Special Steel Moment Frame

8/11/2019 Special Steel Moment Frame

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June 2002

Use of Deep Colum ns

In

Special Steel Moment Frames

By

Jie-Hua Jay Shen, Ph.D., P.E., S.E.

Associate ProfessorDepartment of Civil and Architectural Engineering

Illinois Institute of Technology

Abolhassan Astaneh-Asl, Ph.D., P.E.

ProfessorDepartment of Civil and Environmental Engineering

University of California, Berkeley

David B. McCallen, Ph.D.Director

Center for Complex Distributed SystemsLawrence Livermore National Laboratory

____________________________________________________________________________

(A copy of this report can be downloaded free of charge for personal use from www.aisc.org)


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Use of Deep Columns in Special Steel Moment Frames, J. Shen, A. Astaneh-Asl and D. B. McCallen, 2002. 1

Use of Deep Columns in Special Steel Moment Frames

By Jie-Hua Jay Shen, Abolhassan Astaneh-Asl and David B. McCallen

This report discusses some of the issues related to the use of deep columns in special moment frames.

Since 1994 Northridge earthquake significant amount of research and development projects have been donein U.S., Japan and elsewhere on seismic behavior and design of steel moment frames. In almost all of these

research projects, the column used in testing or analyses have been W14 or smaller sections. One of the

most important research projects during this period was the SAC Steel joint Venture Project where a large

number of moment connections were tested and analyzed and design recommendations were formulated. In

this project, almost all specimens had a column with depth of no more than 14-16 inches. However, since in

many cases of moment frames, the governing design requirement is the stiffness to control the drift, the use

of deep columns with a depth of 24, 27 and even 30 inches, becomes very economical. Unfortunately, there

is no extensive and reliable information on actual cyclic behavior and design of moment frames with deep

columns. This report discusses: (a) the issues that need to be considered in using deep columns in moment

frames, (b) a comparison of seismic behavior of two 10 story moment frames designed using W14 and

W27 respectively, (c) the results of a series of realistic non-linear finite element analysis of moment-

rotation behavior of connections with deep columns and; (d) the conclusions.

First Printing, June 2002.

__________________________________________________________________________________Jie-Hua Jay Shen, Ph.D., P.E., S.E. Associate Professor, Department of Civil and Architectural Engineering,

Illinois Institute of Technology, 3201 South Dearborne Street, Chicago, IL, 60616.

Phone: (312) 567-5860, Fax: (312) 567-3579.

E-mail: [email protected].

____________________________________________________________________________________________

Abolhassan Astaneh-Asl, Ph.D., P.E., Professor, 781 Davis Hall, Univ. of California, Berkeley, CA 94720-1710,

Phone: (510) 642-4528, Fax: (925) 946-0903,

E-mail: [email protected] , Web page: www.ce.berkeley.edu/~astaneh

____________________________________________________________________________________________David B. McCallen, Ph.D., Director, Center for Complex Distributed Systems, Lawrence Livermore National

Laboratory, 7000 East Avenue, MS L-151, Livermore, CA 94550.

Phone: (925) 423-1219


____________________________________________________________________________________________

Disclaimer: The information presented in this publication has been prepared in accordance with recognized engineering

principles and is for general information only. While it is believed to be accurate, this information should not be used or relied

upon for any specific application without competent professional examination and verification of its accuracy, suitability, and

applicability by a licensed professional engineer, designer or architect. The publication of the material contained herein is not

intended as a representation or warranty on the part of the Structural Steel Educational Council or of any other person named

herein, that this information is suitable for any general or particular use or of freedom from infringement of any patent or

patents. Anyone making use of this information assumes all liability arising from such use.

Caution must be exercised when relying upon specifications and codes developed by others and incorporated by reference

herein since such material may be modified or amended from time to time subsequent to the printing of this document. The

Structural Steel Educational Council or the authors bears no responsibility for such material other than to refer to it and

incorporate it by reference at the time of the initial publication of this document.


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ACKNOWLEDGMENTS

The publication of this report was made possible in part by the support of the Structural

Steel Educational Council (SSEC). The authors wish to thank all SSEC members for their

valuable comments. Particularly, special thanks are due to Fred Boettler, Jeff Eandi, Lanny Flynn,Pat Hassett, William Honeck, Brett Manning and James Putkey for their valuable and detailed

review comments. The authors also appreciate the review comments provided by James Malley

of Degenkolb Engineers and Dr. Farzad Naeim of John A. Martin Associates.

The opinions expressed in this report are solely those of the authors and do not necessarily

reflect the views of the Illinois Institute of Technology, the University of California Berkeley, the

Lawrence Livermore National Laboratory where authors are employed nor the Structural Steel

Educational Council or other agencies and individuals whose names appear in this document.

A portion of this work was performed at Lawrence Livermore National Laboratory under

the auspices of DOE Contract W-7405-Eng-48. The analyses and design of the 10-story frames

were done using the latest version of the SAP-2000n program. The generous donation of the

program by Computers and Structures Inc. of Berkeley (www.csiberkeley.com) is sincerelyappreciated. The finite element analyses of connections were conducted using ABAQUAS and

NIKE-3D program.


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USE OF DEEP COLUMNS IN

SPECIAL STEEL MOMENT FRAMES

By:

JