the concrete encasement to meet the detailing requirements inACI 318 Section 14.3. The reinforcement ratio in both directions shall not be lessthan 0.0025; the maximum spacing between bars shall not exceed 18 in. (457 mm).

The steel plate shall be continuously connected on all edges to structural steelframing and Boundary Members with welds and/or slip-critical high-strength boltsto develop the nominal shear strength of the plate. The Design Strength of weldedand bolted connectors shall meet the additional requirements in Part I Section 7.


Structural steel and composite Boundary Members shall be designed to meet therequirements in Section 16.2.

Boundary Members shall be provided around openings as required by analysis.




66



PB262A-03 PB262/Seismic.cls December 18, 2002 15:51 Char Count=

67PART III. ALLOWABLE STRESSDESIGN (ASD) ALTERNATIVE

As an alternative to the Load and Resistance Factor Design (LRFD) provisions for struc-tural steel design in Part I, the use of the Allowable Stress Design (ASD) provisionsin this Part is permitted. All requirements of Part I shall be met except as modified orsupplemented in this Part. When using this Part, the terms “LRFD Specification”,“FR” and “PR” in Part I shall be taken as “ASD Specification”, “Type 1”, and “Type 3”,respectively.

1. SCOPE

Substitute the following for PART I Section 1 in its entirety:

These Provisions are intended for the design and construction of structural steelmembers and connections in the Seismic Force Resisting Systems in buildings forwhich the design forces resulting from earthquake motions have been determinedon the basis of various levels of energy dissipation in the inelastic range of response.These Provisions shall apply to buildings that are classified in the ApplicableBuilding Code as Seismic Design Category D (or equivalent) and higher or whenrequired by the Engineer of Record.

These Provisions shall be applied in conjunction with the AISC Specification forStructural Steel Buildings—Allowable Stress Design and Plastic Design includingSupplement No. 1, hereinafter referred to as the ASD Specification. All membersand connections in the Seismic Force Resisting System shall be proportioned asrequired in the ASD Specification to resist the applicable load combinations andshall meet the requirements in these Provisions.

Part III includes the Part I Glossary and Appendix S.


Substitute the following for the first two paragraphs of Part I Section 2:

The documents referenced in these Provisions shall include those listed in ASDSpecification Section A6 with the following additions and modifications:

American Institute of Steel Construction

Specification for Structural Steel Buildings—Allowable Stress Design and PlasticDesign, June 1, 1989 including Supplement No. 1, December 17, 2001

Substitute the following for the last paragraph of Part I Section 2:

Research Council on Structural Connections

Specification for Structural Joints Using ASTM A325 or A490 Bolts, June 23, 2000,Appendix B


Substitute the following for Part I Section 4.2 in its entirety:




68 PART III – LOADS, LOAD COMBINATIONS, AND NOMINAL STRENGTHS [Sect. 4.

TABLE III-4-1Resistance Factors for ASDLimit State Resistance Factor

TensionYielding 0.90Rupture 0.75

Compression buckling 0.85Flexure

Yielding 0.90Rupture 0.75

ShearYielding 0.90Rupture 0.75

TorsionYielding 0.90Buckling 0.90

Complete-joint-penetration groove weldsTension or compression normal to effective area Base metal 0.90

Weld metal 0.90Shear on effective area Base metal 0.90

Weld metal 0.80Partial-joint-penetration groove welds

Compression normal to effective area Base metal 0.90Weld metal 0.90

Tension normal to effective area Base metal 0.90Weld metal 0.80

Shear parallel to axis of weld Weld metal 0.75Fillet welds

Shear on effective area Weld metal 0.75Plug or slot welds

Shear parallel to faying surface (on effective area) Weld metal 0.75Bolts

Tension rupture, shear rupture, combined tension and shear 0.75Slip resistance for bolts in standard holes,

oversized holes, and short-slotted holes 1.0Slip resistance for bolts in long-slotted holes

with the slot perpendicular to the direction of the slot 1.0Slip resistance for bolts in long-slotted holes with

the slot parallel to the direction of the slot 0.85Connecting elements

Tension yielding, shear yielding 0.90Bearing strength at bolt holes, tension rupture,

shear rupture, block shear rupture 0.75Contact bearing Bearing on steel 0.75

Bearing on concrete 0.60Flanges and webs with concentrated forces

Local flange bending, compression buckling of web 0.90Local web yielding 1.0Web crippling, Panel Zone web shear 0.75Sidesway web buckling 0.85

4.2. Nominal Strength

The Nominal Strengths of members and connections shall be determined asfollows:

Replace ASD Specification Section A5.2 with the following: “The NominalStrength of structural steel members and connections for resisting seismic forces




Sect. 9.] PART III – SPECIAL MOMENT FRAMES 69

acting alone or in combination with dead and live loads shall be determined bymultiplying 1.7 times the allowable stresses in Section D, E, F, G, H, J, and K.

Amend the first paragraph of ASD Specification Section N1 by deleting “or earth-quake” and adding: “The Nominal Strength of members and connections shallbe determined by the requirements contained herein. Except as modified in theseprovisions, all pertinent requirements of Chapters A through M shall govern.”

In ASD Specification Section H1 the definition of F ′e shall read as follows:

F ′e = 2 Es

(Klb/rb)2(4-1)

where:

lb = the actual length in the plane of bending, in. (mm)rb = the corresponding radius of gyration, in. (mm)K = the effective length factor in the plane of bending

4.3. Design Strength

The Design Strength of structural steel members and connections subjected toseismic forces in combination with other prescribed loads shall be determinedby converting allowable stresses into Nominal Strengths and multiplying suchNominal Strengths by the Resistance Factors given in Table III-4-1.

7. CONNECTIONS, JOINTS, AND FASTENERS

7.2. Bolted Joints

Substitute the following for Part I Section 7.2 fourth paragraph in its entirety:

The design resistance to shear and combined tension and shear of bolted jointsshall be determined in accordance with the ASD Specification Sections J3.5 andJ3.7, except that the allowable bearing stress at bolt holes Fp shall not be takengreater than 1.2Fu .

9. SPECIAL MOMENT FRAMES

9.3. Panel Zone of Beam-to-Column Connections(beam web parallel to column web)

Substitute the following for Part I Section 9.3a in its entirety:

The required thickness of the panel zone shall be determined in accordance withthe method used in proportioning the panel zone of the tested connection. As aminimum, the required shear strength Ru of the panel zone shall be determined fromthe summation of the moments at the column faces as determined by projectingthe expected moments at the plastic hinge points to the column faces. The designshear strength v Rv of the panel zone shall be determined using v = 1.0.




70 PART III – SPECIAL TRUSS MOMENT FRAMES [Sect. 12.

When Pu ≤ 0.75Py ,

Rv = 0.6Fydctp

[1 + 3bcf t2

c f

dbdctp

](9-1)

When Pu > 0.75Py ,

Rv = 0.6Fydctp

[1 + 3bcf t2

c f

dbdctp

] [1.9 − 1.2Pu

Py

]. (9-1a)

where:

tp = total thickness of Panel Zone including doubler plate(s), in. (mm)dc = overall column depth, in. (mm)

bcf = width of the column flange, in. (mm)tc f = thickness of the column flange, in. (mm)db = overall beam depth, in. (mm)Fy = specified minimum yield strength of the Panel Zone steel, ksi (MPa)


Substitute the following for Part I Section 9.7b(1) in its entirety:

The required column strength shall be determined from the ASD load combinationsstipulated in the Applicable Building Code, except that E shall be taken as the lesserof:

(a) The Amplified Seismic Load

(b) 125 percent of the frame Design Strength based upon either the beam designflexural strength or Panel Zone design shear strength

12. SPECIAL TRUSS MOMENT FRAMES

12.4. Nominal Strength of Non-special Segment Members

Substitute the following for the first sentence in Part I Section 12.4:

Members and connections of STMF, except those in the special segment definedin Section 12.2, shall have a Design Strength to resist ASD load combinations asstipulated by the Applicable Building Code replacing the earthquake load termE with the lateral loads necessary to develop the expected vertical nominal shearstrength in the special segment Vne given as: [balance to remain unchanged]

12.6. Lateral Bracing

Substitute the following for the first sentence in Part I Section 12.6:

The top and bottom chords of the trusses shall be laterally braced at the ends of thespecial segment, and at intervals not to exceed Lc according to ASD SpecificationSection F1, along the entire length of the truss.




Sect. 14.] PART III – ORDINARY CONCENTRICALLY BRACED FRAMES (OCBF) 71

13. SPECIAL CONCENTRICALLY BRACED FRAMES (SCBF)

Substitute the following for Part I Section 13.4a(2) in its entirety:

(2) A beam that is intersected by braces shall be designed to support the effectsof all tributary dead and live loads assuming that the bracing is not present.

Substitute the following for Part I Section 13.4a(3) in its entirety:

(3) A beam that is intersected by braces shall be designed to resist the effectsof ASD load combinations as stipulated by the Applicable Building Code,except that a load Qb shall be substituted for the term E . Qb is the maximumunbalanced vertical load effect applied to the beam by the braces. This loadeffect shall be calculated using a minimum of Py for the brace in tension anda maximum of 0.3 times c Pn for the brace in compression.

14. ORDINARY CONCENTRICALLY BRACED FRAMES (OCBF)

Substitute the following for Part I Section 14.2 in its entirety:

14.2. Strength

The Required Strength of the members and connections, other than brace connec-tions, in OCBF shall be determined using the ASD load combinations stipulatedby the Applicable Building Code except E shall be taken as the Amplified SeismicLoad. The Design Strength of brace connections shall equal or exceed the expectedtensile strength of the brace, determined as Ry Fy Ag . Braces with Kl/r greater than4.23

√Es/Fy shall not be used in V or inverted-V configurations.




72




73

COMMENTARYon the Seismic Provisionsfor Structural Steel Buildings

May 21, 2002




74




75PART I. STRUCTURAL STEEL BUILDINGS

Experience from the 1994 Northridge and 1995 Kobe earthquakes significantly expandedknowledge regarding the seismic response of structural steel building systems, particu-larly welded steel Moment Frames. (Note: glossary terms are capitalized throughout thisspecification and commentary.) Shortly after the Northridge earthquake, the SAC JointVenture1 initiated a comprehensive study of the seismic performance of steel MomentFrames. Funded by the Federal Emergency Management Agency (FEMA), SAC devel-oped guidelines for structural engineers, building officials and other interested parties forthe evaluation, repair, modification and design of welded steel Moment Frame structuresin seismic regions. AISC actively participated in SAC activities.

Many recommendations in the Recommended Seismic Design Criteria for New SteelMoment-Frame Buildings – FEMA 350 (FEMA, 2000a) formed the basis for SupplementNo. 2 to the 1997 AISC Seismic Provisions for Structural Steel Buildings (AISC, 2000).Supplement No. 2 to the 1997 Provisions was developed simultaneously and cooperativelywith the revisions to the Building Seismic Safety Council (BSSC) NEHRP Provisions.Accordingly, Supplement No. 2 formed the basis for steel seismic design provisions inthe 2000 NEHRP Provisions (FEMA, 2000g) as well as those in the 2000 InternationalBuilding Code (IBC) 2002 Supplement, which has been published by the InternationalCode Council (ICC, 2002).

These 2002 AISC Seismic Provisions (hereinafter referred to as the Provisions) continueincorporating the recommendations of FEMA 350 and other research. While research isongoing, the Committee has prepared this revision of the Provisions using the best availableknowledge to date. Although timing did not permit the adoption of these Provisions inASCE 7 (ASCE, 2002) it is intended that these provisions be used in conjunction with the2002 edition of ASCE 7. It is also anticipated that these Provisions will form the basisfor structural steel seismic requirements in two model building code editions currentlyunder development: the International Building Code and the National Fire ProtectionAssociation (NFPA) Building Code dated 2003. NFPA will reference ASCE 7 for seismicloading.

C1. SCOPE

Structural steel building systems in seismic regions are generally expected to dis-sipate seismic input energy through controlled inelastic deformations of the struc-ture. These Provisions supplement the AISC LRFD Specification (AISC, 1999)for such applications. The seismic design loads specified in the building codeshave been developed considering the energy dissipation generated during inelasticresponse.

It should be noted that these provisions were developed specifically for buildings.The Provisions, therefore, may not be applicable, in whole or in part, to non-building structures. Extrapolation of their use to non-building structures shouldbe done with due consideration of the inherent differences between the responsecharacteristics of buildings and non-building structures.

1A joint venture of the Structural Engineers Association of California (SEAOC), Applied Technology (ATC), and CaliforniaUniversities for Research in Earthquake Engineering (CUREe).




76 PART I – GENERAL SEISMIC DESIGN REQUIREMENTS [Comm. C3.

The Provisions are intended to be mandatory for buildings in Seismic DesignCategory D and above, as defined in the National Earthquake Hazard ReductionProgram (NEHRP) Seismic Provisions (FEMA, 2000e) and ASCE 7 (ASCE,2002). For buildings in Seismic Design Category A to C the designer is givena choice to either solely use the AISC LRFD Specification (AISC, 1999) and the Rfactor given for structural steel buildings not specifically detailed for seismic resis-tance (typically 3) or the designer may choose to assign a higher R factor to a systemdetailed for seismic resistance and follow the requirements of these Provisions.

C2. REFERENCED SPECIFICATIONS, CODES, AND STANDARDS

The specifications, codes and standards referenced in Part I are listed with theappropriate revision date that was used in the development of Part I. Where thesedocuments are referenced in the Provisions, the versions given in Section 2 apply.While most of these documents are also referenced in the LRFD Specification,some have been revised since its publication in 1999.

C3. GENERAL SEISMIC DESIGN REQUIREMENTS

When designing buildings to resist earthquake motions, each building is catego-rized based upon its occupancy and use to establish the potential earthquake hazardthat it represents. Determining the required Design Strength differs significantlyin each specification or building code. The primary purpose of these Provisions isto provide information necessary to determine the Design Strength of steel build-ings. The following discussion provides a basic overview of how several seismiccodes or specifications categorize building structures and how they determine theRequired Strength and stiffness. For the variables required to assign Seismic De-sign Categories, limitations of height, vertical and horizontal irregularities, sitecharacteristics, etc., the Applicable Building Code should be consulted.

In the 2000 NEHRP Provisions (FEMA, 2000g), buildings are assigned to one ofthree Seismic Use Groups, depending upon occupancy or use. Group III includesessential facilities, while Groups I and II include facilities associated with a lesserdegree of public hazard. Buildings are then assigned to a Seismic Design Categorybased upon the Seismic Use Group, the seismicity of the site and the period ofthe building. Seismic Design Categories A, B and C are generally applicable tobuildings in areas of low to moderate seismicity and special seismic provisions likethose in these Provisions are not mandatory unless the Engineer of Record choosesto use an R factor of one of the defined systems prescribed in these Provisions.However, special seismic provisions are mandatory in Seismic Design CategoriesD, E and F, including consideration of system redundancy. Seismic Design Cate-gory D is generally applicable to buildings in areas of high seismicity and SeismicUse Group III buildings in areas of moderate seismicity. Seismic Design Cate-gories E and F are generally applicable to buildings in Seismic Use Groups I and IIand Seismic Use Group III, respectively, in areas of especially high seismicity.

In ASCE 7 (ASCE, 2002), buildings are assigned to one of four Occupancy Cat-egories. Category IV, for example, includes essential facilities. Buildings are thenassigned to a Seismic Use Group based upon the Occupancy Categories andthe seismicity of the site. Seismic Design Categories A, B and C are generallyapplicable to buildings in areas of low to moderate seismicity and special seismic




Comm. C4.] PART I – LOADS, LOAD COMBINATIONS, AND NOMINAL STRENGTH 77

provisions like those in these Provisions are not mandatory. However, special seis-mic provisions are mandatory in Seismic Design Categories D and E, which coverareas of high seismicity.

In the 1997 Uniform Building Code (ICBO, 1997) and the 1999 SEAOC SeismicProvisions (SEAOC, 1999), the detailing required for buildings is based on theSeismic Zone in which the building is located.

A new term “Applicable Building Code” has been introduced into this editionof the Seismic Provisions. While it is the intent that these Provisions be usedin conjunction with the codes and standards previously listed, there can be noguarantee which building code edition a designer may use to design a steel building.To eliminate potential conflicts with the many building codes currently in use,these Provisions refer to the Applicable Building Code to establish loads and loadcombinations, system limitations and system factors (e.g. R, Cd and o). However,where building codes differ from ASCE 7, it is the intent of these Provisions thatthe ASCE 7 criteria apply.

C4. LOADS, LOAD COMBINATIONS, AND NOMINAL STRENGTH

These Provisions are intended for use with load combinations given in the Ap-plicable Building Code. However, since they are written for consistency with theload combinations given in ASCE 7 (ASCE, 2002) and IBC 2000 (ICC, 2000),consistency with the Applicable Building Code should be confirmed.

The earthquake load E is the combination of the horizontal seismic load effect andan approximation of the effect due to the vertical accelerations that accompany thehorizontal earthquake effects.

An amplification or overstrength factor o applied to the horizontal portion of theearthquake load E is prescribed in ASCE 7, the 2000 IBC, 2000 NEHRP Provisionsand the 1997 Uniform Building Code; however, the relevant load combinationsare not all expressed in exactly the same format, as shown in Table C-I-4.1. Inprior editions of these Provisions it was felt that this difference could be clarifiedby including load combinations (Equations 4-1 and 4-2 in the 1997 Seismic Pro-visions (AISC, 1997b)), primarily to account for overstrength inherent in differentsystems or elements when determining the Required Strength of connections. Un-fortunately, due to the difference in the various codes and source documents withwhich these Provisions are intended to be used, the specification of Load Combi-nations 4-1 and 4-2 proved confusing. It is not practical to specifically referenceany load combinations from reference documents, and Load Combinations 4-1and 4-2 were eliminated in favor of a new term “Amplified Seismic Load.” Whenused in these Provisions, this term is intended to refer to the appropriate loadcombinations in the Applicable Building Code that account for overstrength ofmembers of the Seismic Load Resisting System. The load combinations contain-ing the overstrength factor o should be used where these Provisions require useof the Amplified Seismic Load.

The general relationship between the different structural steel systems is illustratedin Table C-I-4.2 based upon similar information in the ASCE 7 load standard. Ris a seismic load reduction factor used to approximate the inherent ductility of theSeismic Load Resisting System. Cd is an amplification factor that is used with the




78 PART I – LOADS, LOAD COMBINATIONS, AND NOMINAL STRENGTH [Comm. C4.

TABLE C-I-4.1Load Combinations

Gravity Effects Additive

Basic CombinationsCode or Ref. including E Definition of E Overstrength

ASCE 7-2002 1.2D + 1.0E + 0.5L + 0.2S E = QE + 0.2SDS D E = o QE + 0.2SDS D

NEHRP 2000 1.2D + 1.0E + 0.5L + 0.2S E = QE + 0.2SDS D E = o QE + 0.2SDS D

IBC 2000 1.2D + 1.0E + f1 L + f2 S E = QE + 0.2SDS D Em = o QE + 0.2SDS D

UBC 1997 1.2D + 1.0E + f1 L + f2 S E = Eh + Ev Em = o Eh

where Ev = 0.5CaID

Gravity Effects Counteraction

Basic CombinationsCode or Ref. including E Definition of E Overstrength

ASCE 7-2002 0.9D + 1.0E + 1.6H E = QE − 0.2SDSD E = o QE − 0.2SDSD

NEHRP 2000 0.9D + 1.0E + 1.6H E = QE − 0.2SDSD E = o QE − 0.2SDSD

IBC 2000 0.9D + 1.0E E = QE − 0.2SDSD Em = o QE − 0.2SDSD

UBC 1997 0.9D + 1.0E E = Eh + Ev Em = o Eh

where Ev = 0.5CaID

Note: For definitions, see applicable code.

TABLE C-I-4.2DESIGN FACTORS FOR STRUCTURAL STEEL

SYSTEMSBASIC STRUCTURAL SYSTEM ANDSEISMIC LOAD RESISTING SYSTEM R Cd

Systems designed and detailed to meet the requirements in theLRFD Specification but not the requirements of Part I 3 3

Systems designed and detailed to meet the requirements of both the LRFD Specification and Part I:

Braced Frame Systems:Special Concentrically Braced Frames (SCBF) 6 5Ordinary Concentrically Braced Frames (OCBF) 5 41/2Eccentrically Braced Frames (EBF)

with moment connections at columns away from Link 8 4without moment connections at columns away from Link 7 4

Moment Frame Systems:Special Moment Frames (SMF) 8 51/2

Intermediate Moment Frames (IMF) 41/2 4Ordinary Moment Frames (OMF) 31/2 3Special Truss Moment Frames (STMF) 7 51/2

Dual Systems with SMF capable of resisting 25 percent of V:Special Concentrically Braced Frames (SCBF) 8 61/2Eccentrically Braced Frames (EBF)

with moment connections at columns away from Link 8 4without moment connections at columns away from Link 7 4

Dual Systems with IMF∗ capable of resisting 25 percent of V:Special Concentrically Braced Frames (SCBF) 41/2 4

∗OMF is permitted in lieu of IMF in Seismic Design Categories A, B and C.




Comm. C5.] PART I – STORY DRIFT 79

loads for strength design to calculate the seismic drift. The use of these factorsshould be consistent with those specified in the Applicable Building Code withdue consideration of the limitations and modifications necessary to account forbuilding category, building height, vertical or horizontal irregularities, and sitecharacteristics.

C5. STORY DRIFT

For non-seismic applications, story drift limits, like deflection limits, are commonlyused in design to assure the serviceability of the structure. They vary because theydepend upon the structural usage and contents. As an example, for wind loadssuch serviceability limit states are regarded as a matter of engineering judgmentrather than absolute design limits (Fisher and West, 1990) and no specific designrequirements are given in the LRFD Specification or these Provisions.

The situation is somewhat different when considering seismic effects. Researchhas shown that story drift limits, although primarily related to serviceability, alsoimprove frame stability (P-) and seismic performance because of the resultingadditional strength and stiffness. Although some building codes, load standardsand resource documents contain specific seismic drift limits, there are major dif-ferences among them as to how the limit is specified and applied. Nevertheless,drift control is important to both the serviceability and the stability of the structure.As a minimum, the designer should use the drift limits specified in the ApplicableBuilding Code.

The analytical model used to estimate building drift should accurately accountfor the stiffness of the frame elements and connections and other structural andnonstructural elements that materially affect the drift. Recent research on steelMoment Frame connections indicates that in most cases Panel Zone deformationshave little effect on analytical estimates of drift and need not be explicitly modeled(FEMA, 2000f). In cases where nonlinear element deformation demands are ofinterest, Panel Zone shear behavior should be represented in the analytical modelwhenever it significantly affects the state of deformation at a beam-to-columnconnection. Mathematical models for the behavior of the Panel Zone in terms ofshear force-shear distortion relationships have been proposed by many researchers.FEMA 355C presents a good discussion of how to incorporate panel zone defor-mations in to the analytical model (FEMA, 2000f).

Adjustment of connection stiffness is usually not required for connections tradi-tionally considered as fixed, although FEMA 350 (FEMA, 2000a) contains recom-mendations for adjusting calculated drift for frames with Reduced Beam Sections.Nonlinear models should contain nonlinear elements where plastic hinging is ex-pected to properly capture the inelastic deformation of the frame.

The story drift limits in ASCE 7 (ASCE, 2002) and the 2000 NEHRP Provisions(FEMA, 2000g) are to be compared to an amplified story drift that approximates thedifference in deflection between the top and bottom of the story under considerationduring a large earthquake. The amplified story drift is determined by multiplyingthe elastic drift caused by the horizontal component of the earthquake load E bya deflection amplification factor Cd , which is dependent upon the type of buildingsystem used; see Table C-I-4.2.




80 PART I – STORY DRIFT [Comm. C5.

The following discussion pertains primarily to Moment Frames (FEMA, 2000a),although other systems where high lateral drifts may occur require a similar anal-ysis. Each story of the structure should be investigated to ensure that lateral driftsinduced by earthquake response do not result in a condition of instability undergravity loads. The analysis of the structure should explicitly consider the geometricnonlinearity introduced by P- effects. The quantity i should be calculated foreach story for each direction of response, as follows:

i = Pi Ri

Vyi H(C5-1)

where:

H = height of story, which may be taken as the distance between the centerline offloor framing at each of the levels above and below, or the distance betweenthe top of floor slabs at each of the levels above and below, in. (mm)

Pi = portion of the total weight of the structure including dead, permanent live,and 25 percent of transient live loads acting on all of the columns withinstory level i , kips (N)

R = design factor used to determine the design seismic loads applicable to theStructural System as defined in the Applicable Building Code. In the absenceof such definition, R is listed in Table C-I-4.1.

i = calculated lateral drift at the center of rigidity of story i , when the designseismic loads are applied in the direction under consideration, in. (mm)

Vyi = total plastic lateral shear restoring capacity in the direction under consider-ation at story i , kips (N)

The plastic story shear quantity, Vyi , should be determined by methods of plasticanalysis. However, Vyi may be approximately calculated from the equation:

Vyi =2

n∑j=1

MpG j

H(C5-2)

when the following conditions apply:

(1) All beam-column connections meet the strong-column-weak-beam criterionin the story

(2) The same number of moment-resisting bays is present at the top and bottomof the frame and

(3) The strength of girders, moment-connected at both ends, at the top and bottomof the frame is similar

where:

MpG j = the plastic moment capacity of girder “j” participating in the moment-resisting framing at the floor level on top of the story, and

n = the number of moment-resisting girders in the framing at the floor levelon top of the story




Comm. C5.] PART I – STORY DRIFT 81

In any story in which all columns do not meet the strong-column-weak-beamcriterion, the plastic story shear quantity, Vyi may be calculated from the equation:

Vyi =2

m∑k=1

MpCk

H(C5-3)

where:

m = the number of columns in moment-resisting framing in the story underconsideration, and

MpCk = the plastic moment capacity of each column “k”, participating in themoment-resisting framing, considering the axial load present on thecolumn

For other conditions, the quantity Vyi shall be calculated by plastic mechanismanalysis, considering the vertical distribution of lateral loads on the structure.

The quantity i is the ratio of the effective story shear produced by first order P-effects at the calculated story drift to the maximum restoring force in the structure.When this ratio has a value greater than 1.0, the structure does not have enoughstrength to resist the P- induced shear forces and may collapse in a sideswaymechanism. If the ratio is less than 1.0, the restoring force in the structure exceedsthe story shear due to P- effects and unless additional displacement is inducedor lateral loads applied, the structure should not collapse. Given the uncertaintyassociated with predicting significance of P- effects, it is recommended that when i in a story exceeds 0.3, the structure be considered unstable, unless a detailedglobal stability capacity evaluation for the structure, considering P- effects, isconducted.

P- effects can have a significant impact on the ability of structures to resistcollapse when subjected to strong ground shaking. When the non-dimensionalquantity, i , calculated in accordance with Equation C5-1 significantly exceeds avalue of about 0.1, the instantaneous stiffness of the structure can be significantlydecreased, and can effectively become negative. If earthquake induced displace-ments are sufficiently large to create negative instantaneous stiffness, collapse islikely to occur.

Analyses reported in FEMA 355F (FEMA, 2000f) included direct consideration ofP- effects in determining the ability of regular, well configured frames designedto modern code provisions to resist P--induced instability and P--inducedcollapse. For regular, well-configured structures, if the value of is maintainedwithin the limits indicated in this section (i.e. 0.3 or less), P--induced instabilityis unlikely to occur. Values of greater than this limit suggest that instability dueto P- effects is possible. In such cases, the frame should be redesigned to providegreater resistance to P--induced instability unless explicit evaluation of theseeffects using the detailed Performance Evaluation methods outlined in AppendixA of FEMA 350 (FEMA, 2000a) are performed.

The evaluation approach for P- effects presented in this section appears similarto but differs substantially from that contained in FEMA 302 (FEMA, 1997a), and




82 PART I – MATERIALS [Comm. C6.

in use in the building codes for many years. The approach contained in FEMA302 and the building codes was an interim formulation. Research indicates thatthis interim approach was not meaningful. Some of this research included theexplicit evaluation of P- effects for buildings of varying heights, subjected tomany different types of ground motion and designed using different building codeprovisions. Using these and other parameters, several tens of thousands of nonlinearanalyses were run to investigate P- effects. Extensive additional discussion onthe issue of P- effects and their importance in the response of structures at largeinterstory drifts is contained in FEMA 355C (FEMA, 2000d).

C6. MATERIALS

C6.1. Material Specifications

The structural steels that are explicitly permitted for use in seismic design havebeen selected based upon their inelastic properties and weldability. In general, theymeet the following characteristics: (1) a ratio of yield stress to tensile stress notgreater than 0.85; (2) a pronounced stress-strain plateau at the yield stress; (3) alarge inelastic strain capability (for example, tensile elongation of 20 percent orgreater in a 2-in. (50 mm) gage length); and (4) good weldability. Other steelsshould not be used without evidence that the above criteria are met.

The 50 ksi (345 MPa) limitation on the specified minimum yield stress for membersexpecting inelastic action refers to inelastic action under the effects of the DesignEarthquake. Modern steels of higher strength, such as A913 Grade 65, are generallyconsidered to have properties acceptable for seismic column applications.

C6.2. Material Properties for Determination of Required Strength

Brittle fracture of beam-to-column moment connections in the Northridge Earth-quake resulted from a complex combination of variables. One of the many con-tributing factors was the failure to recognize that actual beam yield stresses aregenerally higher than the specified minimum yield stress Fy , which elevates theconnection demand. In 1994, the Structural Shape Producers Council (SSPC)conducted a survey to determine the characteristics of current structural steel pro-duction (SSPC, 1994). FEMA (1995) subsequently recommended that the meanvalues of Fy from the SSPC study be used in calculations of demand on momentconnections. It was also recognized that the same overstrength concerns also applyto other systems as well.

Ry is defined as the ratio of Expected Yield Strength Fye to specified minimumyield stress Fy . It is used as a multiplier on the specified minimum yield stresswhen calculating the Required Strength of connections and other members thatmust withstand the development of inelasticity in another member. The specifiedvalues of Ry for rolled shapes are somewhat lower than those that can be calculatedusing the mean values reported in the SSPC survey. Those values were skewedsomewhat by the inclusion of a large number of smaller members, which typicallyhave a higher measured yield stress than the larger members common in seismicdesign. The given values are considered to be reasonable averages, although itis recognized that they are not maxima. The Expected Yield Strength Ry Fy canbe determined by testing conducted in accordance with the requirements for thespecified grade of steel. Such an approach should only be followed in unusual




Comm. C6.] PART I – MATERIALS 83

cases where there is extensive evidence that the values of Ry are significantlyunconservative. It is not expected that this would be the approach followed fortypical building projects. Refer to ASTM A370.

The higher values of Ry for ASTM A36/A36M (Ry = 1.5) and ASTM A572/A572M Grade 42 (290) (Ry = 1.3) shapes are indicative of the most recentlyreported properties of these grades of steel. If the material being used in designwas produced several years ago, it may be possible to use a reduced value of Rybased upon testing of the steel to be used or other supporting data (Galambos andRavindra, 1978). The values of Ry will be periodically monitored to ensure thatcurrent production practice is properly reflected.

A survey of HSS and steel pipe production data in 2000 resulted in Ry values of1.3 and 1.4, respectively.

While ASTM A709/A709M is primarily used in the design and construction ofbridges, it could also be used in building construction. Written as an umbrella spec-ification, its grades are essentially the equivalent of other approved ASTM spec-ifications. For example, ASTM A709/A709 Grade 50 (345) is essentially ASTMA572/A572M Grade 50 (345) and ASTM A709/A709M Grade 50W (345W) is es-sentially ASTM A588/A588M Grade 50 (345). Thus, if used, ASTM A709/A709Mmaterial should be treated as would the corresponding approved ASTM materialgrade.

Specific provisions for some Seismic Load Resisting Systems stipulate that the Re-quired Strength be determined by multiplying the Nominal Strength of a certainmember or connecting element by the value of Ry for the corresponding mate-rial grade. This overstrength is primarily of interest when evaluating the DesignStrength of another connecting element or member. It is not of interest, however,when evaluating the Design Strength of the same member to which the value ofRy was applied in the determination of the Required Strength. Therefore, whenboth the Required Strength and Design Strength calculations are made for thesame member or connecting element, it is also permitted to apply Ry in the de-termination of the Design Strength. An example of such a condition would be thechecking of the Required Strength of the beam outside of the Link in an EBF (seeSection 15.6). Since the Required Strength is generated in the same member (inthis case the Link portion of the beam), the Design Strength should also includethe Ry term.

C6.3. Notch-Toughness Requirements

The LRFD Specification requirements for notch toughness cover Group 4 and5 shapes and plate elements with thickness that is greater than or equal to 2 in.(50 mm) in tension applications. In these Provisions, this requirement is extended tocover: (1) all Group 4 and 5 shapes that are part of the Seismic Load Resisting Sys-tem; (2) ASTM Group 3 shapes that are part of the Seismic Load Resisting Systemwith flange thickness greater than or equal to 11/2 in. (38 mm); and, (3) plate ele-ments with thickness greater than or equal to 2 in. (50 mm) that are part of the Seis-mic Load Resisting System, such as the flanges of built-up girders. Because othershapes and plates are generally subjected to sufficient cross-sectional reductionduring the rolling process such that the resulting notch toughness will exceed thatrequired above (Cattan, 1995), specific requirements have not been included herein.




84 PART I – MATERIALS [Comm. C6.

The requirements of this section may not be necessary for members that resist onlyincidental loads. For example, a designer might include a member in the SeismicLoad Resisting System to develop a more robust load path, but the member willexperience only an insignificant level of seismic demand. An example of sucha member might include a transfer girder with thick plates where its design isdominated by its gravity load demand. It would be inconsistent with the intent ofthis section if the designer were to arbitrarily exclude a member with insignifi-cant seismic loads from the Seismic Load Resisting System that would otherwiseimprove the seismic performance of the building in order to avoid the toughnessrequirements in this section. The LRFD Specification requirements noted abovewould still apply in this case.

For rotary-straightened W-shapes, an area of reduced notch toughness has beendocumented in a limited region of the web immediately adjacent to the flange asillustrated in Figure C-I-6.1. Recommendations issued (AISC, 1997a) by AISCwere followed up by a series of industry sponsored research projects (Kaufmann,Metrovich and Pense, 2001; Uang and Chi, 2001; Kaufmann and Fisher, 2001;Lee, Cotton, Dexter, Hajjar, Ye, and Ojard, 2001; Bartlett, Jelinek, Schmidt, Dexter,Graeser, and Galambos, 2001). This research generally corroborates AISC’s initialfindings and recommendations.

Early investigations of connection fractures in the 1994 Northridge earthquakeidentified a number of fractures that some speculated were the result of inadequatethrough-thickness strength of the column flange material. As a result, in the periodimmediately following the Northridge earthquake, a number of recommendationswere promulgated that suggested limiting the value of through-thickness stress de-mand on column flanges to ensure that through-thickness yielding did not initiatein the column flanges. This limit state often controlled the overall design of theseconnections. However, the actual cause for the fractures that were initially thoughtto be through-thickness failures of the column flange are now considered to beunrelated to this material property. Detailed fracture mechanics investigations con-ducted as part of the FEMA/SAC project confirm that damage initially identifiedas through thickness failures is likely to have occurred as a result of certain combi-nations of filler metal and base material strength and notch toughness, conditionsof stress in the connection, and the presence of critical flaws in the welded joint. Inaddition to the analytical studies, extensive through-thickness testing conductedspecifically to determine the susceptibility to through thickness failures of moderncolumn materials meeting ASTM A572, Gr. 50 and ASTM A913, Gr. 65 specifi-cations did not result in significant through-thickness fractures (FEMA, 2000h).

In addition, none of the more than 100 full scale tests on “post-Northridge” con-nection details have demonstrated any through-thickness column fractures. Thiscombined analytical and laboratory research clearly shows that due to the highrestraint inherent in welded beam flange to column flange joints, the through thick-ness yield and tensile strengths of the column material are significantly elevatedin the region of the connection. For the modern materials tested, these strengthssignificantly exceed those loads that can be delivered to the column by the beamflange. For this reason, no limits are suggested for the through-thickness strengthof the base material by the FEMA/SAC program or in these Provisions.

The preceding discussion assumes that no significant column flange laminations,inclusions or other discontinuities occur in regions adjacent to welded beam




Comm. C7.] PART I – CONNECTIONS, JOINTS, AND FASTENERS 85

Fig. C-I-6.1. “k-area.”

flange-to-column flange joints. Where column shapes with proportions equal tothose defined in this Section are used, the Engineer of Record should considerspecifying ultrasonic examination in these regions prior to welding. FEMA 353(FEMA, 2000b) provides guidance for specifying such examinations.

C7. CONNECTIONS, JOINTS, AND FASTENERS

C7.2. Bolted Joints

The potential for full reversal of design load and likelihood of inelastic deforma-tions of members and/or connected parts necessitates that pretensioned bolts beused in bolted joints in the Seismic Load Resisting System. However, earthquakemotions are such that slip cannot and need not be prevented in all cases, even withslip-critical connections. Accordingly, these Provisions call for bolted joints to beproportioned as pretensioned bearing joints but with faying surfaces prepared asfor Class A or better slip-critical connections. That is, bolted connections can beproportioned with Design Strengths for bearing connections as long as the fayingsurfaces are still prepared to provide a minimum slip coefficient = 0.33. The re-sulting nominal amount of slip resistance will minimize damage in more moderateseismic events. This requirement is intended for joints where the faying surfaceis primarily subjected to shear. Where the faying surface is primarily subjected totension or compression, e.g., in a bolted end plate connection, the requirement onpreparation of the faying surfaces may be relaxed.

The sharing of design load between welds and bolts on the same faying surface isnot permitted. Similarly, sharing of design loads should not be used on elementsof a member that are connected by different means. For example, the web of anaxially loaded wide flange should be connected by the same means as the flangeif the elements are resisting the same load, e.g. axial load in a brace.

To prevent excessive deformations of bolted joints due to slip between the con-nected plies under earthquake motions, the use of holes in bolted joints in theSeismic Load Resisting System is limited to standard holes and short-slotted holeswith the direction of the slot perpendicular to the line of force. An exception isprovided for alternative hole types that are justified as a part of a tested assembly.

Fabrication and erection tolerances often require that oversized holes be used. Thereis no definitive data on the behavior of connections with oversize holes under either




86 PART I – CONNECTIONS, JOINTS, AND FASTENERS [Comm. C7.

dynamic loads or large cyclic load reversals. Static test data indicate that it is diffi-cult to obtain the same preload in bolts in oversized holes as compared to standardholes. In connections for diagonal members in Braced Frames, the reduced slip ca-pacity may lead to unacceptable interstory drifts and designers should incorporatethis effect into their analyses. The loss of preload results in a lower static slip load,but the overall behavior of connections with oversized holes has been shown to besimilar to those with standard holes (Kulak, Fisher and Struik, 1987).

To prevent excessive deformations of bolted joints due to bearing on the connectedmaterial, the bearing strength is limited by the “deformation-considered” optionin LRFD Specification Section J3.10 (Rn = 0.75 × 2.4dtF u). The philosophicalintent of this limitation in the LRFD Specification is to limit the bearing defor-mation to an approximate maximum of 1/4 in. (6 mm). It should be recognized,however, that the actual bearing load in a seismic event may be much larger thanthat anticipated in design and the actual deformation of holes may exceed thistheoretical limit. Nonetheless, this limit should effectively minimize damage inmoderate seismic events.

Tension or shear fracture, bolt shear, and block shear rupture are examples of limitstates that generally result in non-ductile failure of connections. As such, these limitstates are undesirable as the controlling limit state for connections that are part ofthe Seismic Load Resisting System. Accordingly, it is required that connectionsbe configured such that a ductile limit state in the member or connection, such asyielding or bearing deformation, controls the Design Strength.

C7.3. Welded Joints

The general requirements for welded joints are given in AWS D1.1(AWS, 2000),wherein a Welding Procedure Specification (WPS) is required for all welds. Ap-proval by the Engineer of Record of the WPS to be used is required in theseProvisions.

These Provisions contain two significant changes from the 1997 Provisions: (1) forthe critical CJP welds in the special and Intermediate Moment Frame systems, theyrequire that the weld metal be made with filler metals that meet minimum levelsof Charpy V-Notch (CVN) toughness using two different test temperatures andspecified test protocols, and (2) they require weld metal notch toughness in allwelds used in members and connections in the load path of the Seismic LoadResisting System.

FEMA 350 (FEMA, 2000a) and 353 (FEMA, 2000d) also recommend filler metalthat complies with minimum Charpy V-Notch (CVN) requirements using twotest temperatures and specified test protocols for critical welded joints. TheseProvisions adopt the dual CVN requirement suggested in the FEMA documentsbut require a lower temperature than the FEMA recommendations for the AWS A5classification method (i.e. minus 20˚F rather than 0˚F). Successful testing at eithertemperature ensures that some ductile tearing will occur before final fracture. Useof this lower temperature is consistent with the filler metal used in the SAC/FEMAtests and matches the filler metals commercially available. The more critical CVNweld metal property is the minimum of 40 ft-lbf (54J) at 70˚F (21˚C) following theprocedure in Appendix X. Based on the FEMA recommendations, the Engineerof Record may consider applying the 40 ft-lbf (54J) at 70˚F (21˚C) requirements




Comm. C8.] PART I – MEMBERS 87

to other critical welds. For a discussion of special requirements for welds at lowtemperatures, see FEMA (2000a, 2000d, and 2000f).

All other welds in members and connections in the load path of the Seismic LoadResisting System require weld metal with a minimum CVN toughness of 20 ft-lbf(27J) at minus 20˚F (minus 29˚C). Welds carrying only gravity loads such as fillerbeam connections and welds for collateral members of the Seismic Load ResistingSystem such as deck welds, minor collectors, and lateral bracing do not requireweld metal with notch toughness requirements. Following the manufacturer’s es-sential variables, either the AWS classification method in the AWS A5 specificationor manufacturer certification may be used to meet this CVN requirement.

It is not the intent of these Provisions to require project-specific CVN testing ofeither the welding procedure or any production welds. Further, these weld tough-ness requirements are not intended to apply to electric resistance welding (ERW)and submerged arc welding (SAW); welding processes used in the production ofhollow structural sections and pipe (ASTM A500 and A53/A53M). In addition,the control of heat input is not monitored unless specified.

Many operations during fabrication, erection, and the subsequent work of othertrades have the potential to create discontinuities in the Seismic Load ResistingSystem. When located in regions of potential inelasticity, such discontinuities arerequired to be repaired by the responsible subcontractor as required by the En-gineer of Record. Discontinuities should also be repaired in other regions of theSeismic Load Resisting System when the presence of the discontinuity would bedetrimental to its performance. The responsible subcontractor should propose arepair procedure for the approval of the Engineer of Record. Repair may be unnec-essary for some discontinuities, subject to the approval of the Engineer of Record.The Engineer of Record should refer to AWS D1.1 and ASTM A6, Sect. 9 forguidance in establishing the acceptance criteria for repair of discontinuities. Out-side the plastic hinge regions, AWS D1.1 requirements for repair of discontinuitiesshould be applied.

C7.4. Other Connections

The FEMA/SAC testing has demonstrated the sensitivity of regions undergoinglarge inelastic strains due to discontinuities caused by welding, rapid change ofsection, penetrations, or construction caused flaws. For this reason, operations thatcause discontinuities are prohibited in the plastic hinging region. Areas whereplastic hinging is expected include Moment Frame hinging zones, Link beams ofEBFs, the ends and the center of SCBF braces, etc. The beam-column Panel Zoneis a common example of a region experiencing inelastic deformation. It should benoted that yield level strains are not strictly limited to the plastic hinge zones andcaution should also be exercised in creating discontinuities in these regions as well.

C8. MEMBERS

C8.1. Scope

It is intended that Nominal Strengths, Resistance Factors, and Design Strengths ofmembers in the Seismic Load Resisting System be determined in accordance withthe LRFD Specification, unless noted otherwise in these Provisions.




88 PART I – MEMBERS [Comm. C8.

C8.2. Local Buckling

To provide for reliable inelastic deformations in those Seismic Load ResistingSystems that require high levels of inelasticity, the width-thickness ratios ofcompression elements of the system should be less than or equal to those thatare resistant to local buckling when stressed into the inelastic range. Althoughthe width-thickness ratios for compact members, p, given in LRFD Specifica-tion Table B5.1, are sufficient to prevent local buckling before onset of yielding,the available test data suggest that these limits are not adequate for the requiredinelastic performance in several of the Seismic Load Resisting Systems. The width-thickness ratios for seismically compact members, ps , given in Table I-8-1 aredeemed adequate for ductilities to 6 or 7 (Sawyer, 1961; Lay, 1965; Kemp, 1986;Bansal, 1971). The limiting width-thickness ratios for webs in flexural compressionhave been modified (Uang and Fan, 2001) to comply with the recommendationsin FEMA 350 (FEMA, 2000a). Provisions for Special Moment Frames (SMF),members in the special segment of Special Truss Moment Frames (STMF), Spe-cial Concentrically Braced Frames (SCBF), the Links in Eccentrically BracedFrames (EBF), and H-pile design specifically reference Table I-8-1.

Diagonal web members used in the special segments of STMF systems are limitedto flat bars only at this time because of their proven high ductility without buckling.The specified limiting width-thickness ratio of 2.5 in Table I-8-1 does not vary withFy and is intended to be a practical method to limit the aspect ratio of flat bar cross-sections.

During the service life of a steel H-pile it is primarily subjected to axial compressionand acts as an axially-loaded column. Therefore, the b/t ratio limitations given inTable B5.1 of the AISC LRFD Specification (AISC, 1999) should suffice. Duringa major earthquake, because of lateral movements of pile cap and foundation, thesteel H-pile becomes a beam-column and may have to resist large bending momentsand uplift. Cyclic tests (Astaneh-Asl and Ravat, 1997) indicated that local bucklingof piles satisfying the width-thickness limitations in Table I-8-1 occurs after manycycles of loading. In addition, this local buckling did not have much effect on thecyclic performance of the pile during cyclic testing or after cyclic testing stoppedand the piles were once again under only axial load. Thus, in keeping with theScope of the Provisions, the values in Table I-8-1 are specified for Seismic DesignCategories D, E, and F, while for lower Seismic Design Categories A, B, and C,the less stringent values in LRFD Table B5.1 are permitted.

Round HSS sections used as piles in high seismic areas should also satisfy thelimiting width-thickness ratio value given in Table I-8-1.

In Section 6.2, the Expected Yield Strength, Ry Fy , of the material used in a memberis required for the purpose of determining the effect of the actual member strengthon its connections to other members of the Seismic Load Resisting System. Thewidth-thickness requirements in Table I-8-1, calculated using specified minimumyield stress, are expected to permit inelastic behavior without local buckling andneed not be computed using the Expected Yield Strength.

C8.3. Column Strength

The axial loads generated during earthquake motions in columns that are part of theSeismic Load Resisting System are expected to exceed those calculated using the





code-specified seismic loads for several reasons: (1) the reduction in lateral loadfor use in analysis of an elastic model of the structure; (2) the underestimation ofthe overturning forces in the analysis; and (3) the concurrent vertical accelerationsthat are not explicitly specified as a required load. The amplifications required inthis Section represent an approximation of these actions and are meant to providean upper bound for the required axial strength. The use of the Amplified SeismicLoad combinations account for these effects with a minimum required compressiveand tensile strength, and are to be applied without consideration of any concurrentflexural loads on the column.

The exceptions provided in the last paragraph of Section 8.3 represent self-limitingconditions wherein the required axial strength need not exceed the capability ofthe Structural System to transmit axial loads to the column. For example, if aspread footing foundation can only provide a certain resistance to uplift, there isa natural limit to the load that the system can transmit to a column. Conversely,the uplift resistance of a pile foundation designed primarily for compressive loadsmay significantly exceed the required tensile strength for the column. This wouldnot represent a system strength limit.

C8.4. Column Splices

C8.4a. General

Except for Moment Frames, the Design Strength of a column splice is requiredto equal or exceed both the Required Strength determined in Section 8.3 andthe Required Strength for axial, flexural and shear effects at the splice loca-tion determined from load combinations stipulated by the Applicable BuildingCode.

Column splices should be located away from the beam-to-column connection toreduce the effects of flexure. For typical buildings, the 4-ft (1.2 m) minimum dis-tance requirement will control. When located 4 to 5 ft (1.2 to 1.5 m) above thefloor level, field erection and construction of the column splice will generally besimplified due to improved accessibility and convenience. In general, it is rec-ommended that the splice be within the middle third of the story height. For lesstypical buildings, where the floor-to-floor height is insufficient to accommodatethis requirement, the splice should be placed as close as practicable to the midpointof the clear distance between the finished floor and the bottom flange of the beamabove. It is not intended that these column splice requirements be in conflict withapplicable safety regulations, such as the OSHA Safety Standards for Steel Erec-tion developed by the Steel Erection Negotiated Rulemaking Advisory Committee(SENRAC).

Partial-joint-penetration groove welded splices of thick column flanges exhibitvirtually no ductility under tensile loading (Popov and Steven, 1977; Bruneau,Mahin, and Popov, 1987). Consequently, column splices made with partial-joint-penetration groove welds require a 100 percent increase in Required Strength andmust be made using weld metal with minimum CVN toughness properties.

The calculation of the minimum Design Strength in Section 8.4a(2) includes theoverstrength factor Ry . This results in a minimum Design Strength that is notless than 50 percent of the expected axial yield strength of the column flanges.





A complete-joint-penetration groove weld may be considered as satisfying thisrequirement.

The possible occurrence of tensile loads in column splices utilizing partial-joint-penetration groove welds during a maximum considered earthquake should beevaluated. When tensile loads are possible, it is suggested that some restraint beprovided against relative lateral movement between the spliced column shafts. Forexample, this can be achieved with the use of flange splice plates. Alternatively, websplice plates that are wide enough to maintain the general alignment of the splicedcolumns can be used. Shake-table experiments have shown that, when columnsthat are unattached at the base reseat themselves after lifting, the performance ofa steel frame remains tolerable (Huckelbridge and Clough, 1977).

These provisions are applicable to common frame configurations. Additional con-siderations may be necessary when flexure dominates over axial compression incolumns in Moment Frames, and in end columns of tall narrow frames whereoverturning forces can be very significant. The designer should review the con-ditions found in columns in buildings with tall story heights, when large changesin column sizes occur at the splice, or when the possibility of column buckling insingle curvature over multiple stories exists. In these and similar cases, special col-umn splice requirements may be necessary for minimum Design Strength and/ordetailing.

In the 1992 AISC Seismic Provisions, beveled transitions between elements ofdiffering thickness and or width were not specifically required for butt splicesin columns subject to seismic loads. Although no column splices are known tohave failed in the Northridge Earthquake or previous earthquakes, this provisionis no longer considered to be prudent given the concern over stress concentrations,particularly at welds. Moment Frame systems are included in this requirementbecause inelastic analyses consistently suggest that large moments can be expectedat any point along the column length, despite the results of elastic analysis showingthat moments are low at the mid-height of columns in Moment Frames subjected tolateral loads. Column splices in Braced Frames can also be subject to tension due tooverturning effects. Accordingly, beveled transitions are required for all systemswith CJP groove-welded column splices. An exception to the requirements forbeveled transitions is provided when partial-joint-penetration groove welds arepermitted.

Where CJP groove welds are not used, the connection is likely to be a partial-joint-penetration groove weld. The unwelded portion of the partial-joint-penetrationgroove weld forms a crack-like notch that induces stress concentrations. A partial-joint-penetration groove weld made from one side would produce an edge crack-like notch (Barsom and Rolfe, 1999). A partial-joint-penetration groove weld madefrom both sides would produce a buried crack-like notch. The strength of suchcrack-like notches may be computed by using fracture mechanics methodology.Depending on the specific characteristics of the particular design configuration,geometry and deformation, the analysis may require elastic-plastic or plastic finiteelement analysis of the joint. The accuracy of the computed strength will dependon the finite element model and mesh size used, the assumed strength and fracturetoughness of the base metal, heat affected zone and weld metal, and on the residualstress magnitude and distribution in the joint.





C8.4b. Column Web Splices

Column web splices should be concentric with the column loads. Bolted columnweb splices are required to have connection plates on both sides of the web tominimize eccentricities.

C8.4c. Columns Not Part of the Seismic Load Resisting System

Inelastic analyses (FEMA, 2000f; FEMA, 2000g) of Moment Frame buildingshave shown the importance of the columns that are not part of the Seismic LoadResisting System in helping to distribute the seismic shears between the floors.Even columns that have beam connections considered to be pinned connectionsmay develop large bending moments and shears due to non-uniform drifts ofadjacent levels. For this reason, it is recommended that splices of such columns beadequate to develop the shear forces corresponding to these large column momentsin both orthogonal directions.

FEMA 350 (FEMA, 2000a) recommends that: “Splices of columns that are notpart of the Seismic Load Resisting System should be made in the center one-thirdof the column height, and should have sufficient shear capacity in both orthogonaldirections to maintain the alignment of the column at the maximum shear force thatthe column is capable of producing.” The corresponding commentary suggests thatthis shear should be calculated assuming plastic hinges at the ends of the columnsin both orthogonal directions.

Further review (Krawinkler, 2001) of non-linear analyses cited in FEMA 355C(FEMA, 2000d) showed that, in general, shears in such columns will be less thanone-half of the shear calculated from 2Mpc/H . For this reason Section 8.4c requiresthat the calculated shear in the splices be not less than Mpc/H .

Bolted web connections are preferred by many engineers and contractors becausethey have advantages for erection, and, when plates are placed on both sides ofthe web, they are expected to maintain alignment of the column in the eventof a flange splice fracture. Partial-joint-penetration groove welded webs are notrecommended, because fracture of a flange splice would likely lead to fracture ofthe web splice, considering the stress concentrations inherent in such welded joints.

C8.5. Column Bases

A ductile moment frame is usually expected to develop a hinge at the base of thecolumn. The Column Base detail must accommodate the required hinging rotationswhile maintaining the strength required to provide the mechanism envisioned bythe designer. These conditions are similar to the requirements for beam-to-columnconnections.

Column Bases for Moment Frames can be of several different types, as follows:(1) A rigid base assembly may be provided which is strong enough to force

yielding in the column. The designer should employ the same guidelines asgiven for the rigid fully-restrained connections. Such connections may employthick base plates, haunches, cover plates, or other strengthening as requiredto develop the column hinge. Where haunched type connections are used,hinging occurs above the haunch, and appropriate consideration should be





Fig. C-I-8.5.1. Example “rigid base” plate assembly for moment frames.

given to the stability of the column section at the hinge. See Figure C-I-8.5.1for examples of rigid base assemblies that can be designed to be capable offorcing column hinging.

(2) Large columns may be provided at the bottom level to limit the drift, and a“pinned base” may be utilized. The designer should ensure that the requiredshear capacity of the column, base plate, and anchor rods can be maintainedup to the maximum rotation that may occur. It should be recognized, however,that without taking special measures, column base connection will generallyprovide partial rotational fixity.

(3) A connection which provides “partial fixity” may be provided, so that theColumn Base is fixed up to some column moment, but the base itself yieldsbefore the column hinges. In designing a base with partial fixity, the designershould consider the principles used in the design of partially-restrained con-nections. This type of base may rely on bending of the base plate (similarto an end plate connection), bending of angles or tees, or yielding of anchorrods. In the latter case, it is necessary to provide anchor rods with adequateelongation capacity to permit the required rotation and sufficient unrestrainedlength for the yielding to occur. Shear capacity of the base plate to foundationconnection must be assured at the maximum rotation.

(4) The column may continue below the assumed seismic base (e.g. into a base-ment, crawl space, or grade beam) in such a way that the column’s fixity isassured without the need for a rigid base plate connection. The designer shouldrecognize that hinging will occur in the column, just above the seismic base.The horizontal shear to be resisted at the ends of the column below the seismicbase should be calculated considering the probable Overstrength (Ry Fy) ofthe framing. See Fig. C-I-8.5.2 for an example of a Moment Frame ColumnBase fixed within a grade beam.

For both braced frame and Moment Frame Column Bases, the designer should con-sider the base connection as similar to a beam-to-column connection and apply





Fig. C-I-8.5.2. Example of moment frame column base fixity in a grade beam.

similar principles of design and detailing. However, there are also significant dif-ferences that must be considered:

(1) Long anchor rods embedded in concrete will strain much more than the steelbolts or welds of the beam-to-column connections.

(2) Column Base plates are bearing on grout or concrete that is more compressiblethan the column flanges of beam-to-column connections.

(3) Column Base connections have significantly more longitudinal load in theplane of the flanges and less transverse load in the plane of the web, whencompared to beam-to-column connections.

(4) The shear mechanism between the Column Base and grout or concrete isdifferent from the shear mechanism between beam end plate and column flange.

(5) AISC standard Column Base anchor rod hole diameter is different from AISCstandard steel-to-steel bolt holes.

(6) Foundation rocking and rotation may be an issue, especially for isolated col-umn footings.

The Column Base connection is one of the most important elements in steel struc-tures. Damage at Column Bases during past earthquakes has been reported bymany observers (Northridge Reconnaissance Team, 1996; Midorikawa, Hasegawa,Mukai, Nishiyama, Fukuta, and Yamanouchi, 1997). Seismic design practice forthis class of connections has not been well developed (DeWolf and Ricker, 1990;Drake and Elkin, 1999) mainly because of the rather limited number of analyticaland experimental studies that have been carried out to-date (DeWolf and Sarisley,1980; Picard and Beaulieu, 1985; Thambiratnam and Paramasivam, 1986; Satoand Kamagata, 1988; Astaneh-Asl, Bergsma, and Shen, 1992; Targowski, Lamblin,and Guerlement, 1993; Ermopoulos and Stamatopoulos, 1996; Jaspart andVandegans, 1998; Stojadinovic, Spacone, Goel, and Kwon, 1998; Burda and Itani,1999; Adany, Calado, and Dunai, 2000).

Most of the experimental studies have been performed on reduced scale speci-mens representing basic types of connections simulating a column welded to anexposed base plate, which in turn is connected to a concrete foundation throughanchor rods. Test specimens have been subjected to axial loading combined with





cyclic bending to simulate the Column Base behavior in Moment Frames. Tworecent studies (Fahmy, Stojadinovic, and Goel, 2000; Lee and Goel, 2001) havenoted the importance of weld metal toughness and axis of bending of wide flangecolumn sections on ductility and energy dissipation capacity of the test specimens.Also, relative strength and stiffness of the base plate and anchor rods can signifi-cantly influence the stress distribution and failure modes. The performance of thebase connection also depends on the cyclic performance of the anchors and thesurrounding concrete (Klingner and Graces, 2001).

Many different types of Column Base connections are used in current practice.Much research work is needed in order to better understand their behavior underseismic loading and to formulate improved design procedures. Designers shoulduse caution and good judgment in design and detailing in order to achieve desiredstrength, stiffness and ductility of this very important class of connections.

The Provisions are silent on the use of Amplified Seismic Loads for ColumnBase design since this is under the purview of the Applicable Building Code.In general, Amplified Seismic Loads are prescribed to assure sufficient DesignStrength to control the locations of inelastic deformations. When a connection isrequired to be designed for Amplified Seismic Loads, the intent is to assure that theconnection is strong enough and stiff enough to allow yielding of the connectedmember. In the case of Moment Frames, if the building system performance intendscolumn yielding at the base plate, the connection between the column and the baseplate should be designed for the Amplified Seismic Load. In the case of BracedFrames, if the building system performance intends brace yielding at the baseplate, the connection between the brace and the base plate should be designed forthe Amplified Seismic Load.

C8.6. H-Piles

The provisions on seismic design of H-piles are based on the data collected onthe actual behavior of H-piles during recent earthquakes, including the 1994Northridge earthquake (Astaneh-Asl, Bolt, McMullin, Donikaian, Modjtahedi,and Cho, 1994) and the results of cyclic tests of full-scale pile tests (Astaneh-Asland Ravat, 1998; Astaneh-Asl, 2001). In the test program, five full size H-Pileswith reinforced concrete pile caps were subjected to realistic cyclic vertical andhorizontal displacements expected in a major earthquake. Three specimens werevertical piles and two specimens were batter piles. The tests established that dur-ing cyclic loading for all three vertical pile specimens a very ductile and stableplastic hinge formed in the steel pile just below the reinforced concrete pile cap.When very large inelastic cycles were applied, local buckling of flanges withinthe plastic hinge area occurred. Eventually, low cycle fatigue fracture of flangesor overall buckling of the pile occurred. However, before the piles experiencedfracture through locally buckled areas, vertical piles tolerated from 40 to 65 largeinelastic cyclic vertical and horizontal displacements with rotation of the plastichinge exceeding 0.06 radian for more than 20 cycles.

C8.6a. Design of H-Piles

Prior to an earthquake, piles, particularly vertical piles, are primarily subjectedto gravity axial load. During an earthquake, piles are subjected to horizontal and





Fig. C-I-8.6.1. Deformations of Piles and Forces Acting on an Individual Pile.

vertical displacements as shown in Figure C-I-8.6.1. The horizontal and verti-cal displacements of piles generate axial load (compression and possibly uplifttension), bending moment, and shear in the pile.

During tests of H-piles, realistic cyclic horizontal and vertical displacements wereapplied to the pile specimens. Figure C-I-8.6.2 shows test results in terms ofaxial load, bending moment interaction curves for one of the specimens. Basedon performance of test specimens, it was concluded (Astaneh-Asl, 2000) thatH-piles should be designed following the provisions of the AISC LRFD Specifi-cation regarding members subjected to combined loads.

C8.6b. Batter H-Piles

The vertical pile specimens demonstrated very large cyclic ductility as well asconsiderable energy dissipation capacity. A case study of performance of H-pilesduring the 1994 Northridge earthquake (Astaneh-Asl et al., 1994) indicated excel-lent performance for pile groups with vertical piles only. However, the batter pilespecimens did not show as much ductility as the vertical piles. The batter piles tol-erated from 7 to 17 large inelastic cycles before failure. Based on relatively limitedinformation on actual seismic behavior of batter piles, it is possible that during amajor earthquake, batter piles in a pile group fail and are no longer able to supportthe gravity load after the earthquake. Because of this possibility, the use of batterpiles to carry gravity loads is discouraged. Unless, through realistic cyclic tests, itis shown that batter piles will be capable of carrying their share of the gravity loadsafter a major earthquake, the vertical piles in Seismic Design Categories D, E, andF should be designed to support the gravity load alone, without participation ofthe batter piles.




96 PART I – SPECIAL MOMENT FRAMES (SMF) [Comm. C9.

Fig. C-I-8.6.2. Axial-Moment Interaction for H-Pile Test (Astaneh-Asl, 2000).

C8.6c. Tension in H-Piles

Due to overturning moment, piles can be subjected to tension. Piles subjectedto tension should have sufficient mechanical attachments within their embeddedarea to transfer the tension force in the pile to the pile cap or foundation. Sinceit is expected that a plastic hinge will form in the pile just under the pile cap orfoundation, the use of mechanical attachment and welds over a length of pile belowthe pile cap equal to the depth of cross section of the pile is prohibited.

C9. SPECIAL MOMENT FRAMES (SMF)

General Comments for Commentary Sections C9, C10 and C11

These Provisions include three types of steel Moment Frames: Special MomentFrames (SMF) in Section 9, Intermediate Moment Frames (IMF) (new) in Section10, and Ordinary Moment Frames (OMF) in Section 11. The provisions for thesethree moment-frame types reflect lessons learned from the Northridge and KobeEarthquakes, and from subsequent research performed by the SAC Joint Venturefor FEMA. The reader is referred to FEMA 350 (FEMA, 2000a) for an extensivediscussion of these lessons and recommendations to mitigate the conditions ob-served. Commentary on specific provisions in Section C9 is based primarily onFEMA 350 (FEMA, 2000a).

The prescriptive moment-frame connection included in the 1992 AISC SeismicProvisions was based primarily upon testing that was conducted in the early 1970s




Comm. C9.] PART I – SPECIAL MOMENT FRAMES (SMF) 97

(Popov and Stephen, 1972) indicating that for the sizes and material strengths test-ed, a moment connection with complete-joint-penetration groove welded flangesand a welded or bolted web connection could accommodate inelastic rotations inthe range of 0.01 to 0.015 radian. It was judged by engineers at the time that suchrotations, which corresponded to building drifts in the range of 2 to 2.5 percent weresufficient for adequate frame performance. Investigations conducted subsequentto the Northridge earthquake emphasized that the many changes that took place inmaterials, welding, frame configurations and member sizes since the 1970s makethe original results unsuitable as a basis for current design. Additionally, recentanalyses using time histories from certain near-fault earthquakes, including P-effects, demonstrate that drift demands may be larger than previously assumed(Krawinkler and Gupta, 1998).

The three frame types included in these Provisions offer three different levels ofexpected seismic inelastic rotation capability. SMF and IMF are designed to ac-commodate approximately 0.03and 0.01 radian, respectively. OMFs are designedto remain essentially elastic and are assumed to have only very small inelasticdemands. It is assumed that the elastic drift of typical Moment Frames is usuallyin the range of 0.01 radian and that the inelastic rotation of the beams is approxi-mately equal to the inelastic drift. These frames are assumed to accommodate totalinterstory drifts in the range of 0.04, 0.02 and 0.01 radian, respectively.

Although it is common to visualize inelastic rotations in Moment Frames occurringat beam or column “hinges”, analyses and testing demonstrate that the inelasticrotations actually combine flexural deformations at the hinges, and shear defor-mations of the Panel Zones, and deformations from other sources depending onthe configuration, unless the column webs are unusually thick. The contributionof Panel Zone deformation to inelastic rotation is considered beneficial, providedthat it neither leads to significant local column flange bending at the beam-flange-to-column-flange welds nor leads to significant column damage. Currently, theamount of Panel Zone deformation that a given connection will have and howmuch it will accommodate appears to be most reliably determined by testing.

Based upon the recommendations in FEMA 350 (FEMA, 2000a), these Provisionsrequire connections in SMF and IMF be qualified for use by testing. (Note that theIMF as defined in these Provisions is equivalent to the OMF as defined in FEMA350.) It is not the intent of these Provisions to require specific tests for each design,except where the design is sufficiently unique such that there are no publishedor otherwise available tests adequately representing the proposed configuration.For many commonly employed combinations of beam and column sizes, there arereadily available test reports in publications of AISC, FEMA, and others, includingFEMA 355D (FEMA, 2000e) and NIST/AISC (1998). Qualification testing isnot required for OMF connections, which may be proportioned following a setof prescriptive design rules that have been demonstrated by testing to provideadequate performance for the limited inelastic rotation expected for such frames.

C9.1. Scope

Special Moment Frames (SMF) are generally expected to experience significantinelastic deformation during large seismic events. It is expected that most of theinelastic deformation will take place as rotation in beam “hinges,” with some





inelastic deformation in the Panel Zone of the column. The amount of inelasticdeformation is dependent on the connection types used, the configuration, and anumber of other variables. The connections for these frames are to be qualifiedbased upon tests that demonstrate that the connection can sustain an InterstoryDrift Angle of at least 0.04 radian based upon a specified loading protocol. Otherprovisions are intended to limit or prevent excessive Panel Zone distortion, columnhinging and local buckling that may lead to inadequate frame performance in spiteof good connection performance.

C9.2. Beam-to-Column Joints and Connections

C9.2a. Requirements

Sections 9.2a and 9.2b have been rewritten to clarify the requirements and coor-dinate the requirements with Appendices P and S. Section 9.2a gives the perfor-mance and design requirements for the connections, and Section 9.2b provides therequirements for verifying that the selected connections will meet the performancerequirements.

FEMA 350 (FEMA, 2000a) recommends two criteria for qualifying drift angle(QDA) for Special Moment Frames. The “Strength degradation” drift angle, asdefined in FEMA 350, means the angle where “either failure of the connectionoccurs, or the strength of the connection degrades to less than the nominal plasticcapacity, whichever is less”. The “Ultimate” drift angle capacity is defined as theangle “at which connection damage is so severe that continued ability to remainstable under gravity loading is uncertain”. Testing to this level can be hazardous tolaboratory equipment and staff, which is part of the reason that it is seldom done.The strength degradation QDA is set at 0.04 radian and the ultimate QDA is set at0.06 radian. These values formed the basis for extensive probabilistic evaluations ofthe performance capability of various Structural Systems (FEMA, 2000f) demon-strating with high statistical confidence that frames with these types of connectionscan meet the intended performance goals. For the sake of simplicity, and becausemany connections have not been tested to the ultimate QDA, these Provisions adoptthe single criterion of the strength degradation QDA . In addition, the ultimate QDAis more appropriately used for the design of high performance structures.

Although connection qualification primarily focuses on the level of plastic rotationachieved, the tendency for connections to experience strength degradation with in-creased deformation is also of concern. Strength degradation can increase momentdemands from P- effects and the likelihood of frame instability. In the absence ofadditional information, it is recommended that this degradation should not reduceflexural strength, measured at a drift angle of 0.04 radian, to less than the nominalflexural strength, Mp, calculated using the specified minimum yield strength, Fy .Figure C-I-9.1 illustrates this behavior. Note that 0.03 radian plastic rotation isequivalent to 0.04 radian drift angle for frames with an elastic drift of 0.01 radian.

The required shear strength Vu of the beam-to-column joint is defined as thesummation of the factored gravity loads and the shear that results from the requiredflexural strengths on the two ends of the beam segment between the hinge points,which can be determined as 1.1Ry Fy Z . However, in some cases, such as whenlarge gravity loads occur or when Panel Zones are weak, rational analysis mayindicate that lower combinations of end moments are justified.





Fig. C-I-9.1. Acceptable strength degradation during hysteretic behavior, per Section 9.2b.

It should be recognized that truss Moment Frames can be designed with bottom-chord members or connections that can deform inelastically and such frames arepermitted as SMF if all of the provisions of Section 9 are met.

C9.2b. Conformance Demonstration

Section 9.2b provides requirements for demonstrating conformance with the re-quirements of Section 9.2a. It permits use of Prequalified Connections meetingthe requirements of Appendix P or use of connections qualified by tests. Thesemay be from previously documented tests or project-specific tests that meet therequirements of Appendix S.

FEMA 350 (FEMA, 2000a) includes recommendations for design and fabricationof several types of connections that are deemed prequalified for use in SpecialMoment Frames. These connection designs are based on extensive testing andanalysis performed by the SAC Joint Venture under a program funded by FEMA.When used within the limitations listed in FEMA 350 and with quality control andquality assurance requirements in FEMA 353 (FEMA, 2000b), these connectionsare commonly considered to comply with the requirements of Section 9.2. Com-parison of the proposed frame configurations with the SAC tested connections isrecommended to insure that the results are applicable.

C9.3. Panel Zone of Beam-to-Column Connections(beam web parallel to column web)

Cyclic testing has demonstrated that significant ductility can be obtained throughshear yielding in column Panel Zones through many cycles of inelastic distortion(Popov, Blondet, Stepanov, and Stojadinovic, 1996; Slutter, 1981; Becker, 1971;





Fielding and Huang, 1971; Krawinkler, 1978). Consequently, it is not generallynecessary to provide a Panel Zone that is capable of developing the full flexuralstrength of the connected beams if the Design Strength of the Panel Zone can bepredicted. However, the usual assumption that Von Mises criterion applies andthe shear strength is 0.55Fydct does not match the actual behavior observed inmany tests into the inelastic range. Due to the presence of the column flanges,strain hardening and other phenomena, Panel Zone shear strengths in excess ofFydct have been observed. Accordingly, Equation 9-1 accounts for the significantstrength contribution of thick column flanges.

Despite the ductility demonstrated by properly proportioned Panel-Zones in previ-ous studies, excessive Panel Zone distortions can adversely affect the performanceof beam-to-column connections (Englekirk, 1999). Consequently, the provisionsrequire that the Panel Zone design match that of the successfully tested connectionsused to qualify the connection being used. The balance of the procedure of Sec-tion 9.3a is intended to provide a minimum strength level to prevent excessivelyweak Panel Zones, which may lead to unacceptable column distortion. WherePrequalified Connections described in FEMA 350 (FEMA, 2000a) are used, thedesign of Panel Zones according to the methods given therein, generally meet therequirements in Section 9.3a. This should be verified by the designer.

Equation 9-1 represents a Design Strength in the inelastic range and, therefore,is for comparison to limiting strengths of connected members. v has been setequal to unity to allow a direct comparison between Design Strength of the beamand the column Panel Zone. In the LRFD Specification, the engineer is giventhe option to consider inelastic deformations of the Panel Zone in the analysis.Separate sets of equations are provided for use when these deformations are andare not considered. In the AISC Seismic Provisions, however, one set of equationsis provided and consideration of the inelastic deformation of the Panel Zone inthe analysis is required. The application of the moments at the column face todetermine the required shear strength of the Panel Zone recognizes that the beamhinging will take place at a location away from the beam-to-column connection,which will result in amplified effects on the Panel Zone shear. The previous versionof this provision included a reduction factor of 0.8 on the beam yielding effects,which was intended to recognize that, in some cases, gravity loads might inhibitthe development of plastic hinges on both sides of a column. However, there is noassurance that this will be the case, especially for one-sided connections and atperimeter frames where gravity loads may be relatively small.

This provision requires that the Panel Zone thickness be determined using themethod used to determine the Panel Zone thickness in the tested connection, witha minimum value as described in the remainder of the section. The intent is that thelocal deformation demands on the various elements in the structure be consistentwith the results of the tests that justify the use of the connection. The expected shearstrength of the Panel Zone in relation to the maximum expected demands that canbe generated by the beam(s) framing into the column should be consistent with therelative strengths that existed in the tested connection configuration. Many of theconnection tests were performed with a one-sided configuration. If the structurehas a two-sided connection configuration with the same beam and column sizesas a one-sided connection test, the Panel Zone shear demand will be about twice





that of the test. Therefore, in order to obtain the same relative strength, the PanelZone thickness to be provided in the structure should be approximately twice thatof the test.

To minimize shear buckling of the Panel Zone during inelastic deformations, theminimum Panel Zone thickness is set at one-ninetieth of the sum of its depthand width. Thus, when the column web and web doubler plate(s) each meet therequirements of Equation 9-2, their interconnection with plug welds is not required.Otherwise, the column web and web doubler plate(s) can be interconnected withplug welds as illustrated in Figure C-I-9.2 and the total Panel Zone thickness canbe used in Equation 9-2.

In the 1992 AISC Seismic Provisions, it was required that web doubler platesbe placed directly against the column web in all cases. In the 1997 revision(AISC, 1997b), the commentary noted an alternative; to place web doubler platessymmetrically in pairs spaced away from the column web. In the latter configura-tion, both the web doubler plates and the column web are required to all indepen-dently meet Equation 9-2 in order to be considered as effective.

Web doubler plates may extend between top and bottom Continuity Plates weldeddirectly to the column web and web doubler plate, or they may extend above andbelow top and bottom Continuity Plates welded to the doubler plate only. In theformer case, the welded joint connecting the Continuity Plate to the column weband web doubler plate is required to be configured to transmit the proportionate loadfrom the Continuity Plate to each element of the Panel Zone. In the latter case, thewelded joint connecting the Continuity Plate to the web doubler plate is requiredto be sized to transmit the load from the Continuity Plate to the web doubler plateand the web doubler plate thickness is required to be selected to transmit this sameload; minimum-size fillet welds per LRFD Specification Table J2.4 are used toconnect along the column-web edges.

Fig. C-I-9.2. Connecting web doubler plates with plug welds.





Shear loads transmitted to the web doubler plates from the Continuity Plates areequilibrated by shear loads along column-flange edges of the web doubler plate.It is anticipated that the Panel Zone will yield in a seismic event, and the weldsconnecting the web doubler plate to the column flanges are required to be sizedto develop the shear strength of the full web doubler plate thickness. Either acomplete-joint-penetration groove-welded joint or a fillet-welded joint can be usedas illustrated in Figure C-I-9.3. The plate thickness and column fillet radius shouldbe considered before selecting the fillet-welded joint.

The use of diagonal stiffeners for strengthening and stiffening of the Panel Zonehas not been adequately tested for low-cycle reversed loading into the inelasticrange. Thus, no specific recommendations are made at this time for special seismicrequirements for this detail.

C9.4. Beam and Column Limitations

Reliable inelastic deformation requires that width-thickness ratios of projectingelements be within those providing a cross-section resistant to local buckling intothe inelastic range. Although the width-thickness ratios for compact elements inLRFD Specification Table B5.1 are sufficient to prevent local buckling before theonset of yielding, the available test data suggest that these limits are not adequatefor the required inelastic performance in SMF. The limits given in Table I-8-1 aredeemed adequate for ductilities to 6 or 7 (Sawyer, 1961; Lay, 1965; Kemp, 1986;Bansal, 1971).

The choice of the ratio in Equation 9-3 of 2.0 (see Sect. 9.7a) as a trigger for preclud-ing this limit is based upon studies of inelastic analyses by Gupta and Krawinkler(1999), Bondy (1996) and others, that indicate that hinging of columns may notbe precluded at ratios below 2.0. Hinging of columns that do not comply with psmay result in flange local buckling, which is detrimental to column performance.

C9.5. Continuity Plates

When subjected to seismic loads, an interior column (i.e., one with adjacent mo-ment connections to both flanges) in a Moment Frame receives a tensile flangeforce on one flange and a compressive flange force on the opposite side. Whenstiffeners are required, it is normal to place a full-depth transverse stiffener oneach side of the column web. As this stiffener provides a load path for the flangeson both sides of the column, it is commonly called a Continuity Plate. The stiff-ener also serves as a boundary to the very highly stressed Panel Zone. When theformation of a plastic hinge is anticipated adjacent to the column, the RequiredStrength is the flange force exerted when the full beam plastic moment has beenreached, including the effects of overstrength and strain hardening, as well as shearamplification from the hinge location to the column face.

Post-Northridge studies have shown that even when Continuity Plates of substan-tial thickness are used, inelastic strains across the weld of the beam flange to thecolumn flange are substantially higher opposite the column web than they are at theflange tips. Some studies have indicated stress concentrations higher than 4, whichcan cause the weld stress at the center of the flange to exceed its tensile strengthbefore the flange force exceeds its yield strength based on a uniform average stress.





Fig. C-I-9.3. Web doubler plates.





This condition will be exacerbated if relatively thin Continuity Plates are used or ifContinuity Plates are omitted entirely. For this reason, an earlier formula that per-mitted elimination of Continuity Plates where column flanges were very thick wasrescinded in FEMA 267 (FEMA,1995) and the Supplement to FEMA 267 (FEMA,1997b). The use of Continuity Plates was recommended in all cases unless testsshowed that other design features of a given connection are so effective in reducingor redistributing flange stresses that the connection will work without them.

Given the stress distribution cited above, there is little justification for some of theold rules for sizing and connecting Continuity Plates, such as selecting its thicknessequal to one-half the thickness of the beam flange. On the other hand, the use ofexcessively thick Continuity Plates will likely result in large residual stresses,which also may be detrimental. Because of these apparently conflicting concepts,it is recommended that Continuity Plate usage and sizing be based on tests.

The FEMA-sponsored SAC steel project studied this issue in depth. Continuityplates are not required according to FEMA 350 (FEMA, 2000a) when:

tc f >

√1.8b f t f

Fyb Ryb

Fyc Ryc(C9-3)

and

tc f > b f /6 (C9-4)

Equation C9-3 is similar to the equation in older codes, except for the Ry factors.Justification for the use of Equation C9-3 and C9-4 is based on studies by Riclesincluded in FEMA 355D (FEMA, 2000e).

These equations will be considered for incorporation in future editions of theseProvisions.

C9.6. Column-Beam Moment Ratio

The strong-column weak-beam (SC/WB) concept is perhaps one of the least un-derstood seismic provisions in steel design. It is often mistakenly assumed that itis formulated to prevent any column flange yielding in a frame and that if suchyielding occurs, the column will fail. Tests have shown that yielding of columnsin Moment Frame subassemblages does not necessarily reduce the lateral strengthat the expected seismic displacement levels.

The SC/WB concept represents more of a global frame concern than a concern atthe interconnections of individual beams and columns. Schneider, Roeder, andCarpenter (1991) and Roeder (1987) showed that the real benefit of meetingSC/WB requirements is that the columns are generally strong enough to forceflexural yielding in beams in multiple levels of the frame, thereby achieving ahigher level of energy dissipation. Weak column frames, particularly those withweak or soft stories, are likely to exhibit an undesirable response at those storieswith the highest column demand-to-capacity ratios.

It should be noted that compliance with the SC/WB concept and Equation 9-3gives no assurance that individual columns will not yield, even when all connec-tion locations in the frame comply. It can be shown by nonlinear analysis that, as





the frame deforms inelastically, points of inflection shift and the distribution ofmoments varies from the idealized condition. Nonetheless, yielding of the beamsrather than the columns will predominate and the desired inelastic performancewill be achieved in frames composed of members that meet the requirement inEquation 9-3.

Previous formulations of this relationship idealized the beam/column intersectionas a point at the intersection of the member centerlines. Post-Northridge beam-to-column moment connections are generally configured to shift the plastic hingelocation into the beam away from the column face and a more general formula-tion was needed. FEMA 350 provides recommendations regarding the assumedlocation of plastic hinges for different connection configurations. Recognition ofpotential beam overstrength (see Commentary Section C6.2) is also incorporatedinto Equation 9-3.

Exceptions to the requirement in Equation 9-3 are given in Sections 9.6a and 9.6b.The compactness requirements in Section 9.4 must be met for columns using theseexceptions because it is expected that flexural yielding will occur in the columns.

In Section 9.6a, columns with low axial loads used in one-story buildings or in thetop story of a multi-story building need not meet Equation 9-3 because concernsfor inelastic soft or weak stories are not significant in such cases. Also excepted is alimited percentage of columns that is low enough to limit undesirable performancewhile still providing reasonable flexibility where the requirement in Equation 9-3would be impractical, such as at large transfer girders. Finally, Section 9.6 providesan exception for columns in levels that are significantly stronger than in the levelabove because column yielding at the stronger level would be unlikely.

C9.7. Beam-to-Column Connection Restraint

Columns are required to be braced to prevent rotation out of the plane of theMoment Frame, particularly if inelastic behavior is expected in or adjacent to thebeam-to-column connection during high seismic activity.

C9.7a. Restrained Connections

Beam-to-column connections are usually restrained laterally by the floor or roofframing. When this is the case and it can be shown that the column remains elasticoutside of the Panel Zone, lateral bracing of the column flanges is required onlyat the level of the top flanges of the beams. If it cannot be shown that the columnremains elastic, lateral bracing is required at both the top and bottom beam flangesbecause of the potential for flexural yielding, and consequent lateral-torsionalbuckling of the column.

The Required Strength for lateral bracing at the beam-to-column connection is 2percent of the Nominal Strength of the beam flange. In addition, the element(s)providing lateral bracing are required to have adequate stiffness to inhibit lateralmovement of the column flanges (Bansal, 1971). In some cases, a bracing memberwill be required for such lateral bracing. Alternatively, it can be shown that adequatelateral bracing can be provided by the column web and Continuity Plates or by theflanges of perpendicular beams.





The 1997 AISC Seismic Provisions required column lateral bracing when the ratioin Equation 9-3 was less than 1.25. The intent of this provision was to requirebracing to prevent lateral-torsional buckling for cases where it cannot be assuredthat the column will not hinge. Studies utilizing inelastic analyses (Gupta andKrawinkler, 1999; Bondy, 1996) have shown that, in severe earthquakes, plastichinging can occur in the columns even when this ratio is significantly larger than1.25. The revised limit of 2.0 was selected as a reasonable cut-off because columnplastic hinging for values greater than 2.0 only occurs in the case of extremely largestory drifts. The intent of the revisions to this section is to encourage appropriatebracing of column flanges rather than force the use of much heavier columns.

C9.7b. Unrestrained Connections

Unrestrained connections occur in special cases, such as in two-story frames,at mechanical floors or in atriums and similar architectural spaces. When suchconnections occur, the potential for out-of-plane buckling at the connection shouldbe minimized. Three provisions are given for the columns to limit the likelihoodof column buckling.

C9.8. Lateral Bracing of Beams

General requirements for lateral bracing of beams are given in LRFD Specifica-tion Chapter F. In Moment Frames, the beams are nearly always bent in reversecurvature between columns unless one end is pinned. Using a plastic design modelas a guide and assuming that the moment at one end of a beam is Mp and a pinnedend exists at the other, LRFD Specification Equation F1-17 indicates a maximumdistance between points of lateral bracing of 0.12ry Es /Fy . However, there remainsthe uncertainty in locating plastic hinges due to earthquake motions. Consequently,the maximum distance between points of lateral bracing is more conservativelyspecified as 0.086ry Es /Fy for both top and bottom flanges.

The provisions of this section call for the placement of lateral bracing to be near thelocation of expected plastic hinges. Such guidance dates to the original develop-ment of plastic design procedures in the early 1960’s. In Moment Frame structures,many connection details attempt to move the plastic hinge a short distance awayfrom the beam-to-column connection. Testing carried out as part of the SAC pro-gram (FEMA, 2000a) indicated that the bracing provided by typical compositefloor slabs is adequate to avoid excessive strength deterioration up to the requiredInterstory Drift Angle of 0.04 radian. As such, the FEMA recommendations do notrequire the placement of supplemental lateral bracing at plastic hinge locations ad-jacent to column connections for beams with composite floor construction. Theseprovisions allow the placement of lateral braces to be consistent with the testedconnections that are used to justify the design. For conditions where larger driftsare anticipated or improved performance is desired, the designer may decide toprovide additional lateral bracing near these plastic hinges. If lateral braces are pro-vided, they should provide a Design Strength of 6 percent of the expected capacityof the beam flange at the plastic hinge location. If a Reduced Beam Section con-nection detail is used, the reduced flange width may be considered in calculationof the bracing force.




Comm. C10.] PART I – INTERMEDIATE MOMENT FRAMES (IMF) 107

C10. INTERMEDIATE MOMENT FRAMES (IMF)

General Commentary for Sections C10 and C11

As a result of studies conducted under the SAC program (FEMA, 2000f), the In-termediate Moment Frame (IMF) as defined in the 1997 AISC Seismic Provisions(AISC, 1997b) was not referenced in FEMA 350 (FEMA, 2000a). In these Provi-sions, the previously defined Ordinary Moment Frame (OMF) has been split intotwo systems: the IMF based on a tested connection design and the OMF based ona prescriptive design procedure. Both systems are intended primarily for construc-tion limited to certain Seismic Design Categories and height restrictions (FEMA,2000e). It is intended that the new IMF and OMF will not experience the largerInterstory Drift Angles expected of SMF or the previous IMF through the use ofmore or larger framing members or because of less demanding Seismic DesignCategories. Many of the restrictions applied to the SMF are not applied to the IMFand the OMF because limited inelastic action is required.

The statement, “No additional requirements beyond the AISC LRFD Specifica-tion.” which appears in Sections 10.3, 10.4, 10.6, 10.7, 10.8, 11.3, 11.4, 11.6, 11.7,and 11.8 indicates that these Provisions require no limitations or provisions be-yond what is in the AISC Load and Resistance Factor Design Specification (AISC,1999) on that particular topic.

C10.1. Scope

The Intermediate Moment Frame (IMF) currently specified is essentially the sameas the Ordinary Moment Frame (OMF) system defined in the 1997 AISC SeismicProvisions for Structural Steel Buildings. This new IMF is intended to providelimited levels of inelastic rotation capability and is based on tested designs.

The following building height and system limitations are given in the 2000 NEHRPProvisions (FEMA, 2000g) for the IMF:

(1) There is no height limit in Seismic Design Categories (SDC) B and C.

(2) The IMF can be used in buildings up to 35 ft in height (10.7 m) regardless offloor and/or wall weight for SDC D.

(3) The IMF is not permitted in SDC’s E, and F, except as described in referencefootnote i and j.

(4) Footnote i reads “Steel Ordinary Moment Frames and Intermediate MomentFrames are permitted in single-story buildings up to a height of 60 feet (18.3 m)when the moment joints of field connections are constructed of bolted endplates and the dead load of the roof does not exceed 15 psf (0.72 kPa).”

Additionally, IMF’s are permitted in buildings up to a height of 35 feet (10.7 m)where the dead load for each of the following elements: walls, floors, and roofs,does not exceed 15 psf (0.72 kPa).


The minimum Interstory Drift Angle required for IMF connections is 0.02 radianwhile that for SMF connections is 0.04 radian. This level of Interstory Drift Anglehas been established for this type of frame based on engineering judgment applied




108 PART I – ORDINARY MOMENT FRAMES (OMF) [Comm. C11.

to available tests and analytical studies, primarily those included in FEMA (2000dand 2000f).

C11. ORDINARY MOMENT FRAMES (OMF)

C11.1. Scope

The Ordinary Moment Frame (OMF) is intended to provide for limited levels ofinelastic rotation capability that are less than those of the IMF. Unlike the IMF,the OMF is based on a prescriptive design procedure.

The following building height and system limitations are given in the 2000 NEHRPProvisions (FEMA, 2000g) for the OMF:

(1) There is no height limit on Seismic Design Categories (SDC) B and C.

(2) The OMF is not permitted in SDC’s D, E, and F, except as described inreference footnotes i and j (see Section C10.1).


Even though the inelastic rotation demands on OMF are expected to be low, theNorthridge Earthquake damage demonstrated that little, if any, inelastic rotationalcapacity was available in the connection prescribed by the codes prior to 1994.Thus, even for OMF, new connection requirements are needed, and these areprovided in this section.

The specific requirements given for connections are given for both fully-restrained(FR) and partially-restrained (PR) moment connections. For FR moment connec-tions, a minimum calculated strength of 1.1Ry Mp is required to recognize potentialoverstrength and strain hardening of the beam. Additionally, detailing enhance-ments are required that have been demonstrated by FEMA 350 (FEMA, 2000a) tosignificantly improve the connection performance over past steel Moment Frameconstruction.

One such enhancement is the prescribed weld access hole given in Figure 11-1and in FEMA 350 (FEMA, 2000a). The requirement to use the above weld ac-cess hole configuration is not stipulated for SMF nor IMF connections since theapproved joints are based on testing. For OMF joints only, a minor increase incertain tolerances might be appropriate to standardize the shape of access holesfor different beam sizes. Provided that the slope of the access hole cut to the flangedoes not exceed 25˚, such adjustments to Figure 11-1 could include increasing theplus tolerance of Note 2 to 3/4 tb f , Note 3 to + 1/2 in. (12 mm), and Note 5 to + 1in. (25 mm). Other weld access holes may provide good performance with lowerfabrication costs and can be used if verified by testing.

The testing completed by the SAC Joint Venture found that improved performanceinto the inelastic range can be obtained with the following improvements overthe prescriptive pre-Northridge connection detail: (1) the use of notch-tough weldmetal; (2) the removal of backing bars, backgouging of the weld root, and reweldingwith a reinforcing fillet weld; (3) the use of a welded web connection; (4) the useof Continuity Plates; and (5) use of the weld access hole detail as noted above.Where the top flange steel backing is left in place, the steel backing is welded to theflange with a continuous fillet. (See Figure C-I-11.1(d).) The steel backing should




Comm. C11.] PART I – ORDINARY MOMENT FRAMES (OMF) 109

Fig. C-I-11.1. Schematic illustrations of strong-axis moment connections.

not be welded to the underside of the beam flange. Discussion of the connectiondetailing is provided in FEMA documents 350 and 353 (FEMA, 2000a; FEMA,2000b).

FEMA 350 (FEMA, 2000a) did not prequalify welded connections of beams tothe weak axis of columns due to lack of sufficient test data. Use of moment con-nections to the weak axis of columns require that several adjustments be madeto Section 11.2(1). The bottom flange Continuity Plate should be thicker than thebeam flange and set lower than theoretical underside of beam to facilitate beam




110 PART I – SPECIAL TRUSS MOMENT FRAMES (STMF) [Comm. C12.

depth tolerance. The Continuity Plates should project 3 in. beyond the columnflange and be tapered to the width of the beam flange. Continuity Plates shouldbe provided on the far side of the column web. The bottom flange steel backingshould be removed, and a weld transition made to the thickened Continuity Plate.The steel backing may remain at the top flange. See LRFD Manual of Steel Con-struction (AISC, 2001), Driscoll and Beedle (1982), and Gilton and Uang (2002)for information on fully-rigid connections to the column weak axis.

For information on bolted moment end-plate connections in seismic applications,refer to Meng and Murray (1997) and FEMA 355D (FEMA, 2000e).

For information on PR connections, the reader is referred to Leon (1990 and 1994),Leon and Ammerman (1990), Leon and Forcier (1992), Bjorhovde, Colson, andBrozzetti (1990), Hsieh and Deierlein (1991), Leon, Hoffman, and Staeger (1996)and FEMA (2000e).

A welded beam-to-column moment connection in a strong-axis configurationsimilar to one tested at Lehigh University for the SAC Project is illustrated inFigure C-I-11.1(d). FEMA350 (FEMA, 2000a) recommends this detail for use inOMF with similar member sizes and other conditions.

C11.5. Continuity Plates

For all welded OMF connections that are not based upon tests, Continuity Platesare required.

When welding Continuity Plates to the column flanges with two-sided partial-joint-penetration groove welds combined with reinforcing fillet welds, refer toAWS D1.1, Article 2.6.2 and Annex I for an explanation of effective throat forreinforced partial-joint-penetration groove welds.

The “contact area” referred to in this section is the thickness of the ContinuityPlate times its length, after reductions in length for access holes.

C12. SPECIAL TRUSS MOMENT FRAMES (STMF)

C12.1. Scope

Truss-girder Moment Frames have often been designed with little or no regardfor truss ductility. Research has shown that such truss Moment Frames have verypoor hysteretic behavior with large, sudden reductions in strength and stiffnessdue to buckling and fracture of web members prior to or early in the dissipationof energy through inelastic deformations (Itani and Goel, 1991; Goel and Itani,1994a). The resulting hysteretic degradation as illustrated in Figure C-I-12.1 resultsin excessively large story drifts in building frames subjected to earthquake groundmotions with peak accelerations on the order of 0.4g to 0.5g.

Research work led to the development of special truss girders that limit inelasticdeformations to a special segment of the truss (Itani and Goel, 1991; Goel andItani, 1994b; Basha and Goel, 1994). As illustrated in Figure C-I-12.2, the chordsand web members (arranged in an X pattern) of the special segment are designedto withstand large inelastic deformations, while the rest of the structure remains




Comm. C12.] PART I – SPECIAL TRUSS MOMENT FRAMES (STMF) 111

Fig. C-I-12.1. Strength degradation in undetailed truss girder.

elastic. Special Truss Moment Frames (STMF) have been validated by extensivetesting of full-scale subassemblages with story-high columns and full-span spe-cial truss girders. As illustrated in Figure C-I-12.3, STMF are ductile with stablehysteretic behavior for a large number of cycles up to 3 percent story drifts.

Because STMF are relatively new and unique, the span length and depth of thetruss girders are limited at this time to the range used in the test program.

C12.2. Special Segment

It is desirable to locate the STMF special segment near mid-span of the trussgirder because shear due to gravity loads is generally lower in that region. Thelower limit on special segment length of 10 percent of the truss span length providesa reasonable limit on the ductility demand, while the upper limit of 50 percent ofthe truss span length represents more of a practical limit.

The Required Strength of interconnection for X-diagonals is intended to accountfor buckling over half the full diagonal length (El-Tayem and Goel, 1986; Goeland Itani, 1994b). It is recommended that half the full diagonal length be usedin calculating the design compression strength of the interconnected X-diagonalmembers in the special segment.

Because it is intended that the yield mechanism in the special segment form overits full length, no major structural loads should be applied within the length of the




112 PART I – SPECIAL TRUSS MOMENT FRAMES (STMF) [Comm. C12.

Fig. C-I-12.2. Cross-braced truss.

Fig. C-I-12.3. Hysteretic behavior of STMF.

special segment. In special segments with open Vierendeel panels, i.e. when nodiagonal web members are used, any structural loads should be avoided. Accord-ingly, a restrictive upper limit is placed on the axial load in diagonal web membersdue to gravity loads applied directly within the special segment.

C12.3. Nominal Strength of Special Segment Members

STMF are intended to dissipate energy through flexural yielding of the chordmembers and axial yielding and buckling of the diagonal web members in thespecial segment. It is desirable to provide minimum shear strength in the specialsegment through flexural yielding of the chord members and to limit the axial




Comm. C13.] PART I – SPECIAL CONCENTRICALLY BRACED FRAMES (SCBF) 113

load to a maximum value. Plastic analysis can be used to determine the requiredshear strength of the truss special segments under the factored earthquake loadcombination.

C12.4. Nominal Strength of Non-Special Segment Members

STMF are required to be designed to maintain elastic behavior of the truss mem-bers, columns, and all connections, except for the members of the special segmentthat are involved in the formation of the yield mechanism. Therefore, all membersand connections outside the special segments are to be designed for calculatedloads by applying the combination of gravity loads and equivalent lateral loadsthat are necessary to develop the maximum expected nominal shear strength ofthe special segment Vne in its fully yielded and strain-hardened state. Thus, Equa-tion 12-1, as formulated, accounts for uncertainties in the actual yield strength ofsteel and the effects of strain hardening of yielded web members and hinged chordmembers. It is based upon approximate analysis and test results of special trussgirder assemblies that were subjected to story drifts up to 3 percent (Basha andGoel, 1994). Tests (Jain, Goel, and Hanson, 1978) on axially loaded members haveshown that 0.3Pnc is representative of the average nominal post-buckling strengthunder cyclic loading.

C12.5. Compactness

The ductility demand on diagonal web members in the special segment can berather large. Flat bars are suggested at this time because of their high ductility.Tests (Itani and Goel, 1991) have shown that single angles with width-thicknessratios that are less than 0.18

√Es/Fy also possess adequate ductility for use as

web members in an X configuration. Chord members in the special segment arerequired to be compact cross-sections to facilitate the formation of plastic hinges.

C12.6. Lateral Bracing

The top and bottom chords are required to be laterally braced to provide for thestability of the special segment during cyclic yielding. The lateral bracing limit forflexural members L p as specified in the LRFD Specification has been found to beadequate for this purpose.

C13. SPECIAL CONCENTRICALLY BRACED FRAMES (SCBF)

C13.1. Scope

Concentrically Braced Frames are those Braced Frames in which the centerlinesof members that meet at a joint intersect at a point to form a vertical truss systemthat resists lateral loads. A few common types of concentrically Braced Framesare shown in Figure C-I-13.1, including diagonally braced, cross-braced (X), V-braced (or inverted-V-braced) and K-braced configurations. Because of their ge-ometry, concentrically Braced Frames provide complete truss action with memberssubjected primarily to axial loads in the elastic range. However, during a moderateto severe earthquake, the bracing members and their connections are expected toundergo significant inelastic deformations into the post-buckling range.




114 PART I – SPECIAL CONCENTRICALLY BRACED FRAMES (SCBF) [Comm. C13.

Fig. C-I-13.1. Examples of concentric bracing configurations.

Since the initial adoption of concentrically Braced Frames into seismic designcodes, more emphasis has been placed on increasing brace strength and stiffness,primarily through the use of higher design loads in order to minimize inelasticdemand. More recently, requirements for ductility and energy dissipation capabilityhave also been added. Accordingly, provisions for Special Concentrically BracedFrames (SCBF) were developed to exhibit stable and ductile behavior in the eventof a major earthquake. Earlier design provisions have been retained for OrdinaryConcentrically Braced Frames (OCBF) in Section 14.

During a severe earthquake, bracing members in a concentrically Braced Frameare subjected to large deformations in cyclic tension and compression into thepost-buckling range. As a result, reversed cyclic rotations occur at plastic hingesin much the same way as they do in beams and columns in Moment Frames. Infact, braces in a typical concentrically Braced Frame can be expected to yield andbuckle at rather moderate story drifts of about 0.3 percent to 0.5 percent. In a severeearthquake, the braces could undergo post-buckling axial deformations 10 to 20times their yield deformation. In order to survive such large cyclic deformationswithout premature failure the bracing members and their connections must beproperly detailed.

Damage during past earthquakes and that observed in laboratory tests of con-centrically Braced Frames has generally resulted from the limited ductility andcorresponding brittle failures, which are usually manifested in the fracture of con-nection elements or bracing members. The lack of compactness in braces results insevere local buckling, the resulting high concentration of flexural strains at theselocations and reduced ductility. Braces in concentrically Braced Frames are subjectto severe local buckling, with diminished effectiveness in the nonlinear range atlow story drifts. Large story drifts that result from early brace fractures can imposeexcessive ductility demands on the beams and columns, or their connections.

Research has demonstrated that concentrically Braced Frames, with proper con-figuration, member design and detailing can possess ductility far in excess of thatpreviously ascribed to such systems. Extensive analytical and experimental workby Goel and others has shown that improved design parameters, such as limitingwidth/thickness ratios (to minimize local buckling), closer spacing of stitches andspecial design and detailing of end connections greatly improve the post-bucklingbehavior of concentrically Braced Frames. The design requirements for SCBF arebased on those developments.





Previous requirements for concentrically Braced Frames sought reliable behaviorby limiting global buckling. Cyclic testing of Diagonal Bracing systems verifiesthat energy can be dissipated after the onset of global buckling if brittle failures dueto local buckling, stability problems and connection fractures are prevented. Whenproperly detailed for ductility as prescribed in these Provisions, diagonal bracescan sustain large inelastic cyclic deformations without experiencing prematurefailures.

Analytical studies (Tang and Goel, 1987; Hassan and Goel, 1991) on bracing sys-tems designed in strict accordance with earlier code requirements for concentricallyBraced Frames predicted brace failures without the development of significant en-ergy dissipation. Failures occurred most often at plastic hinges (local buckling dueto lack of compactness) or in the connections. Plastic hinges normally occur at theends of a brace and at the brace midspan. Analytical models of bracing systemsthat were designed to ensure stable ductile behavior when subjected to the sameground motion records as the previous concentrically Braced Frame designs ex-hibited full and stable hysteresis without fracture. Similar results were observedin full-scale tests by Wallace and Krawinkler (1985) and Tang and Goel (1989).

For double-angle and double-channel braces, closer stitch spacing, in addition tomore stringent compactness criteria, is required to achieve improved ductility andenergy dissipation. This is especially critical for double-angle and double-channelbraces that buckle imposing large shear forces on the stitches. Studies also showedthat placement of double angles in a toe-to-toe configuration reduces bendingstrains and local buckling (Aslani and Goel; 1991).

Many of the failures reported in concentrically Braced Frames due to strong groundmotions have been in the connections. Similarly, cyclic testing of specimens de-signed and detailed in accordance with typical provisions for concentrically BracedFrames has produced connection failures (Astaneh-Asl, Goel, and Hanson, 1986).Although typical design practice has been to design connections only for axialloads, good post-buckling response demands that eccentricities be accounted forin the connection design, which should be based upon the maximum loads the con-nection may be required to resist. Good connection performance can be expectedif the effects of brace member cyclic post-buckling behavior are considered (Goel,1992c).

For brace buckling in the plane of the gusset plates, the end connections should bedesigned for the full axial load and flexural strength of the brace (Astaneh-Asl et al.,1986). Note that a realistic value of K should be used to represent the connectionfixity.

For brace buckling out of the plane of single plate gussets, weak-axis bending in thegusset is induced by member end rotations. This results in flexible end conditionswith plastic hinges at midspan in addition to the hinges that form in the gusset plate.Satisfactory performance can be ensured by allowing the gusset plate to developrestraint-free plastic rotations. This requires that the free length between the endof the brace and the assumed line of restraint for the gusset be sufficiently longto permit plastic rotations, yet short enough to preclude the occurrence of platebuckling prior to member buckling. A length of two times the plate thickness isrecommended (Astaneh-Asl et al., 1986). Note that this free distance is measuredfrom the end of the brace to a line that is perpendicular to the brace centerline,





Fig. C-I-13.2. Brace-to-gusset plate requirement for buckling out-of-plane bracing system.

drawn from the point on the gusset plate nearest to the brace end that is constrainedfrom out-of-plane rotation. See Figure C-I-13.2. Alternatively, connections withstiffness in two directions, such as crossed gusset plates, can be detailed. Test resultsindicate that forcing the plastic hinge to occur in the brace rather than the connectionplate results in greater energy dissipation capacity (Lee and Goel, 1987).

Since the stringent design and detailing requirements for SCBF are expected toproduce more reliable performance when subjected to high energy demands im-posed by severe earthquakes, model building codes have reduced the design loadlevel below that required for OCBF.

Bracing connections should not be configured in such a way that beams or columnsof the frame are interrupted to allow for a continuous brace element. This provisionis necessary to improve the out-of-plane stability of the bracing system at thoseconnections.

A Zipper Column system and a two-story X system are illustrated inFigure C-I-13.3. Two-story X and zipper-Braced Frames can be designed withpost-elastic behavior consistent with the expected behavior of V-braced SCBF.These configurations can also capture the increase in post-elastic axial loads onbeams at other levels. It is possible to design 2-story X and zipper frames withpost-elastic behavior that is superior to the expected behavior of V-braced SCBFby proportioning elements to discourage single-story mechanisms.





Fig. C-I-13.3. (a) Two-story X-Braced Frame, (b) “Zipper-Column” with Inverted-V bracing.

C13.2. Bracing Members

C13.2a. Slenderness

The slenderness (Kl/r ) limit has been raised to 5.87√

Es/Fy for SCBF. Themore restrictive limit of 4.23

√Es/Fy as specified for OCBF in Section 14.2

is not necessary when the bracing members are detailed for ductile behavior.Tang and Goel (1989) and Goel and Lee (1992) showed that the post-bucklingcyclic fracture life of bracing members generally increases with an increase inslenderness ratio. An upper limit is provided to maintain a reasonable level ofcompressive strength.

C13.2c. Lateral Force Distribution

This provision attempts to balance the tensile and compressive resistance acrossthe width and breadth of the building since the buckling and post-bucklingstrength of the bracing members in compression can be substantially less thanthat in tension. Good balance helps prevent the accumulation of inelastic driftsin one direction. An exception is provided for cases where the bracing membersare sufficiently oversized to provide essentially elastic response.

C13.2d. Width-thickness Ratios

Width-thickness ratios of compression elements in bracing members have beenreduced to be at or below the requirements for compact sections in order to min-imize the detrimental effects of local buckling and subsequent fracture duringrepeated inelastic cycles. Tests have shown this failure mode to be especiallyprevalent in rectangular HSS with width-thickness ratios larger than the pre-scribed limits (Hassan and Goel, 1991; Tang and Goel, 1989).





C13.2e. Built-up Members

Closer spacing of stitches and higher stitch strength requirements are specifiedfor built-up bracing members in SCBF (Aslani and Goel, 1991; Xu and Goel,1990) than those required for OCBF. These are intended to restrict individualelement bending between the stitch points and consequent premature fracture ofbracing members. Wider spacing is permitted under an exception when bucklingdoes not cause shear in the stitches. Bolted stitches are not permitted within themiddle one-fourth of the clear brace length as the presence of bolt holes in thatregion may cause premature fractures due to the formation of a plastic hinge inthe post-buckling range.

C13.3. Bracing Connections

C13.3a. Required Strength

In concentrically Braced Frames, the bracing members normally carry most ofthe seismic story shear, particularly if not used as a part of a Dual System. TheRequired Strength of bracing connections should be adequate so that failure byout-of-plane gusset buckling or brittle fracture of the brace are not critical failuremechanisms.

The minimum of two criteria, (i.e. the nominal expected axial tension strengthof the bracing member and the maximum force that could be generated by theoverall system) determines the Required Strength of both the bracing connectionand the beam-to-column connection if it is part of the bracing system. This upperlimit is included in the specification for structures where elements other thanthe tension bracing limit the system strength; for example, foundation elementsdesigned in systems based on the application of load combinations using theAmplified Seismic Load. Ry has been added to the first provision to recognizethe material overstrength of the member.

C13.3b. Tensile Strength

Braces in Special Concentrically Braced Frames are required to have gross-section tension yielding as their governing limit state so that they will yield in aductile manner. Local connection failure modes such as net-section fracture andblock-shear rupture must be precluded. Therefore, the calculations for these fail-ure modes must use the maximum load that the brace can deliver.

It should be noted that some, if not all, steel materials commonly used for braceshave Expected Yield Strengths significantly higher than their specified minimumyield strengths; some have Expected Yield Strengths almost as high as their tensilestrength. For such cases, no significant reduction of the brace section is permis-sible and connections may require local reinforcement of the brace section. Thisis the case for knife-plate connections between gusset plates and A53 or A500braces (e.g. pipe braces or square, rectangular, or round hollow structural steel(tube) braces), where the over-slot of the brace required for erection leaves areduced section. If this section is left unreinforced, net-section fracture will bethe governing limit state and brace ductility may be significantly reduced (Korol,1996; Cheng, Kulak, and Khoo, 1998). Reinforcement may be provided in theform of steel plates welded to the tube, increasing the effective area at the reduced





Fig. C-I-13.4. P- diagram for a strut.

brace section. Braces with two continuous welds to the gusset wrapped aroundits edge (instead of the more typical detail with four welds stopping short of thegusset edge) performed adequately in the tests by Cheng. However, this practicemay be difficult to implement in field conditions; it also creates a potential stressriser that may lead to crack initiation.

C13.3c. Flexural Strength

Braces with “fixed” end connections have been shown to dissipate more energythan those that are “pin” connected, because buckling requires the formation ofthree plastic hinges in the brace. Nonetheless, end connections that can accommo-date the rotations associated with brace buckling deformations while maintainingadequate strength have also been shown to have acceptable performance. Testinghas demonstrated that where a single gusset plate connection is used, the rotationscan be accommodated as long as the brace end is separated by at least two timesthe gusset thickness from a line perpendicular to the brace axis about which thegusset plate may bend unrestrained by the beam, column, or other brace joints(Astaneh-Asl et al., 1986). This condition is illustrated in Figure C-I-13.2 andprovides hysteretic behavior as illustrated in Figure C-I-13.4. The distance of “2t”shown in Figure C-I-13.2 should be considered the minimum offset distance. Inpractice, it may be advisable to specify a slightly larger distance (perhaps “3t”)on construction documents to provide for erection tolerances. More informationon seismic design of gusset plates can be obtained from Astaneh-Asl (1998).

Where “fixed” end connections are used in one axis with “pinned” connectionsin the other axis, the effect of the fixity should be considered in determining thecritical buckling axis.





C13.4. Special Bracing Configuration Special Requirements

C13.4a. V-Type and Inverted V-Type Bracing

V-braced and Inverted-V-Braced Frames exhibit a special problem that sets themapart from Braced Frames in which both ends of the braces frame into beam-column joints. Upon continued lateral displacement as the compression bracebuckles, its force drops while that in the tension brace continues to increase up tothe point of yielding. This creates an unbalanced vertical force on the intersectingbeam. In order to prevent undesirable deterioration of lateral strength of the frame,the SCBF provisions require that the beam possess adequate strength to resist thispotentially significant post-buckling load redistribution (the unbalanced load) incombination with appropriate gravity loads. Tests have shown that typical brac-ing members demonstrate a residual post-buckling compressive strength of about30 percent of the initial compressive strength (Hassan and Goel, 1991). This is themaximum compression load that should be combined with the full yield load ofthe adjacent tension brace. The full tension load can be expected to be in the rangeofPy . In addition, configurations where the beam-to-brace connection is signif-icantly offset from the midspan location should be avoided whenever possible,since such a configuration exacerbates the unbalanced conditions cited above.The adverse effect of this unbalanced load can be mitigated by using bracingconfigurations, such as V- and Inverted-V-braces in alternate stories creating anX- configuration over two story modules, or by using a “Zipper Column” withV- or Inverted-V bracing (Khatib, Mahin, and Pister, 1988). See Figure C-I-13.3.Adequate lateral bracing at the brace-to-beam intersection is necessary in orderto prevent adverse effects of possible lateral-torsional buckling of the beam.

The requirements in Sections 13.4a(1) and 13.4a(2) provide for a minimumstrength of the beams to support gravity loads in the event of loss of bracecapacities.

The limitations of Sections 13.4a(2) and 13.4a(3) need not be applied on thebeam strength of roof stories, penthouses, and one-story structures as the lifesafety consequences of excessive beam deformations may not be as severe as forfloors.

C13.4b. K-Type Bracing

K-bracing is generally not considered desirable in concentrically Braced Framesand is prohibited entirely for SCBF because it is considered undesirable to havecolumns that are subjected to unbalanced lateral forces from the braces, as theseforces may contribute to column failures.

C13.5. Columns

In the event of a major earthquake, columns in concentrically Braced Frames canundergo significant bending beyond the elastic range after buckling and yieldingof the braces. Even though their bending strength is not utilized in the design pro-cess when elastic design methods are used, columns in SCBF are required to haveadequate compactness and shear and flexural strength in order to maintain theirlateral strength during large cyclic deformations of the frame. Analytical studieson SCBF that are not part of a Dual System have shown that columns can carry




Comm. C14.] PART I – ORDINARY CONCENTRICALLY BRACED FRAMES (OCBF) 121

Fig. C-I-13.5. Base shear vs. story drift of a SCBF.

as much as 40 percent of the story shear (Tang and Goel, 1987; Hassan and Goel,1991). When columns are common to both SCBF and SMF in a Dual System, theircontribution to story shear may be as high as 50 percent. This feature of SCBFgreatly helps in making the overall frame hysteretic loops “full” when comparedwith those of individual bracing members which are generally “pinched” (Hassanand Goel, 1991; Black, Wenger, and Popov, 1980). See Figure C-I-13.5.

C14. ORDINARY CONCENTRICALLY BRACED FRAMES (OCBF)

C14.1. Scope

These Provisions assume that the Applicable Building Code significantly restrictsthe permitted use of OCBF because of their limited ductility. Specifically, it is as-sumed that the restrictions given in the NEHRP Seismic Provisions (FEMA, 2000g)govern the design of the structure. The NEHRP Seismic Provisions effectively re-strict the use of OBCF as described in Section C14.2. If similar restrictions are notfound in the Applicable Building Code used for a given structure (e.g. 1997 Uni-form Building Code (ICBO, 1997)), then the design of the concentrically BracedFrame should comply with the detailing and design requirements for SCBF.

C14.2. Strength

In the 1997 AISC Seismic Provisions, there were relatively few differences be-tween Ordinary Concentrically Braced Frames (OCBFs) and Special Concentri-cally Braced Frames (SCBFs). Despite the lower R-value given in the 1997 NEHRPseismic provisions (FEMA, 1997a), these systems may not perform well in largeground motions. Consequently the OCBF provisions, except those previously givenfor “Low Buildings,” have been eliminated.




122 PART I – ORDINARY CONCENTRICALLY BRACED FRAMES (OCBF) [Comm. C14.

The specific reasons for elimination of most of the OCBF provisions that were inthe 1997 AISC Seismic Provisions include:

(1) Section 14.3a.b allowed connections to be designed for a strength that may beless than that of the braces themselves. This will preclude ductile performanceof the system.

(2) Section 14.4a.1 required that braces in V-Type and Inverted V-Type bracingsystems be designed for “at least 1.5 times the strength using LRFD Specifi-cation [1993 version] Load combinations A4-5 and A4-6.” This may lead tooverly strong bracing, which will be capable of buckling the columns of theBraced Frame, and may lead to collapse.

(3) Section 14.4a.3 does not provide for sufficient beam strength to maintain thestrength of the tension brace after buckling of the compression brace. Theresult is that buckling of the compression brace or buckling and/or rotationof the gusset plate can lead to a sudden and dramatic reduction in the storystrength.

The provisions in Section 13 for SCBF’s attempt to prevent all the above unde-sirable characteristics. It is the intent of the Provisions that SCBF be used for allconcentrically Braced Frames where significant ductility is needed. In order toaccomplish this, the following items are included in the 2000 NEHRP Provisions(FEMA, 2000g):

(1) The use of Ordinary Steel Concentrically Braced Frames (OCBF’s) in theDual Systems is not included.

(2) The height limit of Seismic Design Categories (SDC) D and E are limited to35 ft and OCBF’S are not permitted for SDC F, except as noted below.

(3) Each of these three categories has a reference footnote k.

(4) Footnote k reads “Steel ordinary Braced Frames are permitted in single storybuildings up to a height of 60 ft when the dead load of the roof does not exceed15 psf, and in penthouse structures.”

The term “brace connection” includes the brace-to-gusset connection as well asthe gusset plate-to-column, beam, or base plate connection.

The application of the Amplified Seismic Load to determine the member size andconnections other than bracing connections, in SDC D and E buildings with anR factor of about 2.5 would provide sufficient strength to preclude the need forsignificant ductility of the system.

The effect of these modifications on the design of steel concentrically BracedFrames in comparison to those designed in accordance with these Seismic Provi-sions will be as follows:

(1) Most concentrically Braced Frames in the higher Seismic Design Categorieswill be classified as SCBF.

(2) V-Type and Inverted V-Type SCBF frames will have lighter braces, but sig-nificantly heavier floor beams.

(3) All configurations will be permitted to use larger Kl/r values, which may resultin lighter braces. Connections of the braces may be heavier, depending upon




Comm. C15.] PART I – ECCENTRICALLY BRACED FRAMES (EBF) 123

whether or not the requirement to develop the strength of the braces for SCBFis offset by the lighter bracing.

C15. ECCENTRICALLY BRACED FRAMES (EBF)

C15.1. Scope

Research has shown that EBF can provide an elastic stiffness that is comparableto that of SCBFs and OCBFs, particularly when short Link lengths are used, andexcellent ductility and energy dissipation capacity in the inelastic range, compara-ble to that of SMF, provided that Links are not too short (Roeder and Popov; 1978;Libby, 1981; Merovich, Nicoletti, and Hartle, 1982; Hjelmstad and Popov, 1983;Malley and Popov, 1984; Kasai and Popov, 1986a and 1986b; Ricles and Popov,1987a and 1987b; Engelhardt and Popov, 1989a and 1989b; Popov, Engelhardt,and Ricles, 1989). EBF are composed of columns, beams, and braces in whichat least one end of each bracing member connects to a beam at a short distancefrom an adjacent beam-to-brace connection or a beam-to-column connection as il-lustrated in Figure C-I-15.1. This short beam segment, called the Link, is intendedas the primary zone of inelasticity. These provisions are intended to ensure thatcyclic yielding in the Links can occur in a stable manner while the diagonal braces,columns, and portions of the beam outside of the Link remain essentially elasticunder the forces that can be generated by fully-yielded and strain-hardened Links.

Figure C-I-15.1 identifies the key components of an EBF: the Links, the beamsegments outside of the Links, the diagonal braces, and the columns. Requirementsfor Links are provided in Sections 15.2 to 15.5; requirements for beam segmentsoutside of the Links and for the diagonal braces are provided in Sections 15.6 and15.7; requirements for columns are provided in Section 15.8.

In some bracing arrangements, such as that illustrated in Figure C-I-15.2 withLinks at each end of the brace, Links may not be fully effective. If the upper Linkhas a significantly lower design shear strength than that for the Link in the storybelow, the upper Link will deform inelastically and limit the force that can bedelivered to the brace and to the lower Link. When this condition occurs the upperLink is termed an active Link and the lower Link is termed an inactive Link. Thepresence of potentially inactive Links in an EBF increases the difficulty of analysis.

It can be shown with plastic frame analyses that, in some cases, an inactive Linkwill yield under the combined effect of dead, live and earthquake loads, therebyreducing the frame strength below that expected (Kasai and Popov, 1984). Fur-thermore, because inactive Links are required to be detailed and constructed asif they were active, and because a predictably inactive Link could otherwise bedesigned as a pin, the cost of construction is needlessly increased. Thus, an EBFconfiguration that ensures that all Links will be active, such as those illustrated inFigure C-I-15.1(d), are recommended. Further recommendations for the design ofEBF are available (Popov et al., 1989).

These Provisions are primarily intended to cover the design of EBF in which theLink is a horizontal framing member located between the column and a braceor between two braces. For the Inverted Y-Braced EBF configuration shown inFigure C-I-15.1(d), the Link is attached underneath the beam. If this configurationis to be used, lateral bracing should be provided at the intersection of the diagonal




124 PART I – ECCENTRICALLY BRACED FRAMES (EBF) [Comm. C15.

Fig. C-I-15.1. Examples of eccentrically braced frames.

braces and the vertical link, unless calculations are provided to justify the designwith such bracing.

Columns in EBF should be designed following capacity design principles so thatthe full strength and deformation capacity of the frame can be developed withoutfailure of any individual column and without the formation of a soft story. Plastichinge formation in columns should be avoided because, when combined with





Fig. C-I-15.2. EBF – active and inactive link.

hinge formation in the Links, it can result in the formation of a soft story. Therequirements of Sections 8.3 and 15.8 address column design.

C15.2. Links

Inelastic action in EBF is intended to occur primarily within the Links. The generalprovisions in this section are intended to ensure that stable inelasticity can occurin the Link.

The Link cross-section is required to meet the same width-thickness criteria as isspecified for beams in SMF (Table I-8-1).

The majority of experiments conducted on EBF have used Links with a specifiedminimum yield stress of 50 ksi (345 MPa) or less. In order to stay within thebounds of this experimental database, Links are not permitted to be made of steelswith a minimum specified yield stress in excess of 50 ksi (345 MPa).

The reinforcement of Links with web doubler plates is not permitted as suchreinforcement may not fully participate as intended in inelastic deformations. Ad-ditionally, beam web penetrations within the Link are not permitted because theymay adversely affect the inelastic behavior of the Link.

The Link design shear strength Vn is the lesser of that determined from the plasticshear strength of the Link section or twice the plastic moment divided by the Linklength, as dictated by statics assuming equalization of end moments. Accordingly,





Link design shear strength can be computed as follows:

Vn =

Vp for e ≤ 2Mp

Vp

2Mp

e for e >2Mp

Vp

(C15.3-1)

This design shear strength should then be greater than or equal to the required shearstrength determined from the load combinations as required by the ApplicableBuilding Code, but without the inclusion of Amplified Seismic Loads.

The effects of axial load on the Link can be ignored if the required axial strengthon the Link does not exceed 15 percent of the nominal yield strength of the LinkPy . In general, such an axial load is negligible because the horizontal componentof the brace load is transmitted to the beam segment outside of the Link. However,when the framing arrangement is such that larger axial forces can develop inthe Link, such as from drag struts or a modified EBF configuration, the additionalrequirements in Section C15.2 apply and the design shear strength and Link lengthsare required to be reduced to ensure stable inelastic behavior.

To assure satisfactory behavior of an EBF, the inelastic deformation expected tooccur in the Links in a severe earthquake should not exceed the inelastic defor-mation capacity of the Links. In these Provisions, the Link Rotation Angle is theprimary variable used to describe inelastic Link deformation. The Link RotationAngle is the plastic rotation angle between the Link and the portion of the beamoutside of the Link.

The Link Rotation Angle can be estimated by assuming that the EBF bay willdeform in a rigid-plastic mechanism as illustrated for various EBF configurationsin Figure C-I-15.3. In this figure, the Link Rotation Angle is denoted by the symbolp. The Link Rotation Angle can be related to the plastic story drift angle, p,using the relationships shown in the Figure C-I-15.3. The plastic story drift angle,in turn, can be computed as the plastic story drift, p divided by the story height,h. The plastic story drift can conservatively be taken equal to the Design StoryDrift. Alternatively, the Link Rotation Angle can be determined more accuratelyby inelastic dynamic analyses.

The inelastic response of a Link is strongly influenced by the length of the Linkas related to the ratio Mp/Vp of the Link cross-section. When the Link lengthis selected not greater than 1.6Mp/Vp, shear yielding will dominate the inelasticresponse. If the Link length is selected greater than 2.6Mp/Vp, flexural yield-ing will dominate the inelastic response. For Link lengths intermediate betweenthese values, the inelastic response will occur through some combination of shearand flexural yielding. The inelastic deformation capacity of Links is generallygreatest for shear yielding Links, and smallest for flexural yielding Links. Basedon experimental evidence, the Link Rotation Angle is limited to 0.08 radian forshear yielding Links (e ≤ 1.6Mp/Vp) and 0.02 radian for flexural yielding Links(e ≥ 2.6Mp/Vp). For Links in the combined shear and flexural yielding range(1.6Mp/Vp < e < 2.6Mp/Vp), the limit on Link Rotation Angle is determinedaccording to Link length by linear interpolation between 0.08 and 0.02 radian.

It has been demonstrated experimentally (Whittaker, Uang, and Bertero, 1987;Foutch, 1989) as well as analytically (Popov et al., 1989) that Links in the first floor





Fig. C-I-15.3. Link Rotation Angle.

usually undergo the largest inelastic deformation. In extreme cases this may resultin a tendency to develop a soft story. The plastic Link rotations tend to attenuate athigher floors, and decrease with the increasing frame periods. Therefore for severeseismic applications, a conservative design for the Links in the first two or threefloors is recommended. This can be achieved by increasing the minimum designshear strengths of these Links on the order of 10 percent over that specified inSection 15.2.

C15.3. Link Stiffeners

A properly detailed and restrained Link web can provide stable, ductile, and pre-dictable behavior under severe cyclic loading. The design of the Link requiresclose attention to the detailing of the Link web thickness and stiffeners.





Full-depth stiffeners are required at the ends of all Links and serve to transfer theLink shear forces to the reacting elements as well as restrain the Link web againstbuckling.

The maximum spacing of Link Intermediate Web Stiffeners in shear yieldingLinks (e ≤ 1.6 Mp/Vp) is dependent upon the size of the Link Rotation Angle(Kasai and Popov, 1986b) with a closer spacing required as the rotation angleincreases. Intermediate Web Stiffeners in shear yielding Links are provided todelay the onset of inelastic shear buckling of the web. Flexural yielding Linkshaving lengths greater than 2.6Mp/Vp but less than 5Mp/Vp are required to havean intermediate stiffener at a distance from the Link end equal to 1.5 times the beamflange width to limit strength degradation due to flange local buckling and lateral-torsional buckling. Links of a length that are between the shear and flexural limitsare required to meet the stiffener requirements for both shear and flexural yieldingLinks. When the Link length exceeds 5Mp/Vp, Link Intermediate Web Stiffenersare not required. Link Intermediate Web Stiffeners are required to extend full depthin order to effectively resist shear buckling of the web and to effectively limitstrength degradation due to flange local buckling and lateral-torsional buckling.Link Intermediate Web Stiffeners are required on both sides of the web for Links25 in. (635 mm) in depth or greater. For Links that are less than 25 in. (635 mm)deep, the stiffener need be on one side only.

All Link stiffeners are required to be fillet welded to the Link web and flanges. Thewelds to the Link web are required to provide a Design Strength that is equal to thenominal vertical tensile yield strength of the stiffener in a section perpendicular toboth the plane of the web and the plane of the stiffener or the shear yield strengthof the stiffener, whichever is less. The connection to the Link flanges is designedfor correspondingly similar forces.

C15.4. Link-to-Column Connections

Prior to the 1994 Northridge Earthquake, Link-to-column connections were typi-cally constructed in a manner substantially similar to beam-to-column connectionsin SMFs. Link-to-column connections in EBF are therefore likely to share manyof the same problems observed in Moment Frame connections. Consequently, in amanner similar to beam-to-column connections in SMFs, these Provisions requirethat the performance of Link-to-column connections be verified by testing in accor-dance with Appendix S, or by the use of prequalified Link-to-column connectionsin accordance with Appendix P.

The load and deformation demands at a Link-to-column connection in an EBFare substantially different from those at a beam-to-column connection in an SMF.Consequently, beam-to-column connections which qualify for use in an SMF maynot necessarily perform adequately when used as a Link-to-column connection inan EBF. Link-to-column connections must therefore be tested in a manner thatproperly simulates the forces and inelastic deformations expected in an EBF.

These Provisions permit the use of Link-to-column connections without the needfor qualification testing for shear yielding Links when the connection is reinforcedwith haunches or other suitable reinforcement designed to preclude inelastic ac-tion in the reinforced zone adjacent to the column. This reinforced region shouldremain essentially elastic for the fully yielded and strain hardened Link strength as





defined in Section 15.6 for the design of the diagonal brace. That is, the reinforcedconnection should be designed to resist the Link shear and moment generated bythe expected shear strength of the Link Ry Vn increased by 125 percent to accountfor strain hardening. Alternatively, the EBF can be configured to avoid Link-to-column connections entirely.

These Provisions do not explicitly address the column Panel Zone design require-ments at Link-to-column connections, as little research is available on this issue.However, based on research on Panel Zones for SMF systems, limited yielding ofPanel Zones in EBF systems would not be detrimental. Pending future researchon this topic, it is recommended that the required shear strength of the Panel Zonebe determined from Equation 9-1 with the flexural demand at the column end ofthe Link given by the equations in Commentary Section 15.6.

C15.5. Lateral Bracing of the Link

Lateral restraint against out-of-plane displacement and twist is required at the endsof the Link to ensure stable inelastic behavior. The Required Strength for suchlateral bracing is 6 percent of the Expected Yield Strength of the beam flange.In typical applications, a composite deck can likely be counted upon to provideadequate lateral bracing at the top flange of the Link. However, a composite deckalone cannot be counted on to provide adequate lateral bracing at the bottom flangeof the Link and direct bracing through transverse beams or a suitable alternativeis recommended.

C15.6. Diagonal Brace and Beam Outside of Links

This section addresses design requirements for the diagonal brace and the beamsegment outside of the Link in EBF. The intent of these provisions is to assure thatyielding and energy dissipation in an EBF occur primarily in the Links. Conse-quently, the diagonal brace and beam segment outside of the Link must be designedto resist the loads generated by the fully yielded and strain hardened Link. Thatis, the brace and beam should be designed following capacity design principlesto develop the full inelastic capacity of the Links. Limited yielding outside of theLinks, particularly in the beams, is sometimes unavoidable in an EBF. Such yield-ing is likely not detrimental to the performance of the EBF, as long as the beamand brace have sufficient strength to develop the Link’s full inelastic strength anddeformation capacity.

In most EBF configurations, the diagonal brace and the beam are subject to largeaxial loads combined with significant bending moments. Consequently, both thediagonal brace and the beam should be designed as beam-columns.

A diagonal brace in a concentrically Braced Frame is subject to cyclic bucklingand is the primary source of energy dissipation in such a frame. Many of thedesign provisions for OCBF and SCBF systems are intended to permit stablecyclic buckling behavior of the diagonal braces. A properly designed diagonalbrace in an EBF, on the other hand, should not buckle, regardless of the intensityof the earthquake ground motion. As long as the brace is designed to be strongerthan the Link, as is the intent of these provisions, then the Link will serve as afuse to limit the maximum load transferred to the brace, thereby precluding the





possibility of brace buckling. Consequently, many of the design provisions forbraces in OCBF and SCBF systems intended to permit stable cyclic buckling ofbraces are not needed in EBF. Similarly, the Link also limits the loads transferredto the beam beyond the link, thereby precluding failure of this portion of the beamif it is stronger than the Link.

The diagonal brace and beam segment outside of the Link must be designed forsome reasonable estimate of the maximum forces that can be generated by the fullyyielded and strain hardened Link. For this purpose, the nominal shear strength ofthe Link, Vn , as defined by Equation C15.3-1 is increased by two factors. First, thenominal shear strength is increased by Ry to account for the possibility that theLink material may have actual yield strength in excess of the specified minimumvalue. Secondly, the resulting expected shear strength of the Link, Ry Vn is furtherincreased to account for strain hardening in the Link.

Experiments have shown that Links can exhibit a high degree of strain hardening.Recent tests on rolled wide-flange Links constructed of A992 steel (Arce, 2002)showed strength increases due to strain hardening ranging from 1.2 to 1.45, withan average value of about 1.30. Past tests on rolled wide-flange Links constructedof A36 steel have sometimes shown strength increases due to strain hardening inexcess of 1.5 (Hjelmstad and Popov, 1983; Engelhardt and Popov, 1989a). Further,recent tests on very large welded built-up wide-flange Links for use in major bridgestructures have shown strain hardening factors close to 2.0 (McDaniel, Uang, andSeible, 2002; Dusicka and Itani, 2002). These sections, however, typically haveproportions significantly different from rolled shapes.

For purposes of designing the diagonal brace, these Provisions have adopted a stre-ngth increase due to strain hardening equal to 1.25. This value is close to but some-what below the average measured strain hardening factor for rolled wide-flangeLinks of A992 steel. Designers should recognize that strain hardening in Links maysometimes exceed this value, and so a conservative design of the diagonal braceis appropriate. Further, if large built-up Link sections are used, designers shouldconsider the possibility of strain hardening factors substantially in excess of 1.25.

Based on the above, the Required Strength of the diagonal brace can be taken asthe forces generated by the following values of Link Shear and Link end moment:

For e ≤ 2Mp/Vp:Link shear = 1.25Ry VpLink end moment = e(1.25Ry Vp)/2

For e > 2Mp/Vp:Link shear = 2(1.25Ry Mp)/eLink end moment = 1.25Ry Mp

The above equations assume Link end moments will equalize as the Link yieldsand deforms plastically. For Link lengths less than 1.6Mp/Vp attached to columns,Link end moments do not fully equalize (Kasai and Popov, 1986a). For this situa-tion, the Link ultimate forces can be estimated as follows:

For Links attached to columns with e ≤ 1.6Mp/Vp:Link shear = 1.25Ry VpLink end moment at column = Ry MpLink end moment at brace = [e(1.25 Ry Vp) − Ry Mp] ≥ 0.75Ry Mp





The Link shear force will generate axial force in the diagonal brace, and for mostEBF configurations, will also generate substantial axial force in the beam segmentoutside of the Link. The ratio of beam or brace axial force to Link shear force iscontrolled primarily by the geometry of the EBF and is therefore not affected byinelastic activity within the EBF (Engelhardt and Popov, 1989a). Consequently,this ratio can be determined from an elastic frame analysis and can be used toamplify the beam and brace axial forces to a level that corresponds to the Linkshear force specified in the above equations. Further, as long as the beam and braceare designed to remain essentially elastic, the distribution of Link end moment tothe beam and brace can be estimated from an elastic frame analysis. For example,if an elastic analysis of the EBF under lateral load shows that 80 percent of theLink end moment is resisted by the beam and the remaining 20 percent is resistedby the brace, the ultimate Link end moments given by the above equations canbe distributed to the beam and brace in the same proportions. Alternatively, aninelastic frame analysis can be conducted for a more accurate estimate of howLink end moment is distributed to the beam and brace in the inelastic range.

As described above, these Provisions assume that as a Link deforms to large plasticrotations, the Link expected shear strength will increase by a factor of 1.25 dueto strain hardening. However, for the design of the beam segment outside of theLink, these Provisions permit calculation of beam Required Strength based onLink ultimate forces equal to only 1.1 times the Link expected shear strength. Thisrelaxation on Link ultimate forces for purposes of designing the beam segmentreflects the view that beam strength will be substantially enhanced by the presenceof a composite floor slab, and also that limited yielding in the beam will not likelybe detrimental to EBF performance, as long as stability of the beam is assured.Consequently, designers should recognize that the actual forces that will develop inthe beam will be substantially greater than computed using this 1.1 factor, but thislow value of required beam strength will be mitigated by contributions of the floorslab in resisting axial load and bending moment in the beam and by limited yieldingin the beam. Based on this approach, the required axial and flexural strength of thebeam can be first computed as described above for the diagonal brace, assuminga strain hardening factor of 1.25. The resulting axial force and bending momentin the beam can then be reduced by a factor of 1.1/1.25 = 0.88. In cases whereno composite slab is present, designers should consider computing required beamstrength based on a Link strain hardening factor of 1.25.

For most EBF configurations, the beam and the Link are a single continuous wideflange member. If this is the case, the Design Strength of the beam can be increasedby Ry . If the Link and the beam are the same member, any increase in yield strengthpresent in the Link will also be present in the beam segment outside of the Link.

Design of the beam segment outside of the Link can sometimes be problematic inEBF. In some cases, the beam segment outside of the Link is inadequate to resist theRequired Strength based on the Link ultimate forces. For such cases, increasing thesize of the beam may not provide a solution. This is because the beam and the Linkare typically the same member. Increasing the beam size therefore increases theLink size, which in turn, increases the Link ultimate forces and therefore increasesthe beam Required Strength. The relaxation in beam Required Strength based onthe 1.1 factor on Link strength was adopted by these Provisions largely as a resultof such problems reported by designers, and by the view that EBF performance





would not likely be degraded by such a relaxation due to beneficial effects of thefloor slab and limited beam yielding, as discussed above. Design problems withthe beam can also be minimized by using shear yielding Links (e ≤1.6Mp/Vp)as opposed to longer Links. The end moments for shear yielding Links will besmaller than for longer Links, and consequently less moment will be transferredto the beam. Beam moments can be further reduced by locating the intersection ofthe brace and beam centerlines inside of the Link, as described below. Providing adiagonal brace with a large flexural stiffness so that a larger portion of the Link endmoment is transferred to the brace and away from the beam can also substantiallyreduce beam moment. In such cases, the brace must be designed to resist theselarger moments. Further, the connection between the brace and the Link must bedesigned as a Fully Restrained moment resisting connection. Test results on severalbrace connection details subject to axial load and bending moment are reported inEngelhardt and Popov (1989a).

Avoiding very shallow angles between the diagonal brace and the beam can alsomitigate problems with beam design. As the angle between the diagonal braceand the beam decreases, the axial load generated in the beam increases. Usingangles between the diagonal brace and the beam of at least about 40 degreeswill often be beneficial in reducing beam required axial strength. Problems withdesign of the beam segment outside of the Link can also be addressed by choosingEBF configurations that minimize axial loads in the beam. An example of such aconfiguration is illustrated in Engelhardt and Popov (1989b).

Typically in EBF design, the intersection of the brace and beam centerlines islocated at the end of the Link. However, as permitted in Section 15.6, the braceconnection may be designed with an eccentricity so that the brace and beam cen-terlines intersect inside of the Link. This eccentricity in the connection generates amoment that is opposite in sign to the Link end moment. Consequently, the valuegiven above for the Link end moment can be reduced by the moment generated bythis brace connection eccentricity. This may substantially reduce the moment thatwill be required to be resisted by the beam and brace, and may be advantageous indesign. The intersection of the brace and beam centerlines should not be locatedoutside of the Link, as this increases the bending moment generated in the beamand brace. See Figures C-I-15.4 and C-I-15.5.

C15.7. Beam-to-Column Connections

If the arrangement of the EBF system is such that a Link is not adjacent to acolumn and large axial loads are not present in the beam, a simple connection canbe adequate if the connection provides some restraint against torsion in the beam.The magnitude of torsion to be considered is calculated from a pair of equal andopposite forces equal to 2 percent of the nominal beam flange yield strength.

C15.8. Required Column Strength

Similar to the diagonal brace and beam segment outside of the Link, the columnsof an EBF should also be designed using capacity design principles. That is, thecolumns should be designed to resist the maximum forces generated by the fullyyielded and strain hardened Links. As discussed in Section C15.6, the maximumshear force developed by a fully yielded and strain hardened Link can be estimated





Fig. C-I-15.4. EBF with W-shape bracing.

Fig. C-I-15.5. EBF with HSS bracing.

as 1.25Ry times the Link nominal shear strength Vn , where the 1.25 factor accountsfor strain hardening. For capacity design of the columns, this section permitsreduction of the strain hardening factor to 1.1. This relaxation reflects the viewthat all Links above the level of the column under consideration will not likelyreach their maximum shear strength simultaneously. Consequently, applying the1.25 strain hardening factor to all Links above the level of the column underconsideration is likely too conservative for a multistory EBF. For a low rise EBF




134 PART I – QUALITY ASSURANCE [Comm. C16.

with only a few stories, designers should consider increasing the strain hardeningfactor on Links to 1.25 for capacity design of the columns, since there is a greaterlikelihood that all Links may simultaneously reach their maximum shear strength.In addition to the requirements of this section, columns in EBF must also bechecked in accordance with the requirements of Section 8.3, which are applicableto all systems.

C16. QUALITY ASSURANCE

To assure ductile seismic response, steel framing is required to meet the qualityrequirements as appropriate for the various components of the structure. ASCE7 (ASCE, 2002) provides special requirements for inspection and testing basedupon the Seismic Design Category. Additionally, these Provisions, the AISC LRFDSpecification for Structural Steel Buildings, AISC Code of Standard Practice, AWSD1.1, and the RCSC Specification for Structural Joints Using ASTM A325 or A490Bolts provide acceptance criteria for steel building structures.

These Provisions require that a quality assurance plan be implemented as requiredby the Engineer of Record. In some cases, the contractor implements such a qualitycontrol plan as part of normal operations, particularly contractors that participatein the AISC Quality Certification Program for steel fabricators. The Engineerof Record should evaluate the quality assurance needs for each project with dueconsideration of what is already a part of the contractor’s quality control plan.Where additional needs are identified, such as for innovative connection details orunfamiliar construction methods, supplementary requirements should be specifiedas appropriate. It should be noted that site observation by the Engineer of Recordis a critical portion of the quality assurance plan.

Visual inspection by a qualified inspector prior to, during, and after welding isemphasized as the primary method used to evaluate the conformance of weldedjoints to the applicable quality requirements. Joints are examined prior to thecommencement of welding to check fit-up, preparation bevels, gaps, alignment,and other variables. During welding, adherence to the WPS is maintained. Afterthe joint is welded, it is then visually inspected to the requirements of AWS D1.1.

The subsequent use of other nondestructive testing methods as required by theEngineer of Record is recommended to verify the soundness of welds that aresubject to tensile loads as a part of the Seismic Load Resisting Systems describedin Sections 9 through 15. Ultrasonic testing (UT) is capable of detecting seriousembedded flaws in groove welds in all standard welded joint configurations. Aprogram of UT of critical groove welds in tension should be specified by theEngineer of Record as a part of the quality assurance plan, and may be required forcertain joints by the Authority Having Jurisdiction. UT is not suitable for inspectingmost fillet welds, nor should it be relied upon for the detection of surface or near-surface flaws. Magnetic particle testing (MT) is capable of detecting serious flawson or near the surface of all types of welds, and should be considered by theEngineer of Record for the inspection of critical fillet welded joints and for thesurface examination of critical groove welds subjected to tension loads. The use ofPenetrant Testing (PT) is not recommended for general weld inspection, but may beused for crack detection in specific locations such as weld access holes and in the k-Area of welded shapes, or for the location of crack tips for cracks detected visually.




Comm. C16.] PART I – QUALITY ASSURANCE 135

Commentary Section C6.3 indicates that the k-Area of rotary-straightened wide-flange columns may have reduced notch toughness. Preliminary recommendations(AISC, 1997a) discouraged the placement of welds in this area because of post-weld cracking that occurred on past projects. Where such welds are to be placed, itis recommended to inspect these areas to verify that such cracking has not occurred.Typically, such inspections would incorporate magnetic particle or dye penetranttesting with acceptance criteria as specified in AWS D1.1. The required frequencyof such inspections should be specified in the contract documents.

FEMA 353, Recommended Specifications and Quality Assurance Guidelines forSteel Moment-Frame Construction for Seismic Applications (FEMA, 2000b), is areference for the preparation of a quality assurance plan for steel SMF and IMF,as well as for other Seismic Load Resisting Systems. In addition, the AuthorityHaving Jurisdiction may have specific quality assurance plan requirements.




136

APPENDIX P

PREQUALIFICATION OF BEAM-TO-COLUMN ANDLINK-TO-COLUMN CONNECTIONS

PART I

CP1. SCOPE

Appendix P describes requirements for prequalification of beam-to-column con-nections in Special and Intermediate Moment Frames and Link-to-column con-nections in Eccentrically Braced Frames. The concept of prequalified beam-to-column connections for Moment Frame systems, as used in these Provisions, hasbeen adopted from FEMA 350 (FEMA, 2000a), and has been extended to includeprequalified Link-to-column connections for EBF.

Previous editions of these Provisions released since the 1994 Northridge Earth-quake required that the design of beam-to-column and Link-to-column connectionsbe based on qualifying cyclic tests per Appendix S. This requirement was basedon the view that the behavior of connections under severe cyclic loading cannot bepredicted by analytical means alone, and consequently, that the satisfactory per-formance of all connections be confirmed by laboratory testing. To satisfy theserequirements, designers were required to provide substantiating test data, eitherfrom project specific tests or from tests reported in the literature, on connectionsmatching project conditions within the limits specified in Appendix S.

The introduction of Prequalified Connections in these Provisions does not alter thefundamental view that the performance of beam-to-column and Link-to-columnconnections should be confirmed by testing. However, it is recognized that requir-ing designers to provide substantiating test data for each new project is unneces-sarily burdensome, particularly when the same connections are used on a repeatedbasis that have already received extensive testing, evaluation and review.

It is the intent of these Provisions that designers be permitted to use PrequalifiedConnections without the need to present laboratory test data, as long as the con-nection design, detailing and quality assurance measures conform to the limits andrequirements of the prequalification. The introduction of Prequalified Connectionsis intended to simplify the design and design approval process by removing theonus on designers to present test data, and by removing the onus on the AuthorityHaving Jurisdiction to review and interpret test data. The use of Prequalified Con-nections is not intended as a guarantee against damage to, or failure of, connectionsin major earthquakes. A licensed professional engineer, designer, or architect shalldesign such connections, as with any others, for any specific application with com-petent professional examination and verification of their accuracy, suitability, andapplicability.

The use of Prequalified Connections is permitted, but not required, by theseProvisions. Connections that are not prequalified are still permitted, as long as




Comm. CP2.] PART I – SCOPE 137

qualifying cyclic test data is presented in accordance with the requirements ofAppendix S.

CP2.1. Basis for Prequalification

In general terms, a Prequalified Connection is one that has undergone sufficienttesting, analysis, evaluation and review so that a high level of confidence exists thatthe connection can fulfill the performance requirements specified in Section 9.2for Special Moment Frames, in Section 10.2 for Intermediate Moment Frames, orin Section 15.4 for Eccentrically Braced Frames. Prequalification should be basedprimarily on laboratory test data, but supported by analytical studies of connec-tion performance and by the development of detailed design criteria and designprocedures. The behavior and expected performance of a Prequalified Connec-tion should be well understood and predictable. Further, a sufficient body of testdata should be available to ensure that a Prequalified Connection will perform asintended on a consistent and reliable basis.

Further guidance on prequalification of connections is provided by the commentaryfor FEMA 350, which indicates that the following four criteria should be satisfiedfor a Prequalified Connection:

1. There is sufficient experimental and analytical data on the connection per-formance to establish the likely yield mechanisms and failure modes for theconnection.

2. Rational models for predicting the resistance associated with each mechanismand failure mode have been developed.

3. Given the material properties and geometry of the connection, a rational proce-dure can be used to estimate which mode and mechanism controls the behaviorand deformation capacity (that is, Interstory Drift Angle) that can be attainedfor the controlling conditions.

4. Given the models and procedures, the existing database is adequate to permitassessment of the statistical reliability of the connection.

CP2.2. Authority for Prequalification

While the general basis for prequalification is outlined in Section P2.1, it is notpossible to provide highly detailed and specific criteria for prequalification, consid-ering the wide variety of possible connection configurations, and considering thecontinually changing state-of-the-art in the understanding of connection perfor-mance. It is also recognized that decisions on whether or not a particular connectionshould be prequalified, and decisions on establishing limits on prequalification,will ultimately entail a considerable degree of professional engineering judgment.Consequently, a fundamental premise of these provisions is that prequalificationcan only be established based on an evaluation of the connection by a panel ofknowledgeable individuals. Thus, these Provisions call for the establishment of aConnection Prequalification Review Panel (CPRP). Such a panel should consistof individuals with a high degree of experience, knowledge and expertise in con-nection behavior, design and construction. It is the responsibility of the CPRP toreview all available data on a connection, and then determine if the connection




138 PART I – TESTING REQUIREMENTS [Comm. CP3.

warrants prequalification and determine the associated limits of prequalification,in accordance with Appendix P. It is the intent of these Provisions that only a sin-gle, nationally recognized CPRP be established in order to maintain consistencyin the prequalification process.

CP3. TESTING REQUIREMENTS

It is the intent of these Provisions that laboratory test data form the primary basisof prequalification, and that the connection testing conforms to the requirements ofAppendix S. FEMA 350 specifies the minimum number of tests on non-identicalspecimens needed to establish prequalification of a connection, or subsequently tochange the limits of prequalification. However, in these Provisions, the number oftests needed to support prequalification or to support changes in prequalificationlimits is not specified. The number of tests and range of testing variables neededto support prequalification decisions will be highly dependent on the particularfeatures of the connection and on the availability of other supporting data. Con-sequently, this section requires that the CPRP determine if the number and typeof tests conducted on a connection are sufficient to warrant prequalification orto warrant a change in prequalification limits. Both FEMA 350 and these Pro-visions refer to “non-identical” test specimens, indicating that a broad range ofvariables potentially affecting connection performance should be investigated ina prequalification test program. It may also be desirable to test replicates of nom-inally identical specimens in order to investigate repeatability of performance.Individuals planning a test program to support prequalification of a connection areencouraged to consult with the CPRP, in advance, for a preliminary assessment ofthe planned testing program.

Tests used to support prequalification are required to comply with Appendix S.This appendix requires test specimens be loaded at least to an Interstory DriftAngle as specified in Section 9.2 for Special Moment Frames or in Section 10.2for Intermediate Moment Frames, or a Link Rotation Angle as specified in Section15.4 for Eccentrically Braced Frames. These provisions do not include the ad-ditional requirement for connection rotation capacity at failure, as recommendedin FEMA 350 (FEMA, 2000a). For purposes of prequalification, however, it isdesirable to load specimens to larger deformation levels in order to reveal the ul-timate controlling failure modes. Prequalification of a connection requires a clearunderstanding of the controlling failure modes for a connection, i.e., the failuremodes that control the strength and deformation capacity of the connection. Con-sequently, test data must be available to support connection behavior models overthe full range of loading, from the initial elastic response to the inelastic range ofbehavior, and finally through to the ultimate failure of the connection.

When a connection is being considered for prequalification by the CPRP, all testdata for that connection must be available for review by the CPRP. This alsoincludes data on unsuccessful tests. Individuals seeking prequalification of a con-nection are obliged to present the entire known database of tests for the connection.Such data is essential for an assessment of the reliability of a connection. Notethat unsuccessful tests do not necessarily preclude prequalification, particularlyif the reasons for unsuccessful performance have been identified and addressedin the connection design procedures. For example, if ten tests are conducted onvarying sized members and one test is unsuccessful, the cause for the “failure”




Comm. CP6.] PART I – PREQUALIFICATION RECORD 139

should be determined. If possible, the connection design procedure should be ad-justed in such a way to preclude the failure and not invalidate the other nine tests.Subsequent tests should then be performed to validate the final proposed designprocedure.

CP4. PREQUALIFICATION VARIABLES

This section provides a list of variables that can affect connection performance,and that should be considered in the prequalification of connections. The CPRPshould consider the possible effects of each variable on connection performance,and establish limits of application for each variable. Laboratory tests or analyticalstudies investigating the full range of all variables listed in this section are notrequired and would not be practical. Connection testing and/or analytical studiesinvestigating the effects of these variables are only required where deemed neces-sary by the CPRP. However, regardless of which variables are explicitly consideredin testing or analytical studies, the CPRP should still consider the possible effectsof all variables listed in this section, and assign appropriate limits.

CP5. DESIGN PROCEDURE

In order to prequalify a connection, a detailed and comprehensive design procedureconsistent with the test results and addressing all pertinent limit states must beavailable for the connection. This design procedure must be included as part ofthe prequalification record, as required in Section P6. Examples of the format andtypical content of such design procedures can be found in FEMA 350 (FEMA,2000a).

CP6. PREQUALIFICATION RECORD

A written prequalification record is required for a Prequalified Connection. As aminimum, the prequalification record must include the information listed in Sec-tion P6. The prequalification record should provide a comprehensive listing of allinformation needed by a designer to determine the applicability and limitations ofthe connection, and information needed to design the connection. The prequali-fication record need not include detailed records of laboratory tests or analyticalstudies. However, a list of references should be included for all test reports, re-search reports, and other publications used as a basis of prequalification. Thesereferences should, to the extent possible, be available in the public domain to per-mit independent review of the data and to maintain the integrity and credibility ofthe prequalification process. FEMA 350 (FEMA, 2000a) provides an example ofthe type and formatting of information needed for a Prequalified Connection.




140

APPENDIX S

QUALIFYING CYCLIC TESTS OF BEAM-TO-COLUMNAND LINK-TO-COLUMN CONNECTIONS

PART I

CS1. SCOPE AND PURPOSE

The development of testing requirements for beam-to-column moment connec-tions was motivated by the widespread occurrence of flange weld fractures in suchconnections in the 1994 Northridge Earthquake. To improve performance of con-nections in future earthquakes, laboratory testing is required to identify potentialproblems in the design, detailing, materials, or construction methods to be used forthe connection. The requirement for testing reflects the view that the behavior ofconnections under severe cyclic loading cannot be reliably predicted by analyticalmeans only.

It is recognized that testing of connections can be costly and time consuming.Consequently, this Appendix has been written with the simplest testing require-ments possible, while still providing reasonable assurance that connections testedin accordance with these Provisions will perform satisfactorily in an earthquake.Where conditions in the actual building differ significantly from the test condi-tions specified in this Appendix, additional testing beyond the requirements hereinmay be needed to ensure satisfactory connection performance. Many of the factorsaffecting connection performance under earthquake loading are not completelyunderstood. Consequently, testing under conditions that are as close as possibleto those found in the actual building will provide for the best representation ofexpected connection performance.

It is not intended in these Provisions that project-specific connection tests be con-ducted on a routine basis for building construction projects. Rather, it is anticipatedthat most projects would use connection details that have been previously prequal-ified in accordance with Appendix P. If connections are being used that have notbeen prequalified, then connection performance must be verified by testing in ac-cordance with Appendix S. However, even in such cases, tests reported in the litera-ture can be used to demonstrate that a connection satisfies the strength and rotationrequirements of these Provisions, so long as the reported tests satisfy the require-ments of this Appendix. Consequently, it is expected that project-specific connec-tion tests would be conducted for only a very small number of construction projects.

Although the provisions in this Appendix predominantly address the testing ofbeam-to-column connections in Moment Frames, they also apply to qualifyingcyclic tests of Link-to-column connections in EBF. While there are no reports offailures of Link-to-column connections in the Northridge Earthquake, it cannotbe concluded that these similar connections are satisfactory for severe earthquakeloading as it appears that few EBF with a Link-to-column configuration were




Comm. CS3.] PART I – DEFINITIONS 141

subjected to strong ground motion in that earthquake. Many of the conditionsthat contributed to poor performance of moment connections in the NorthridgeEarthquake can also occur in Link-to-column connections in EBF. Consequently,the same testing requirements are applied to both moment connections and toLink-to-column connections.

When developing a test program, the designer should be aware that the AuthorityHaving Jurisdiction may impose additional testing and reporting requirements notcovered in this Appendix. Examples of testing guidelines or requirements deve-loped by other organizations or agencies include those published by SAC (FEMA,2000a; SAC, 1997), by the ICBO Evaluation Service (ICBO Evaluation Service,1997), and by the County of Los Angeles (County of Los Angeles Department ofPublic Works, 1996). Prior to developing a test program, the appropriate AuthorityHaving Jurisdiction should be consulted to ensure the test program meets all appli-cable requirements. Even when not required, the designer may find the informationcontained in the foregoing references to be useful resources in developing a testprogram.

CS3. DEFINITIONS

Inelastic Rotation. One of the key parameters measured in a connection test isthe inelastic rotation that can be developed in the specimen. Previously in theseProvisions, inelastic rotation was the primary acceptance criterion for beam-to-column moment connections in Moment Frames. The acceptance criterion in theProvisions is now based on Interstory Drift Angle, which includes both elastic andinelastic rotations. However, inelastic rotation provides an important indication ofconnection performance in earthquakes and should still be measured and reportedin connection tests. Researchers have used a variety of different definitions forinelastic rotation of moment connection test specimens in the past, making com-parison among tests difficult. In order to promote consistency in how test results arereported, these Provisions require that inelastic rotation for moment connection testspecimens be computed based on the assumption that all inelastic deformation of atest specimen is concentrated at a single point at the intersection of the centerline ofthe beam with the centerline of the column. With this definition, inelastic rotationis equal to the inelastic portion of the Interstory Drift Angle. Previously the Provi-sions defined inelastic rotation of moment connection specimens with respect to theface of the column. The definition has been changed to the centerline of the columnto be consistent with recommendations of SAC (SAC, 1997; FEMA, 2000a).

For tests of Link-to-column connections, the key acceptance parameter is the Linkinelastic rotation, also referred to in these Provisions as the Link Rotation Angle.The Link Rotation Angle is computed based upon an analysis of test specimendeformations, and can normally be computed as the inelastic portion of the rela-tive end displacement between the ends of the Link, divided by the Link length.Examples of such calculations can be found in Kasai and Popov (1986c), Riclesand Popov (1987), Engelhardt and Popov (1989a), and Arce (2002).

Interstory Drift Angle. The Interstory Drift Angle developed by a moment con-nection test specimen is the primary acceptance criterion for a beam-to-columnmoment connection in a Moment Frame. In an actual building, the Interstory DriftAngle is computed as the interstory displacement divided by the story height, and




142 PART I – ESSENTIAL TEST VARIABLES [Comm. CS5.

includes both elastic and inelastic components of deformation. For a test specimen,Interstory Drift Angle can usually be computed in a straightforward manner fromdisplacement measurements on the test specimen. Guidelines for computing theInterstory Drift Angle of a connection test specimen are provided by SAC (1997).

CS4. TEST SUBASSEMBLAGE REQUIREMENTS

A variety of different types of subassemblages and test specimens have been usedfor testing moment connections. A typical subassemblage is planar and consistsof a single column with a beam attached on one or both sides of the column. Thespecimen can be loaded by displacing either the end of the beam(s) or the end ofthe column. Examples of typical subassemblages for moment connections can befound in the literature, for example in SAC (1996) and Popov et al. (1996).

In these Provisions, test specimens generally need not include a composite slabor the application of axial load to the column. However, such effects may have aninfluence on connection performance, and their inclusion in a test program shouldbe considered as a means to obtain more realistic test conditions. An exampleof test subassemblages that include composite floor slabs and/or the applicationof column axial loads can be found in Popov et al. (1996), Leon, Hajjar, andShield (1997), and Tremblay, Tchebotarev, and Filiatrault (1997). A variety ofother types of subassemblages may be appropriate to simulate specific projectconditions, such as a specimen with beams attached in orthogonal directions toa column. A planar bare steel specimen with a single column and a single beamrepresents the minimum acceptable subassemblage for a moment connection test.However, more extensive and realistic subassemblages that better match actualproject conditions should be considered where appropriate and practical, in orderto obtain more reliable test results.

Examples of subassemblages used to test Link-to-column connections can be foundin Hjelmstad and Popov (1983), Kasai and Popov (1986c), Ricles and Popov(1987b), Engelhardt and Popov (1989a), Dusicka and Itani (2002), McDaniel et al.(2002), and Arce (2002).

CS5. ESSENTIAL TEST VARIABLES

CS5.1. Sources of Inelastic Rotation

This section is intended to ensure that the inelastic rotation in the test specimenis developed in the same members and connection elements as anticipated in theprototype. For example, if the prototype moment connection is designed so thatessentially all of the inelastic rotation is developed by yielding of the beam, then thetest specimen should be designed and perform in the same way. A test specimenthat develops nearly all of its inelastic rotation through yielding of the columnPanel Zone would not be acceptable to qualify a prototype connection whereinflexural yielding of the beam is expected to be the predominant inelastic action.

Because of normal variations in material properties, the actual location of inelasticaction may vary somewhat from that intended in either the test specimen or in theprototype. Consequently, by requiring that only 75 percent of the inelastic rotationoccur in the intended elements of the test specimen, some allowance is made forsuch variations. Thus, for the example above where essentially all of the inelastic




Comm. CS5.] PART I – ESSENTIAL TEST VARIABLES 143

rotation in the prototype is expected to be developed by flexural yielding of thebeam, at least 75 percent of the total inelastic rotation of the test specimen isrequired to be developed by flexural yielding of the beam in order to qualify thisconnection.

For many types of connections, yielding or inelastic deformations may occur inmore than a single member or connection element. For example, in some typesof moment connections, yielding may occur within the beam, within the columnPanel Zone, or within both the beam and Panel Zone. The actual distribution ofyielding between the beam and Panel Zone may vary depending upon the beamand column dimensions, web doubler plate thickness, and on the actual yieldstress of the beam, column and web doubler plate. Such a connection design canbe qualified by running two series of tests: one in which at least 75 percent ofthe inelastic rotation is developed by beam yielding; and a second in which atleast 75 percent of the inelastic rotation is developed by Panel Zone yielding. Theconnection design would then be qualified for any distribution of yielding betweenthe beam and the Panel Zone in the prototype.

For Link-to-column connections in EBF, the type of yielding (shear yielding,flexural yielding, or a combination of shear and flexural yielding) expected in thetest specimen Link should be substantially the same as for the Prototype Link.For example, a Link-to-column connection detail which performs satisfactorilyfor a shear-yielding Link (e ≤ 1.6Mp/Vp) may not necessarily perform well for aflexural-yielding Link (e ≥ 2.6Mp/Vp). The load and deformation demands at theLink-to-column connection will differ significantly for these cases.

Satisfying the requirements of this section will require the designer to have aclear understanding of the manner in which inelastic rotation is developed in thePrototype and in the test specimen.

CS5.2. Size of Members

The intent of this section is that the member sizes used in a test specimen should be,as nearly as practical, a full-scale representation of the member sizes used in theprototype. The purpose of this requirement is to ensure that any potentially adversescale effects are adequately represented in the test specimen. As beams becomedeeper and heavier, their ability to develop inelastic rotation may be somewhat di-minished (Roeder and Foutch, 1996; Blodgett, 1995). Although such scale effectsare not yet completely understood, at least two possible detrimental scale effectshave been identified. First, as a beam gets deeper, larger inelastic strains are gen-erally required in order to develop the same level of inelastic rotation. Second, theinherent restraint associated with joining thicker materials can affect joint and con-nection performance. Because of such potentially adverse scale effects, the beamsizes used in test specimens are required to adhere to the limits given in this section.

This section only specifies restrictions on the degree to which test results can bescaled up to deeper or heavier members. There are no restrictions on the degree towhich test results can be scaled down to shallower or lighter members. No suchrestrictions have been imposed in order to avoid excessive testing requirementsand because currently available evidence suggests that adverse scale effects aremore likely to occur when scaling up test results rather than when scaling down.Nonetheless, caution is advised when using test results on very deep or heavy




144 PART I – ESSENTIAL TEST VARIABLES [Comm. CS5.

members to qualify connections for much smaller or lighter members. It is prefer-able to obtain test results using member sizes that are a realistic representation ofthe prototype member sizes.

As an example of applying the requirements of this section, consider a momentconnection test specimen constructed with a W36 × 150 beam. This specimencould be used to qualify any beam with a depth up to 40 in. (= 36/0.9) and aweight up to 200 lbs/ft (= 150/0.75). The limits specified in this section have beenchosen somewhat arbitrarily based on judgment, as no quantitative research resultsare available on scale effects.

When choosing a beam size for a test specimen, several other factors should beconsidered in addition to the depth and weight of the section. One of these factorsis the width-thickness (b/t) ratio of the beam flange and web. The b/t ratios ofthe beam may have an important influence on the performance of specimens thatdevelop plastic rotation by flexural yielding of the beam. Beams with high b/tratios develop local buckling at lower inelastic rotation levels than beams withlow b/t ratios. This local buckling causes strength degradation in the beam, andmay therefore reduce the load demands on the connection. A beam with very lowb/t ratios may experience little if any local buckling, and will therefore subjectthe connection to higher moments. On the other hand, the beam with high b/tratios will experience highly localized deformations at locations of flange and webbuckling, which may in turn initiate a fracture. Consequently, it is desirable to testbeams over a range of b/t ratios in order to evaluate these effects.

In the previous edition of these Provisions, no specific restrictions were placed onthe size of columns used in test specimens in order to avoid excessively burden-some testing requirements. However, restrictions have now been added to theseProvisions requiring that the depth of the test column be at least 90 percent ofthe depth of the Prototype column. Tests conducted as part of the SAC programindicated that performance of connections with deep columns may differ from theperformance with W12 and W14 columns (Chi and Uang, 2002). In addition toadhering to this depth restriction, the column size should also be chosen to produceinelastic deformation in the appropriate elements of the specimen, according tothe requirements of Section S5.1.

CS5.5. Material Strength

The actual yield stress of structural steel can be considerably greater than its spec-ified minimum value. Higher levels of actual yield stress in members that supplyinelastic rotation by yielding can be detrimental to connection performance bydeveloping larger forces at the connection prior to yielding. For example, considera moment connection design in which inelastic rotation is developed by yieldingof the beam, and the beam has been specified to be of ASTM A36/A36M steel. Ifthe beam has an actual yield stress of 55 ksi (380 MPa), the connection is requiredto resist a moment that is 50 percent higher than if the beam had an actual yieldstress of 36 ksi (250 MPa). Consequently, this section requires that the materialsused for the test specimen represent this possible overstrength condition, as thiswill provide for the most severe test of the connection.

As an example of applying these provisions, consider again a test specimen inwhich inelastic rotation is intended to be developed by yielding of the beam. In




Comm. CS6.] PART I – LOADING HISTORY 145

order to qualify this connection for ASTM A36/A36M beams, the test beam isrequired to have a yield stress of at least 46 ksi (317 MPa) (= 0.85Ry Fy for ASTMA36/A36M). This minimum yield stress is required to be exhibited by both theweb and flanges of the test beam.

The requirements of this section are applicable only to members or connection ele-ments of the test specimen that are intended to contribute to the inelastic rotation ofthe specimen through yielding. The requirements of this section are not applicableto members or connection elements that are intended to remain essentially elastic.

CS5.6. Welds

The intent of these Provisions is to ensure that the welds on the test specimenreplicate the welds on the prototype as closely as practicable. Accordingly, it isrequired that the welding parameters, such as current and voltage, be within therange established by the weld metal manufacturer. Other essential variables, suchas steel grade, type of joint, root opening, included angle and preheat level, arerequired to be in accordance with AWS D1.1. It is not the intent of this section thatthe electrodes used to make welds in a test specimen must necessarily be the sameAWS classification, diameter or brand as the electrodes to be used on the Prototype.

CS6. LOADING HISTORY

The loading sequence prescribed in Section S6.2 for beam-to-column momentconnections is taken from SAC/BD-97/02, “Protocol for Fabrication, Inspection,Testing, and Documentation of Beam-to-Column Connection Tests and Other Ex-perimental Specimens” (SAC, 1997). This document should be consulted for fur-ther details of the loading sequence, as well as for further useful information ontesting procedures. The prescribed loading sequence is not intended to representthe demands presented by a particular earthquake ground motion. This loading se-quence was developed based on a series of non-linear time history analyses of steelMoment Frame structures subjected to a range of seismic inputs. The maximum de-formation, as well as the cumulative deformation and dissipated energy sustainedby beam-to-column connections in these analyses, were considered when estab-lishing the prescribed loading sequence and the connection acceptance criteria. If adesigner conducts a non-linear time history analysis of a Moment Frame structurein order to evaluate demands on the beam-to-column connections, considerablejudgment will be needed when comparing the demands on the connection predictedby the analysis with the demands placed on a connection test specimen using theprescribed loading sequence. In general, however, a connection can be expectedto provide satisfactory performance if the cumulative plastic deformation, and thetotal dissipated energy sustained by the test specimen prior to failure are equalor greater to the same quantities predicted by a non-linear time-history analysis.When evaluating the cumulative plastic deformation, both total rotation (elasticplus inelastic) as well as inelastic rotation at the connection should be considered.SAC/BD-00/10 (SAC, 2000) can be consulted for further information on this topic.

The loading sequence specified in SAC/BD-97/02 was specifically developed forconnections in Moment Frames, and may not be appropriate for testing of Link-to-column connections in EBF. Inelastic deformation of EBF generally initiates atmuch lower Interstory Drift Angles than in Moment Frames. The loading cycles




146 PART I – MATERIALS TESTING REQUIREMENTS [Comm. CS8.

prescribed for Moment Frame connection test specimens at Interstory Drift Anglesless than 0.01 radian will generally be in the elastic range. However, for typicalEBF, yielding in the Links may initiate at Interstory Drift Angles less than 0.00375radian. Consequently, using the loading sequence prescribed for Moment Framesmay result in an excessive number of inelastic loading cycles for an EBF test spec-imen. Further, the relationship between Interstory Drift Angle and Link RotationAngle in an EBF is dependent on the frame geometry. Consequently, prescribing aloading history for Link-to-column connection tests based on the Interstory DriftAngle may lead to inconsistent test results. Since acceptance criteria for Link-to-column connections are based on the Link Rotation Angle, then a prescribedloading history based on the Link Rotation Angle will provide for more consis-tent test results. No standard loading sequence has been developed for testing ofLink-to-column connections. The loading sequence prescribed in Section S6.3 waschosen based on judgment and a review of typical loading sequences used in pastEBF testing. The Link Rotation Angle specified in Section S6.3 for controllingtests on Link-to-column connection specimens is intended to be the total LinkRotation Angle, including both elastic and inelastic components of deformation.This can usually be computed by taking the total relative end displacement of theLink and dividing by the Link length. While test control is based on total LinkRotation Angle, the acceptance criteria for Link-to-column connections are basedon the inelastic Link Rotation Angle.

The loading sequence specified in ATC-24, “Guidelines for Cyclic Seismic Test-ing of Components of Steel Structures,” (Applied Technology Council, 1992) isconsidered as an acceptable alternative to those prescribed in Sections S6.2 andS6.3. Further, any other loading sequence may be used for beam-to-column mo-ment connections or Link-to-column connections, as long as the loading sequenceis equivalent or more severe than those prescribed in Sections S6.2 and S6.3. Tobe considered as equivalent or more severe, alternative loading sequences shouldmeet the following requirements: (1) the number of inelastic loading cycles shouldbe at least as large as the number of inelastic loading cycles resulting from theprescribed loading sequence; and (2) the cumulative plastic deformation should beat least as large as the cumulative plastic deformation resulting from the prescribedloading sequence.

Dynamically applied loads are not required in these Provisions. The use of slowlyapplied cyclic loads, as typically reported in the literature for connection tests,are acceptable for the purposes of these Provisions. It is recognized that dynamicloading can considerably increase the cost of testing, and that few laboratory fa-cilities have the capability to dynamically load very large-scale test specimens.Furthermore, the available research on dynamic loading effects on steel connec-tions has not demonstrated a compelling need for dynamic testing. Nonetheless,applying the required loading sequence dynamically, using loading rates typicalof actual earthquake loading, will likely provide a better indication of the expectedperformance of the connection, and should be considered where possible.

CS8. MATERIALS TESTING REQUIREMENTS

Tension testing is required for members and connection elements of the test speci-men that contribute to the inelastic rotation of the specimen by yielding. These testsare required to demonstrate conformance with the requirements of Section S5.5,




Comm. CS10.] PART I – ACCEPTANCE CRITERIA 147

and to permit proper analysis of test specimen response. Tension test results re-ported on certified mill test reports are not permitted to be used for this purpose.Yield stress values reported on a certified mill test report may not adequately repre-sent the actual yield strength of the test specimen members. Variations are possibledue to material sampling locations and tension test methods used for certified milltest reports.

ASTM standards for tension testing permit the reported yield stress to be taken asthe upper yield point. However, for steel members subject to large cyclic inelasticstrains, the upper yield point can provide a misleading representation of the actualmaterial behavior. Thus, while an upper yield point is permitted by ASTM, it isnot permitted for the purposes of this Section. Determination of yield stress usingthe 0.2 percent strain offset method is required in this Appendix.

Only tension tests are required in this section. Additional materials testing, how-ever, can sometimes be a valuable aid for interpreting and extrapolating test results.Examples of additional tests, which may be useful in certain cases, include CharpyV-Notch tests, hardness tests, chemical analysis, and others. Consideration shouldbe given to additional materials testing, where appropriate.

CS10. ACCEPTANCE CRITERIA

A minimum of two tests is required for each condition in the prototype in whichthe variables listed in Section S5 remain unchanged. The designer is cautioned,however, that two tests, in general, cannot provide a thorough assessment of thecapabilities, limitations, and reliability of a connection. Thus, where possible, it ishighly desirable to obtain additional test data to permit a better evaluation of theexpected response of a connection to earthquake loading. Further, when evaluatingthe suitability of a proposed connection, it is advisable to consider a broader rangeof issues other than just inelastic rotation capacity.

One factor to consider is the controlling failure mode after the required inelasticrotation has been achieved. For example, a connection that slowly deteriorates instrength due to local buckling may be preferable to a connection that exhibits amore brittle failure mode such as fracture of a weld, fracture of a beam flange, etc.,even though both connections achieved the required inelastic rotation.

In addition, the designer should also carefully consider the implications of un-successful tests. For example, consider a situation where five tests were run on aparticular type of connection, two tests successfully met the acceptance criteria, butthe other three failed prematurely. This connection could presumably be qualifiedunder these Provisions, since two successful tests are required. Clearly, however,the number of failed tests indicates potential problems with the reliability of theconnection. On the other hand, the failure of a tested connection in the laboratoryshould not, by itself, eliminate that connection from further consideration. As longas the causes of the failure are understood and corrected, and the connection issuccessfully retested, the connection may be quite acceptable. Thus, while theacceptance criteria in these Provisions have intentionally been kept simple, thechoice of a safe, reliable and economical connection still requires considerablejudgment.




148

APPENDIX X

WELD METAL/WELDING PROCEDURE SPECIFICATIONTOUGHNESS VERIFICATION TEST

PART I

FEMA (FEMA, 2000b) first published this procedure for qualifying weld metals to meetthe recommended Charpy V-Notch (CVN) requirements. The procedure and test tempera-tures vary from the existing AWS requirements. Since publication of the FEMA document,the filler metal manufacturers have been conducting tests and are certifying those mate-rials that meet this requirement. As stated, this Appendix is included on an interim basispending adoption of such a procedure by AWS or other accredited organization.

All component tests conducted in the SAC Project were conducted at room temperature,approximately 70˚F (21˚C), at which it was determined that an adequate CVN toughnessis 40 ft-lbf (54 J). The lowest anticipated operating temperature of most buildings is50˚F (10˚C). Considering the difference in loading rates between seismic motions andCVN testing, and the temperature increase of weldments under seismic loads, the CVNrequirement of 40 ft-lbf (54 J) at 70˚F (21˚C) should be adequate for use at 50˚F (10˚C).This additional weld toughness requirement is stipulated in Section 7.3b for CJP welds inSpecial and Intermediate Moment Frames.

During the SAC study (see FEMA 355B, section 2.3.3.5 (FEMA, 2000d)), it was deemedimportant to actually verify this expected higher toughness for performance at the expectedservice temperature. Appendix X testing requirements for 40 ft-lbf (54 J) at 70˚F (21˚C)are intended to verify that at most common service temperatures, the minimum toughnessis maintained to provide satisfactory performance in seismic joints.

For service temperatures below 50˚F (10˚C), it is reasonable to require a 40 ft-lbf (54 J)minimum CVN toughness at 20˚F (11˚C) above the lowest anticipated service temperatureof the building.

Filler metal classification testing is governed by the AWS A5 specifications, which requirespecific tests on the weld metal, deposited using prescribed electrode diameters withprescribed welding conditions. Actual production welding may be done with electrodes ofdifferent diameters and under considerably different welding variables (amperage, voltage,travel speed, electrode extension, position, plate thickness, joint geometry, preheat andinterpass temperatures, shielding gas type and flow rate, for example). Such variables mayconsiderably affect the actual tensile and CVN properties achieved in production welds.Although the requirement of 7.3a is that all filler metals be classified under AWS A5tests for a minimum of 20 ft-lbf at minus 20˚F (27 J at minus 29˚C) ensures that someminimum level of notch toughness will be provided, there is no guarantee that 40 ft-lbf(54 J) at 70˚F (21˚C) CVN will be achieved under either the A5 prescribed conditions, northe wide variety of possible welding procedures and cooling rates. For the critical weldslisted in Section 7.3b, additional testing is required to verify that the production weld will




PART I – WELD METAL / WELDING PROCEDURE 149

achieve the required higher level of CVN toughness, under conditions similar to those tobe encountered in production.

The mechanical properties of welds are affected by the cooling rates experienced by thewelds. Cooling rates are in turn affected by the welding heat input, calculated as 0.06 timesarc voltage times welding current, divided by travel speed in inches per minute. Testing ofwelds is required using high heat input levels, and low heat input levels. By testing usingbracketed heat inputs, requiring that production welding procedures fall within these testedheat inputs, and by testing the actual electrode diameter and production lot to be used inproduction, there is greater confidence that the as-deposited weld metal will provide therequired level of CVN toughness.




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151PART II. COMPOSITE STRUCTURAL STEELAND REINFORCED CONCRETE BUILDINGS

C1. SCOPE

These Provisions for the seismic design of composite structural steel and reinforcedconcrete buildings are based upon the 1994 NEHRP Provisions (FEMA, 1994) andsubsequent modifications made in the 1997 NEHRP Provisions (FEMA, 1997a).Since composite systems are assemblies of steel and concrete components, Part Iof these Provisions, the LRFD Specification (AISC, 1999) and ACI 318 (ACI,2002), form an important basis for Part II.

The available research demonstrates that properly detailed composite membersand connections can perform reliably when subjected to seismic ground motions.However, there is at present limited experience with composite building systemssubjected to extreme seismic loads and many of the recommendations herein arenecessarily of a conservative and/or qualitative nature. Careful attention to allaspects of the design is necessary, particularly the general building layout and de-tailing of members and connections. Composite connection details are illustratedthroughout this Commentary to convey the basic character of the composite sys-tems. However, these details should not necessarily be treated as design standardsand the cited references provide more specific information on the design of com-posite connections. Additionally, refer to Viest, Colaco, Furlong, Griffis, Leon,and Wylie (1997).

The design and construction of composite elements and systems continues toevolve in practice. With further experience and research, it is expected that theseProvisions can be better quantified, refined and expanded. Nonetheless, these Pro-visions are not intended to limit the application of new systems, except whereexplicitly stated, for which testing and analysis demonstrates that the structure hasadequate strength, ductility, and toughness.

It is generally anticipated that the overall behavior of the composite systems hereinwill be similar to that for counterpart structural steel systems or reinforced concretesystems and that inelastic deformations will occur in conventional ways, suchas flexural yielding of beams in FR Moment Frames or axial yielding and/orbuckling of braces in Braced Frames. However, differential stiffness between steeland concrete elements is more significant in the calculation of internal forcesand deformations of composite systems than for structural steel only or reinforcedconcrete only systems. For example, deformations in reinforced concrete elementscan vary considerably due to the effects of cracking.

When systems have both ductile and non-ductile elements, the relative stiffnessof each should be properly modeled; the ductile elements can deform inelasti-cally while the non-ductile elements remain nominally elastic. When using elasticanalysis, member stiffness should be reduced to account for the degree of crack-ing at the onset of significant yielding in the structure. Additionally, it is nec-essary to account for material overstrength that may alter relative strength andstiffness.


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152 PART II – MATERIALS [Comm. C5.

C2. REFERENCED SPECIFICATIONS, CODES, AND STANDARDS

The specifications, codes and standards that are referenced in Part II are listed withthe appropriate revision date that was used in the development of Part II, exceptthose that are already listed in Part I.

C3. SEISMIC DESIGN CATEGORIES

See Part I Commentary Section C3.

C4. LOADS, LOAD COMBINATIONS, AND NOMINAL STRENGTHS

In general, requirements for loads and load combinations for composite structuresare similar to those described in Part I Section C4. Specific seismic design, loadingcriteria and usage limitations for composite structures are specified in the 2000NEHRP Provisions (FEMA, 2000g).

The calculation of seismic loads for composite systems per the 2000 NEHRPProvisions (FEMA, 2000g) is the same as is described for steel structures in PartI Commentary Section C4. Table C-II-4.1 lists the seismic response modificationfactors R and Cd for the 2000 NEHRP Provisions (FEMA, 2000g). The values inTable C-II-4.1 are predicated upon meeting the design and detailing requirementsfor the composite systems specified in these Provisions. Overstrength factors forthe composite systems given in Table II-4-1 of these Provisions are the same asthose specified in the 2000 NEHRP Provisions (FEMA, 2000g).

ACI 318 Appendix C has been included by reference to facilitate the proportioningof building structures that include members made of steel and concrete. Whenreinforced concrete members are proportioned using the minimum design loadsstipulated in LRFD Specification Section A4.1, which is consistent with those inASCE 7 (ASCE, 1998), the strength reduction factors in ACI 318 Appendix Cshould be used in lieu of those in ACI 318 Chapter 9.

The seismic response modification factors R and Cd for composite systems spec-ified by the 2000 NEHRP Provisions (FEMA, 2000g) are similar to those forcomparable systems of steel and reinforced concrete. This is based on the fact that,when carefully designed and detailed according to these Provisions, the overallinelastic response for composite systems should be similar to comparable steel andreinforced concrete systems. Therefore, where specific loading requirements arenot specified in the Applicable Building Code for composite systems, appropriatevalues for the seismic response factors can be inferred from specified values forsteel and/or reinforced concrete systems.

C5. MATERIALS

The limitations in Section 5.1 on structural steel grades used with Part II require-ments are the same as those given in Part I. The limitations in Section 5.2 onspecified concrete compressive strength in composite members are the same asthose given in LRFD Specification Chapter I and ACI 318 Chapter 21. Whilethese limitations are particularly appropriate for construction in Seismic DesignCategories D and higher, they apply in any Seismic Design Category when systemsare designed with the assumption that inelastic deformation will be required.


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Comm. C6.] PART II – COMPOSITE MEMBERS 153

TABLE C-II-4.1Design Factors for Composite Systems

BASIC STRUCTURAL SYSTEM AND R CdSEISMIC LOAD RESISTING SYSTEM

Systems designed and detailed to meet the requirements ofboth the LRFD Specification and Part I:

Braced Frame Systems:Composite Concentrically Braced Frame (C-CBF) 5 41/2

Composite Ordinary Braced Frames (C-OBF) 3 3Composite Eccentrically Braced Frames (C-EBF) 8 4

Shear Wall Systems:Composite Steel Plate Shear Walls (C-SPW) 61/2 51/2

Special Reinforced Concrete Shear WallsComposite with Steel Elements (C-SRCW) 6 5

Ordinary Reinforced Concrete Shear WallsComposite with Steel Elements (C-ORCW) 5 41/2

Moment Frame Systems:Composite Special Moment Frames (C-SMF) 8 51/2

Composite Intermediate Moment Frames (C-IMF) 5 41/2

Composite Partially Restrained Moment Frame (C-PRMF) 6 51/2

Composite Ordinary Moment Frames (C-OMF) 3 21/2

Dual Systems with SMF capable of resisting 25 percent of V:Composite Concentrically Braced Frames (C-CBF) 6 5Composite Eccentrically Braced Frames (C-EBF) 8 4Composite Steel Plate Shear Walls (C-SPW) 8 61/2

Special Reinforced Concrete Shear WallsComposite with Steel Elements (C-SRCW) 8 61/2

Ordinary Reinforced Concrete Shear WallsComposite with Steel Elements (C-ORCW) 7 6

Dual Systems with IMF capable of resisting 25 percent of V:Composite Concentrically Braced Frame (C-CBF) 5 41/2

Composite Ordinary Braced Frame (C-OBF) 4 3Ordinary Reinforced Concrete Shear Walls

Composite with Steel Elements (C-ORCW) 51/2 41/2

C6. COMPOSITE MEMBERS

C6.1. Scope

These Provisions address the seismic design requirements that should be appliedin addition to the basic design requirements for gravity and wind loading.

C6.2. Composite Floor and Roof Slabs

In composite construction, floor and roof slabs typically consist of either compositeor non-composite metal deck slabs that are connected to the structural framing toprovide an in-plane composite diaphragm that collects and distributes seismicloads. Generally, composite action is distinguished from non-composite action onthe basis of the out-of-plane shear and flexural behavior and design assumptions.

Composite metal deck slabs are those for which the concrete fill and metal deckwork together to resist out-of-plane bending and out-of-plane shear. Flexuralstrength design procedures and codes of practice for such slabs are well estab-lished (ASCE, 2002; ASCE, 1991a and 1991b; AISI, 1996; SDI, 1993).


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154 PART II – COMPOSITE MEMBERS [Comm. C6.

Non-composite metal deck slabs are one-way or two-way reinforced concreteslabs for which the metal deck acts as formwork during construction, but is notrelied upon for composite action. Non-composite metal deck slabs, particularlythose used as roofs, can be formed with metal deck and overlaid with insulatingconcrete fill that is not relied upon for out-of-plane strength and stiffness. Whetheror not the slab is designed for composite out-of-plane action, the concrete fillinhibits buckling of the metal deck, increasing the in-plane strength and stiffnessof the diaphragm over that of the bare steel deck.

The diaphragm should be designed to collect and distribute seismic loads to theSeismic Load Resisting System. In some cases, loads from other floors should alsobe included, such as at a level where a change in the structural stiffness results ina redistribution. Recommended diaphragm (in-plane) shear strength and stiffnessvalues for metal deck and composite diaphragms are available for design fromindustry sources that are based upon tests and recommended by the AuthorityHaving Jurisdiction (SDI, 1991; SDI, 2001). In addition, research on compositediaphragms has been reported in the literature (Easterling and Porter, 1994).

As the thickness of concrete over the steel deck is increased, the shear strengthcan approach that for a concrete slab of the same thickness. For example, incomposite floor deck diaphragms having cover depths between 2 in. (51 mm) and6 in. (152 mm), measured shear stresses on the order of 3.5

√f ′c (where

√f ′c and

f ′c are in units of psi) have been reported. In such cases, the diaphragm strength

of concrete metal deck slabs can be conservatively based on the principles ofreinforced concrete design (ACI, 2002) using the concrete and reinforcement abovethe metal deck ribs and ignoring the beneficial effect of the concrete in the flutes.

Shear forces are transferred through welds and/or shear devices in the collectorand boundary elements. Fasteners between the diaphragm and the steel framingshould be capable of transferring forces using either welds or shear devices. Whereconcrete fill is present, it is generally advisable to use mechanical devices suchas headed shear stud connectors to transfer diaphragm forces between the slaband collector/boundary elements, particularly in complex shaped diaphragms withdiscontinuities. However, in low-rise buildings without abrupt discontinuities inthe shape of the diaphragms or in the Seismic Load Resisting System, the standardmetal deck attachment procedures may be acceptable.

C6.3. Composite Beams

These provisions apply only to composite beams that are part of the Seismic LoadResisting System.

When the design of a composite beam satisfies Equation 6-1, the strain in thesteel at the extreme fiber will be at least five times the tensile yield strain prior toconcrete crushing at strain equal to 0.003. It is expected that this ductility limitwill control the beam geometry only in extreme beam/slab proportions.

While these Provisions permit the design of composite beams based solely uponthe requirements in the LRFD Specification, the effects of reversed cyclic load-ing on the strength and stiffness of shear studs should be considered. This isparticularly important for C-SMF where the design loads are calculated assuming


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large member ductility and toughness. In the absence of test data to support spe-cific requirements in these Provisions, the following special measures should beconsidered in C-SMF: (1) implementation of an inspection and quality assuranceplan to insure proper welding of shear stud connectors to the beams; and (2) use ofadditional shear stud connectors beyond those required in the LRFD Specificationimmediately adjacent to regions of the beams where plastic hinging is expected.

C6.4. Reinforced-concrete-encased Composite Columns

The basic requirements and limitations for determining the Design Strength ofencased composite columns are the same as those in the LRFD Specification.Additional requirements for reinforcing bar details of composite columns that arenot covered in the LRFD Specification are included based on provisions in ACI 318.

Composite columns can be an ideal solution for use in seismic regions becauseof their inherent structural redundancy. For example, if a composite column isdesigned such that the structural steel can carry most or all of the dead load actingalone, then an extra degree of protection and safety is afforded, even in a severeearthquake where excursions into the inelastic range can be expected to deteriorateconcrete cover and buckle reinforcing steel. However, as with any column ofconcrete and reinforcement, the designer should be aware of the constructabilityconcerns with the placement of reinforcement and potential for congestion. Thisis particularly true at beam-to-column connections where potential interferencebetween a steel spandrel beam, a perpendicular floor beam, vertical bars, joint ties,and shear stud connectors can cause difficulty in reinforcing bar placement and apotential for honeycombing of the concrete.

Seismic detailing requirements for composite columns are specified in the fol-lowing three categories: ordinary, intermediate, and special. The required levelof detailing is specified in these Provisions for seismic systems in Sections 8through 17. The ordinary detailing requirements of Section 6.4a are intended asbasic requirements for all cases. Intermediate requirements are intended for seis-mic systems permitted in Seismic Design Category C, and special requirementsare intended for seismic systems permitted in Seismic Design Categories D andabove.

C6.4a. Ordinary Seismic System Requirements

These requirements are intended to supplement the basic requirements of the LRFDSpecification for encased composite columns in all Seismic Design Categories.

(1) Specific instructions are given for the determination of the nominal shearstrength in concrete encased steel composite members including assignmentof some shear to the reinforced concrete encasement. Examples for deter-mining the effective shear width bw of the reinforced concrete encasementare illustrated in Figure C-II-6.1. These provisions exclude any strength Vcassigned to concrete alone (Furlong, 1997).

(2) Currently no existing specification in the United States includes requirementsfor shear connectors for encased steel sections. The provisions in this subsec-tion require that shear connectors be provided to transfer all calculated axial


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156 PART II – COMPOSITE MEMBERS [Comm. C6.

Fig. C-II-6.1. Effective widths for shear strength calculation of encased composite columns.

forces between the structural steel and the concrete, neglecting the contributionof bond and friction. Friction between the structural steel and concrete is as-sumed to transfer the longitudinal shear stresses required to develop the plasticbending strength of the cross section. However, minimum shear studs shouldbe provided according to the maximum spacing limit of 16 inches. Furtherinformation regarding the design of shear connectors for encased members isavailable (Furlong, 1997; Griffis, 1992a and 1992b).

(3) The tie requirements in this section are essentially the same as those for com-posite columns in ACI 318 Chapter 10.

(4) The requirements for longitudinal bars are essentially the same as those thatapply to composite columns for low- and non-seismic design as specified inACI 318. The distinction between load-carrying and restraining bars is madeto allow for longitudinal bars (restraining bars) that are provided solely forerection purposes and to improve confinement of the concrete. Due to inter-ference with steel beams framing into the encased members, the restrainingbars are often discontinuous at floor levels and, therefore, are not included indetermining the column strength.

(5) The requirements for the steel core are essentially the same as those for com-posite columns as specified in the LRFD Specification and ACI 318. In addi-tion, earthquake damage to encased composite columns in Japan (Azizinaminiand Ghosh, 1996) highlights the need to consider the effects of abrupt changesin stiffness and strength where encased composite columns transition into re-inforced concrete columns and/or concrete foundations.

C6.4b. Intermediate Seismic System Requirements

The more stringent tie spacing requirements for intermediate seismic systemsfollow those for reinforced concrete columns in regions of moderate seismicity asspecified in ACI 318 Chapter 21 (Section 21.8). These requirements are appliedto all composite columns for systems permitted in Seismic Design Category Cto make the composite column details at least equivalent to the minimum levelof detailing for columns in Intermediate Moment Frames of reinforced concrete(FEMA, 2000e; ICC, 2000).


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C6.4c. Special Seismic System Requirements

The additional requirements for encased composite columns used in special seismicsystems are based upon comparable requirements for structural steel and reinforcedconcrete columns in systems permitted in Seismic Design Categories D and above(FEMA, 2000e; ICC, 2000). For additional explanation of these requirements, seethe Commentaries for Part I in these Provisions and ACI 318 Chapter 21.

The minimum tie area requirement in Equation 6-2 is based upon a similar provisionin ACI 318 Section 21.4.4, except that the required tie area is reduced to take intoaccount the steel core. The tie area requirement in Equation 6-2 and related tiedetailing provisions are waived if the steel core of the composite member can aloneresist the expected (arbitrary point in time) gravity load on the column becauseadditional confinement of the concrete is not necessary if the steel core can inhibitcollapse after an extreme seismic event. The load combination of 1.0D + 0.5L isbased upon a similar combination proposed as loading criteria for structural safetyunder fire conditions (Ellingwood and Corotis, 1991).

The requirements for composite columns in C-SMF are based upon similar re-quirements for steel and reinforced concrete columns in SMF (FEMA, 2000e;ICC, 2000). For additional commentaries, see Part I in these Provisions andASCE 7.

The strong-column/weak-beam (SC/WB) concept follows that used for steel andreinforced concrete columns in SMF. Where the formation of a plastic hinge at theColumn Base is likely or unavoidable, such as with a fixed base, the detailing shouldprovide for adequate plastic rotational ductility. For Seismic Design Category E,special details, such as steel jacketing of the Column Base, should be consideredto avoid spalling and crushing of the concrete.

Closed hoops are required to ensure that the concrete confinement and nominalshear strength are maintained under large inelastic deformations. The hoop de-tailing requirements are equivalent to those for reinforced concrete columns inSMF. The transverse reinforcement provisions are considered to be conservativesince composite columns generally will perform better than comparable reinforcedconcrete columns with similar confinement. However, further research is requiredto determine to what degree the transverse reinforcement requirements can be re-duced for composite columns. It should be recognized that the closed hoop andcross-tie requirements for C-SMF may require special details such as those sug-gested in Figure C-II-6.2 to facilitate the erection of the reinforcement around the

Fig. C-II-6.2. Example of a closed hoop detail for an encased composite column.


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158 PART II – COMPOSITE CONNECTIONS [Comm. C7.

steel core. Ties are required to be anchored into the confined core of the columnto provide effective confinement.

C6.5. Concrete-filled Composite Columns

The basic requirements and limitations for detailing and determining the DesignStrength of filled composite columns are the same as those in LRFD SpecificationChapter I. The limit of As/Ag ≥ 0.04 is the same as that in the LRFD Specificationand defines the limit of applicability of these Provisions. Although it is not intendedin these Provisions that filled composite columns with smaller steel area ratios beprohibited, alternative provisions are not currently available.

The shear strength of the filled member is conservatively limited to the nominalshear yield strength of the steel tube because the actual shear strength contributionof the concrete fill has not yet been determined in testing. This approach is rec-ommended until tests are conducted (Furlong, 1997; ECS, 1994). Even with thisconservative approach, shear strength rarely governs the design of typical filledcomposite columns with cross-sectional dimensions up to 30 in. (762 mm). Alter-natively, the shear strength for filled tubes can be determined in a manner that issimilar to that for reinforced concrete columns with the steel tube considered asshear reinforcement and its shear yielding strength neglected. However, given theupper limit on shear strength as a function of concrete crushing in ACI 318, thisapproach would only be advantageous for columns with low ratios of structuralsteel to concrete areas (Furlong, 1997).

The more stringent slenderness criteria for the wall thickness in square or rectan-gular HSS is based upon comparable requirements from Part I in these Provisionsfor unfilled HSS used in SMF. Comparing the provisions in the LRFD Specifica-tion and Part I in these Provisions, the width/thickness ratio for unfilled HSS inSMF is about 80 percent of those for OMF. This same ratio of 0.8 was appliedto the standard (non-seismic) b/t ratio for filled HSS in the LRFD Specification.The reduced slenderness criterion was imposed as a conservative measure untilfurther research data becomes available on the cyclic response of filled square andrectangular tubes. More stringent D/t ratio limits for circular pipes are not appliedas data are available to show the standard D/t ratio is sufficient for seismic design(Boyd, Cofer, and McLean, 1995; Schneider, 1998).

C7. COMPOSITE CONNECTIONS

C7.1. Scope

The use of composite connections often simplifies some of the special challengesassociated with traditional steel and concrete construction. For example, comparedto structural steel, composite connections often avoid or minimize the use of fieldwelding, and compared to reinforced concrete, there are fewer instances whereanchorage and development of primary beam reinforcement is a problem.

Given the many alternative configurations of composite structures and connec-tions, there are few standard details for connections in composite construction(Griffis, 1992b; Goel, 1992a; Goel, 1993). However, tests are available for severalconnection details that are suitable for seismic design. References are given inthis Section of the Commentary and Commentary Sections C8 to C17. In most


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Comm. C7.] PART II – COMPOSITE CONNECTIONS 159

composite structures built to date, engineers have designed connections using ba-sic mechanics, equilibrium, existing standards for steel and concrete construction,test data, and good judgment. The provisions in this Section are intended to helpstandardize and improve design practice by establishing basic behavioral assump-tions for developing design models that satisfy equilibrium of internal forces inthe connection for seismic design.

C7.2. General Requirements

The requirements for deformation capacity apply to both connections designedfor gravity load only and connections that are part of the Seismic Load ResistingSystem. The ductility requirement for gravity load only connections is intended toavoid failure in gravity connections that may have rotational restraint but limitedrotation capacity. For example, shown in Figure C-II-7.1 is a connection between areinforced concrete wall and steel beam that is designed to resist gravity loads andis not considered to be part of the Seismic Load Resisting System. However, thisconnection is required to be designed to maintain its vertical shear strength underrotations and/or moments that are imposed by inelastic seismic deformations ofthe structure.

In calculating the Required Strength of connections based on the Nominal Strengthof the connected members, allowance should be made for all components of themembers that may increase the Nominal Strength above that usually calculated indesign. For example, this may occur in beams where the negative moment strengthprovided by slab reinforcement is often neglected in design but will increase themoments applied through the beam-to-column connection. Another example is inconcrete-filled tubular braces where the increased tensile and compressive strengthof the brace due to concrete should be considered in determining the requiredconnection strength. Because the evaluation of such conditions is case specific,these provisions do not specify any allowances to account for overstrength. How-ever, as specified in Part I Section 6.2, calculations for the Required Strength of

Fig. C-II-7.1. Steel beam-to-RC wall gravity load shear connection.


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connections should, as a minimum, be made using the Expected Yield Strength ofthe connected steel member. Where connections resist forces imposed by yieldingof steel in reinforced concrete members, ACI 318 Section 21.5 implies an ExpectedYield Strength equal to 1.25Fy for reinforcing bars.

C7.3. Nominal Strength of Connections

In general, forces between structural steel and concrete will be transferred by acombination of bond, adhesion, friction and direct bearing. Transfers by bond andadhesion are not permitted for Nominal Strength calculation purposes because:(1) these mechanisms are not effective in transferring load under inelastic loadreversals; and (2) the effectiveness of the transfer is highly variable depending onthe surface conditions of the steel and shrinkage and consolidation of the concrete.

Transfer by friction shall be calculated using the shear friction provisions in ACI318 where the friction is provided by the clamping action of steel ties or studsor from compressive stresses under applied loads. Since the provisions for shearfriction in ACI 318 are based largely on monotonic tests, the values are reducedby 25 percent where large inelastic stress reversals are expected. This reduction isconsidered to be a conservative requirement that does not appear in ACI 318 butis applied herein due to the relative lack of experience with certain configurationsof composite structures.

In many composite connections, steel components are encased by concrete thatwill inhibit or fully prevent local buckling. For seismic deign where inelastic loadreversals are likely, concrete encasement will be effective only if it is properlyconfined. One method of confinement is with reinforcing bars that are fully an-chored into the confined core of the member (using requirements for hoops in ACI318 Chapter 21). Adequate confinement also may occur without special reinforce-ment where the concrete cover is very thick. The effectiveness of the latter type ofconfinement should be substantiated by tests.

For fully-encased connections between steel (or composite) beams and reinforcedconcrete (or composite) columns such as shown in Figure C-II-7.2, the PanelZone nominal shear strength can be calculated as the sum of contributions from

Fig. C-II-7.2. Reinforced concrete column-to-steel beam moment connection.


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Comm. C7.] PART II – COMPOSITE CONNECTIONS 161

Fig. C-II-7.3. Panel shear mechanisms in steel beam-to-reinforced concrete column connections(Deierlein et al., 1989)

the reinforced concrete and steel shear panels (see Figure C-II-7.3). This super-position of strengths for calculating the Panel Zone nominal shear strength isused in detailed design guidelines (Deierlein, Sheikh, and Yura, 1989; ASCE,1994; Parra-Montesinos and Wight, 2001) for composite connections that are sup-ported by test data (Sheikh, Deierlein, Yura, and Jirsa, 1989; Kanno and Deierlein,1997; Nishiyama, Hasegawa, and Yamanouchi, 1990; Parra-Montesinos and Wight,2001). Further information on the use and design of such connections is includedin Commentary Part II, Section C9.

Reinforcing bars in and around the joint region serve the dual functions of resist-ing calculated internal tension forces and providing confinement to the concrete.Internal tension forces can be calculated using established engineering models thatsatisfy equilibrium (e.g., classical beam-column theory, the truss analogy, strut andtie models). Tie requirements for confinement usually are based on empirical mod-els based on test data and past performance of structures (ACI, 1991; Kitayama,Otani, Aoyama, 1987).

(1) In connections such as those in C-PRMF, the force transfer between theconcrete slab and the steel column requires careful detailing. For C-PRMF


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Fig. C-II-7.4. Composite partially restrained connection.

connections (see Figure C-II-7.4), the strength of the concrete bearing againstthe column flange should be checked. Only the solid portion of the slab (areaabove the ribs) should be counted, and the nominal bearing strength should belimited to 1.2 f ′

c (Ammerman and Leon, 1990). In addition, because the forcetransfer implies the formation of a large compressive strut between the slabbars and the column flange, adequate transverse steel reinforcement shouldbe provided in the slab to form the tension tie. From equilibrium calculations,this amount should be the same as that provided as longitudinal reinforcementand should extend at least 12 in. (305 mm) beyond either side of the effectiveslab width.

(2) Due to the limited size of joints and the congestion of reinforcement, it of-ten is difficult to provide the reinforcing bar development lengths specifiedin ACI 318 for transverse column reinforcement in joints. Therefore, it isimportant to take into account the special requirements and recommenda-tions for tie requirements as specified for reinforced concrete connections inACI 318, Section 21.5 and in ACI (1991), Kitayama et al. (1987), Sheikh


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Comm. C8.] PART II – COMPOSITE PARTIALLY RESTRAINED MOMENT FRAMES 163

and Uzumeri (1980), Park, Priestley, and Gill (1982), and Saatcioglu (1991).Test data (Sheikh et al., 1989; Kanno and Deierlein, 1997; Nishiyama et al.,1990) on composite beam-to-column connections similar to the one shown inFigure C-II-7.2 indicate that the face bearing (stiffener) plates attached to thesteel beam provide effective concrete confinement.

(3) As in reinforced concrete connections, large bond stress transfer of loadsto column bars passing through beam-to-column connections can result inslippage of the bars under extreme loadings. Current practice for reinforcedconcrete connections is to control this slippage by limiting the maximumlongitudinal bar sizes as described in ACI (2002).

C8. COMPOSITE PARTIALLY RESTRAINED (PR)MOMENT FRAMES (C-PRMF)

Composite Partially Restrained (PR) frames consist of structural steel columnsand composite steel beams that are interconnected with PR composite connections(Zandonini and Leon, 1992). PR composite connections utilize traditional steelframe shear and bottom flange connections and the additional strength and stiffnessprovided by the floor slab has been incorporated by adding shear studs to the beamsand slab reinforcement in the negative moment regions adjacent to the columns(see Figure C-II-7.4). This results in a more favorable distribution of strength andstiffness between negative and positive moment regions of the beams and providesfor redistribution of loads under inelastic action.

In the design of PR composite connections, it is assumed that bending and shearloads can be considered separately with the bending assigned to the steel in theslab and a bottom-flange steel angle or plate and the shear assigned to a webangle or plate. Design methodologies and standardized guidelines for C-PRMFframes and connections have been published (Ammerman and Leon, 1990; Leonand Forcier, 1992; Steager and Leon, 1993; Leon, 1990). The performance of thebase connection also depends, of course, on the cyclic performance of the anchorsand the surrounding concrete (Klingner and Graces, 2001).

Subassemblage tests show that when properly detailed, the PR composite con-nections such as those shown in Figure C-II-7.4 can undergo large deformationswithout fracturing. The connections generally are designed with a yield stress thatis less than that of the connected members to prevent local limit states, such aslocal buckling of the flange in compression, web crippling of the beam, Panel Zoneyielding in the column, and bolt or weld failures, from controlling. When theselimit states are avoided, large connection ductilities should ensure excellent frameperformance under large inelastic load reversals.

C-PRMF were originally proposed for areas of low to moderate seismicity inthe eastern United States (Seismic Design Categories C and below). However,with appropriate detailing and analysis, C-PRMF can be used in areas of higherseismicity (Leon, 1990). Tests and analyses of these systems have demonstratedthat the seismically induced loads on PR Moment Frames can be lower than thosefor FR Moment Frames due to: (1) lengthening in the natural period due to yieldingin the connections and (2) stable hysteretic behavior of the connections (Naderand Astaneh-Asl, 1992; DiCorso, Reinhorn, Dickerson, Radziminski, and Harper,


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164 PART II – COMPOSITE SPECIAL MOMENT FRAMES (C-SMF) [Comm. C9.

1989). Thus, in some cases, C-PRMF can be designed for lower seismic loads thanOMF. Because the force transfer relies on bearing of the concrete slab against thecolumn flange-bearing capacity of the concrete should be carefully checked. Thefull nominal slab depth should be available for a distance of at least 6 in. (152 mm)from the column flange.

For frames up to four stories, the design should be made using an analysis that, asa minimum, accounts for the semi-rigid behavior of the connections by utilizinglinear springs with reduced stiffness (Bjorhovde, 1984). The effective connec-tion stiffness should be considered for determining member load distributions anddeflections, calculating the building’s period of vibration, and checking framestability. Frame stability can be addressed using conventional effective bucklinglength procedures. However, the connection flexibility should be considered indetermining the rotational restraint at the ends of the columns. For structurestaller than four stories, drift and stability need to be carefully checked using anal-ysis techniques that incorporate both geometric and connection non-linearities(Ammerman and Leon, 1990; Chen and Lui, 1991). PR composite connectionscan also be used as part of the gravity load system for Braced Frames providedthat minimum design criteria such as those proposed by Leon and Ammerman(1990) are followed. In this case no height limitation applies, and the frame shouldbe designed as a braced system.

Because the moments of inertia for composite beams in the negative and posi-tive regions are different, the use of either value alone for the beam members inthe analysis can lead to significant errors. Therefore, the use of a weighted aver-age is recommended (Ammerman and Leon, 1990; Leon and Ammerman, 1990;Zaremba, 1988).

C9. COMPOSITE SPECIAL MOMENT FRAMES (C-SMF)

C9.1. Scope

Composite Moment Frames include a variety of configurations where steel orcomposite beams are combined with reinforced concrete or composite columns. Inparticular, composite frames with steel floor framing and composite or reinforcedconcrete columns have been used in recent years as a cost-effective alternative toframes with reinforced concrete floors (Furlong, 1997; Griffis, 1992b). For seismicdesign, composite Moment Frames are classified as Special, Intermediate, or Ordi-nary depending upon the detailing requirements for the members and connectionsof the frame. As shown in Table C-II-4.1, C-SMF are primarily intended for usein Seismic Design Categories D and above. Design and detailing provisions forC-SMF are comparable to those required for steel and reinforced concrete SMFand are intended to confine inelastic deformation to the beams. Since the inelasticbehavior of C-SMF is comparable to that for steel or reinforced concrete SMF, theR and Cd values are the same as for those systems.

C9.3. Beams

The use of composite trusses as flexural members in C-SMF is not permitted unlesssubstantiating evidence is provided to demonstrate adequate seismic resistance ofthe system. This limitation applies only to members that are part of the SeismicLoad Resisting System and does not apply to joists and trusses that carry gravity


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Comm. C9.] PART II – COMPOSITE SPECIAL MOMENT FRAMES (C-SMF) 165

loads only. Trusses and open web joists generally are regarded as ineffective asflexural members in lateral load systems unless either (1) the web members havebeen carefully detailed through a limit-state design approach to delay, control, oravoid overall buckling of compression members, local buckling, or failures at theconnections (Itani and Goel, 1991) or (2) a strong-beam/weak-column mechanismis adopted and the truss and its connections proportioned accordingly (Camachoand Galambos, 1993). Both approaches can be used for one-story industrial-typestructures where the gravity loads are small and ductility demands on the criticalmembers can be sustained. Under these conditions and when properly propor-tioned, these systems have been shown to provide adequate ductility and energydissipation capability.

C9.4. Moment Connections

A schematic connection drawing for composite Moment Frames with reinforcedconcrete columns is shown in Figure C-II-7.2 where the steel beam runs contin-uously through the column and is spliced away from the beam-to-column con-nection. Often, a small steel column that is interrupted by the beam is used forerection and is later encased in the reinforced concrete column (Griffis, 1992b).Since the late 1980s, over 60 large-scale tests of this type of connection havebeen conducted in the United States and Japan under both monotonic and cyclicloading (Sheikh et al., 1989; Kanno and Deierlein, 1997; Nishiyama et al., 1990;Parra-Montesinos and Wight, 2000). The results of these tests show that care-fully detailed connections can perform as well as seismically designed steel orreinforced concrete connections. In particular, details such as the one shown inFigure C-II-7.2 avoid the need for field welding of the beam flange at the criti-cal beam-to-column junction. Therefore, these joints are generally not susceptibleto the fracture behavior that is now recognized as a critical aspect of weldedsteel moment connections. Tests have shown that, of the many possible ways ofstrengthening the joint, face bearing plates (see Figure C-II-7.2) and steel bandplates (Figure C-II-9.1) attached to the beam are very effective for both mobilizingthe joint shear strength of reinforced concrete and providing confinement to theconcrete. Further information on design methods and equations for these compos-ite connections is available in guidelines prepared by ASCE (Nishiyama et al.,

Fig. C-II-9.1. Steel band plates used for strengthening the joint.


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166 PART II – COMPOSITE SPECIAL MOMENT FRAMES (C-SMF) [Comm. C9.

Fig. C-II-9.2. Composite (encased) column-to-steel beam moment connection.

1990) and Parra-Montesinos and Wight (2001). Note that while the scope of thecurrent ASCE Guidelines (ASCE, 1994) limits their application to regions of lowto moderate seismicity, recent test data indicate that the ASCE Guidelines are ade-quate for regions of high seismicity as well (Kanno and Deierlein, 1997; Nishiyamaet al., 1990).

Connections between steel beams and encased composite columns (seeFigure C-II-9.2) have been used and tested extensively in Japan where design pro-visions are included in Architectural Institute of Japan standards (AIJ, 1991). Alter-natively, the connection strength can be conservatively calculated as the strength ofthe connection of the steel beam to the steel column. Or, depending upon the jointproportions and detail, where appropriate, the strength can be calculated using anadaptation of design models for connections between steel beams and reinforcedconcrete columns (ASCE, 1994). One disadvantage of this connection detail com-pared to the one shown in Figure C-II-7.2 is that, like standard steel construction,the detail in Figure C-II-9.2 requires welding of the beam flange to the steelcolumn.

Connections to filled composite columns (see Figure C-II-9.3) have been used lessfrequently and only a few tests of this type have been reported (Azizinamini andPrakash, 1993). Where the steel beams run continuously through the compositecolumn, the internal load transfer mechanisms and behavior of these connectionsare similar to those for connections to reinforced concrete columns (Figure C-II-7.2). Otherwise, where the beam is interrupted at the column face, special detailsare needed to transfer the column flange loads through the connection.

These Provisions require that connections in C-SMF meet the same inelastic ro-tation capacity of 0.03 radian as required for steel SMF in Part I. In connectiondetails where the beam runs continuously through the joint (Figure C-II-7.2) andthe connection is not susceptible to fracture, then the connection design can besubstantiated from available test data that is not subjected to requirements such asthose described in Part I, Appendix S. However, where the connection is interruptedand fracture is of concern, then connection performance should be substantiatedfollowing requirements similar to those in Part I, Appendix S.


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Comm. C13.] PART II – COMPOSITE CONCENTRICALLY BRACED FRAMES (C-CBF) 167

Fig. C-II-9.3. Concrete-filled tube column-to-steel beam moment connection.

C10. COMPOSITE INTERMEDIATE MOMENT FRAMES (C-IMF)

The basic construction and connections for C-IMF are similar to C-SMF exceptthat many of the seismic detailing requirements have been relaxed. C-IMF arelimited for use in Seismic Design Category C and below, and provisions for C-IMF are comparable to those required for reinforced concrete IMF and betweenthose for steel IMF and OMF. The R and Cd values for C-IMF are equal to thosefor reinforced concrete IMF and between those for steel IMF and OMF.

C11. COMPOSITE ORDINARY MOMENT FRAMES (C-OMF)

C-OMF represent a type of composite Moment Frame that is designed and detailedfollowing the LRFD Specification and ACI 318, excluding Chapter 21. C-OMFare limited to Seismic Design Categories A and B, and the design provisions arecomparable to those for reinforced concrete and steel frames that are designedwithout any special seismic detailing. The R and Cd values for C-OMF are chosenaccordingly.

C12. COMPOSITE ORDINARY BRACED FRAMES (C-OBF)

Composite Braced Frames consisting of steel, composite and/or reinforced con-crete elements have been used in low- and high-rise buildings in regions of low andmoderate seismicity. The C-OBF category is provided for systems without specialseismic detailing that are used in Seismic Design Categories A and B. Becausesignificant inelastic load redistribution is not relied upon in the design, there isno distinction between frames where braces frame concentrically or eccentricallyinto the beams and columns.

C13. COMPOSITE CONCENTRICALLY BRACED FRAMES (C-CBF)

C-CBF is one of the two types of composite Braced Frames that is speciallydetailed for Sesimic Design Categories C and above; the other is C-EBF (see


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168 PART II – COMPOSITE CONCENTRICALLY BRACED FRAMES (C-CBF) [Comm. C13.

Fig. C-II-13.1. Reinforced concrete (or composite) column-to-steel concentric brace.

Table C-II-4.1). While experience using C-CBF is limited in high seismic regions,the design provisions for C-CBF are intended to result in behavior comparable tosteel SCBF, wherein the braces often are the elements most susceptible to inelasticdeformations (see Part I Commentary Section C13). The R and Cd values andusage limitations for C-CBF are similar to those for steel SCBF.

In cases where composite braces are used (either concrete-filled or concrete-encased), the concrete has the potential to stiffen the steel section and preventor deter brace buckling while at the same time increasing the capability to dissi-pate energy. The filling of steel tubes with concrete has been shown to effectivelystiffen the tube walls and inhibit local buckling (Goel and Lee, 1992). For concrete-encased steel braces, the concrete should be sufficiently reinforced and confinedto prevent the steel shape from buckling. It is recommended that composite bracesbe designed to meet all requirements of composite columns as specified in Part II,Sections 6.4a through 6.4c. Composite braces in tension should be designed basedon the steel section alone unless test data justify higher strengths. Braces that areall steel should be designed to meet all requirements for steel braces in Part I ofthese Provisions. Reinforced concrete and composite columns in C-CBF are de-tailed with similar requirements to columns in C-SMF. With further research, itmay be possible to relax these detailing requirements in the future.

Examples of connections used in C-CBF are shown in Figures C-II-13.1 throughC-II-13.3. Careful design and detailing of the connections in a C-CBF is requiredto prevent failure before developing the strength of the braces in either tension orcompression. All connection strengths should be capable of developing the fullstrength of the braces in tension and compression. Where the brace is composite,the added brace strength afforded by the concrete should be considered. In suchcases, it would be unconservative to base the connection strength on the steelsection alone. Connection design and detailing should recognize that bucklingof the brace could cause excessive rotation at the brace ends and lead to localconnection failure.


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Comm. C15.] PART II – ORDINARY REINFORCED CONCRETE SHEAR WALLS COMPOSITE 169

C14. COMPOSITE ECCENTRICALLY BRACED FRAMES (C-EBF)

Structural steel EBF have been extensively tested and utilized in seismic regionsand are recognized as providing excellent resistance and energy absorption forseismic loads (see Part I Commentary Section C15). While there has been littleuse of C-EBF, the inelastic behavior of the critical steel Link should be essen-tially the same as for steel EBF and inelastic deformations in the composite orreinforced concrete columns should be minimal. Therefore, the R and Cd valuesand usage limitations for C-EBF are the same as those for steel EBF. As describedbelow, careful design and detailing of the brace-to-column and Link-to-columnconnections is essential to the performance of the system.

The basic requirements for C-EBF are the same as those for steel EBF with ad-ditional provisions for the design of composite or reinforced concrete columnsand the composite connections. While the inelastic deformations of the columnsshould be small, as a conservative measure, detailing for the reinforced concreteand encased composite columns are based upon those in ACI 318 Chapter 21. In ad-dition, where Links are adjacent to the column, closely spaced hoop reinforcementis required similar to that used at hinge regions in reinforced concrete SMF. Thisrequirement is in recognition of the large moments and load reversals imposed inthe columns near the Links.

Satisfactory behavior of C-EBF is dependent on making the braces and columnsstrong enough to remain essentially elastic under loads generated by inelasticdeformations of the Links. Since this requires an accurate calculation of the shearLink Nominal Strength, it is important that the shear region of the Link not beencased in concrete. Portions of the beam outside of the Link are permitted to beencased since overstrength outside the Link would not reduce the effectivenessof the system. Shear Links are permitted to be composite with the floor or roofslab since the slab has a minimal effect on the nominal shear strength of the Link.The additional strength provided by composite action with the slab is important toconsider, however, for long Links whose Nominal Strength is governed by flexuralyielding at the ends of the Links (Ricles and Popov, 1989).

In C-EBF where the Link is not adjacent to the column, the concentric brace-to-column connections are similar to those shown for C-CBF (Figures C-II-13.1through C-II-13.3). An example where the Link is adjacent to the column is shownin Figure C-II-14.1. In this case, the Link-to-column connection is similar to com-posite beam-to-column moment connections in C-SMF (see Part II, Section 9) andto steel coupling beam-to-wall connections (see Part II, Section 15).

C15. ORDINARY REINFORCED CONCRETE SHEAR WALLS COMPOSITEWITH STRUCTURAL STEEL ELEMENTS (C-ORCW)

The provisions in this Section apply to three variations of Structural Systems usingreinforced concrete walls. One type is where reinforced concrete walls serve as in-fill panels in what are otherwise steel or composite frames. Examples of typical sec-tions at the wall-to-column interface for such cases are shown in Figures C-II-15.1and C-II-15.2. The details in Figure C-II-15.2 also can occur in the second typeof system where encased steel sections are used as vertical reinforcement in whatare otherwise reinforced concrete shear walls. Finally, the third variation is where


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170 PART II – ORDINARY REINFORCED CONCRETE SHEAR WALLS COMPOSITE [Comm. C15.

Fig. C-II-13.2. Reinforced concrete (or composite) column-to-steel concentric brace.

Fig. C-II-13.3. Concrete filled tube or pipe column-to-steel concentric base.

steel or composite beams are used to couple two or more reinforced concrete walls.Examples of coupling beam-to-wall connections are shown in Figures C-II-15.3and C-II-15.4. When properly designed, each of these systems should have shearstrength and stiffness comparable to those of pure reinforced concrete shear wallsystems. The structural steel sections in the boundary members will, however,increase the in-plane flexural strength of the columns and delay flexural hingingin tall walls. R and Cd values for reinforced concrete shear walls with composite


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Comm. C15.] PART II – ORDINARY REINFORCED CONCRETE SHEAR WALLS COMPOSITE 171

Fig. C-II-14.1. Reinforced concrete (or composite) column-to-steel eccentric brace. (Note:Stiffeners designed according to Part I, Sect. 15.3)

Fig. C-II-15.1. Partially encased steel boundary element.

elements are the same as those for traditional reinforced concrete shear wall sys-tems. Requirements in this section are for ordinary reinforced concrete shear wallsthat are limited to use in Seismic Design Categories C and below; requirements forspecial reinforced concrete shear walls permitted in Seismic Design Categories Dand above are given in Section 16.

For cases where the reinforced concrete walls frame into non-encased steel shapes(Figure C-II-15.1), mechanical connectors are required to transfer vertical shearbetween the wall and column, and to anchor the wall reinforcement. Additionally,if the wall elements are interrupted by steel beams at floor levels, shear connectorsare needed at the wall-to-beam interface. Tests on concrete infill walls have shownthat if shear connectors are not present, story shear loads are carried primarilythrough diagonal compression struts in the wall panel (Chrysostomou, 1991).This behavior often includes high loads in localized areas of the walls, beams,columns, and connections. The shear stud requirements will improve performanceby providing a more uniform transfer of loads between the infill panels and theboundary members (Hajjar, Tong, Schultz, Shield, and Saari, 2002).


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172 PART II – ORDINARY REINFORCED CONCRETE SHEAR WALLS COMPOSITE [Comm. C15.

Fig. C-II-15.2. Fully encased composite boundary element.

Fig. C-II-15.3. Steel coupling beam to reinforced concrete wall.

Two examples of connections between steel coupling beams to concrete walls areshown in Figures C-II-15.3 and C-II-15.4. The requirements for coupling beamsand their connections are based largely on recent tests of unencased steel couplingbeams (Harries, Mitchell, Cook, and Redwood, 1993; Shahrooz, Remmetter, andQin, 1993). These test data and analyses show that properly detailed couplingbeams can be designed to yield at the face of the concrete wall and provide stablehysteretic behavior under reversed cyclic loads. Under high seismic loads, thecoupling beams are likely to undergo large inelastic deformations through eitherflexural and/or shear yielding. However, for the ordinary class of shear wall, thereare no special requirements to limit the slenderness of coupling beams beyond


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Comm. C16.] PART II – SPECIAL REINFORCED CONCRETE SHEAR WALLS COMPOSITE 173

Fig. C-II-15.4. Steel Coupling beam to reinforced concrete wall with composite boundary member.

those in the LRFD Specification. More stringent provisions are required for thespecial class of shear wall (see Part II, Section 16).

C16. SPECIAL REINFORCED CONCRETE SHEAR WALLS COMPOSITEWITH STRUCTURAL STEEL ELEMENTS (C-SRCW)

Additional requirements are given in this section for composite features of rein-forced concrete walls classified as special that are permitted in Seismic DesignCategories D and above. These provisions are applied in addition to those ex-plained in the commentary to Part II, Section 15. As shown in Table C-II-4.1, theR-value for special reinforced concrete walls is larger than for ordinary walls.

Concerns have been raised that walls with encased steel boundary members mayhave a tendency to split along vertical planes inside the wall near the column.Therefore, the provisions require that transverse steel be continued into the wallfor the distance 2h as shown in Figures C-II-15.1 and C-II-15.2.

As a conservative measure until further research data are available, strengths forshear studs to transfer load into the structural steel boundary members are reducedby 25 percent from their Static Yield Strength. This is done because provisions inthe Specification and most other sources for calculating the Nominal Strength ofshear studs are based on static monotonic tests. The 25 percent reduction in studstrengths need not apply to cases where the steel member is fully encased since theprovisions conservatively neglect the contribution of bond and friction betweenthe steel and concrete.

Several of the requirements for Links in steel EBF are applied to coupling beams toinsure more stable yielding behavior under extreme earthquake loading. It shouldbe noted, however, that the Link requirements for steel EBF are intended forunencased steel members. For encased coupling beams, it may be possible toreduce the web stiffener requirements of Part II, Section 16.3, which are the same


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174 PART II – COMPOSITE STEEL PLATE SHEAR WALLS (C-SPW) [Comm. C17.

as those in Part I, Section 15.3, but currently, there are no data available thatprovides design guidance on this.

C17. COMPOSITE STEEL PLATE SHEAR WALLS (C-SPW)

Steel plate reinforced composite shear walls can be used most effectively wherestory shear loads are large and the required thickness of conventionally reinforcedshear walls is excessive. The provisions limit the shear strength of the wall to theyield stress of the plate because there is insufficient basis from which to developdesign rules for combining the yield stress of the steel plate and the reinforcedconcrete panel. Moreover, since the shear strength of the steel plate usually ismuch greater than that of the reinforced concrete encasement, neglecting the con-tribution of the concrete does not have a significant practical impact. The NEHRPProvisions assign structures with composite walls a slightly higher R value thanspecial reinforced concrete walls because the shear yielding mechanism of thesteel plate will result in more stable hysteretic loops than for reinforced concretewalls (see Table C-II-4.1). The R value for C-SPW is also the same as that forlight frame walls with shear panels.

Two examples of connections between composite walls to either steel or compos-ite boundary elements are shown in Figures C-II-17.1, C-II-17.2, and C-II-17.3.The provisions require that the connections between the plate and the boundarymembers (columns and beams) be designed to develop the full yield stress of theplate. Minimum reinforcement in the concrete cover is required to maintain theintegrity of the wall under reversed cyclic loading and out-of-plane loads. Untilfurther research data are available, the minimum required wall reinforcement isbased upon the specified minimum value for reinforced concrete walls in ACI 318.

The thickness of the concrete encasement and the spacing of shear stud connectorsshould be calculated to ensure that the plate can reach yield prior to overall or

Fig. C-II-17.1. Concrete stiffened steel shear wall with steel boundary member.

Fig. C-II-17.2. Concrete stiffened steel shear wall with composite (encased) boundary member.


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Comm. C17.] PART II – COMPOSITE STEEL PLATE SHEAR WALLS (C-SPW) 175

Fig. C-II-17.3. Concrete filled composite shear wall with two steel plates.

local buckling. It is recommended that overall buckling of the composite panel bechecked using elastic buckling theory using a transformed section stiffness of thewall. For plates with concrete on only one side, stud spacing requirements thatwill meet local plate buckling criteria can be calculated based upon h/t provisionsfor the shear design of webs in steel girders. For example, in LRFD SpecificationAppendix F2.2, the limiting h/t value specified for compact webs subjected to shearis h/tw = 1.10

√kv Es/Fyw . Assuming a conservative value of the plate buckling

coefficient kv = 5 and Fy = 50 ksi (345 MPa), this equation gives the limitingvalue of h/t ≤ 59. For a 3/8-in. (10 mm) thick plate, this gives a maximum valueof h = 22 in. (560 mm) that is representative of the maximum center-to-center studspacing that should suffice for the plate to reach its full shear yielding strength.

Careful consideration should be given to the shear and flexural strength of wallpiers and of spandrels adjacent to openings. In particular, composite walls withlarge door openings may require structural steel boundary members attached tothe steel plate around the openings.


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176


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177PART III. ALLOWABLE STRESS DESIGN (ASD)ALTERNATIVE

C1. SCOPE

Part III has been included in these Seismic Provisions for designers that chooseto use ASD in the seismic design of steel structures. A limited Supplement tothe ASD Specification (AISC, 1989) has been issued (AISC, 2001). As notedin Part I, the seismic requirements are collateral provisions related to the LRFDSpecification. Part I is based upon the limit-state seismic load model used in theNEHRP Provisions (FEMA, 2000g). Since the seismic requirements in Part Iare based upon the expected nonlinear performance of a structure, the use ofASD in its traditional form is somewhat complicated because a knowledge ofDesign Strengths, not allowable stresses, is required to assure that connectors havesufficient strength to allow nonlinear behavior of the connected member(s).

The provisions in Part III allow for the selection of members in an ASD formatto provide the performance intended in Part I. Part III is intended as an overlay toPart I and, when using ASD, the designer will use Part I for the seismic design ofa structure except where a section is replaced by or modified by a section shownin Part III.

Provisions have not been included for the use of ASD with the composite structuralsteel and reinforced concrete systems, members, and connections in Part II becauseACI 318 is in limit-states format.

C4.2. Nominal Strength

The procedures in this section provide a methodology for the conversion of al-lowable stresses into Nominal Strengths, in most cases by removing the factor ofsafety from the ASD equations. These Nominal Strengths are converted to DesignStrengths when multiplied by the Resistance Factors given in Part III, Section 4.3.In general, the Resistance Factors given are consistent with those in the LRFDSpecification.

The remainder of the provisions in Part III translate the provisions of Part I intoASD terminology to correlate with the appropriate sections of ASD.


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178


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179

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Documents

AISC SEISMIC PROVISIONS FOR STRUCTURAL STEEL BUILDINGS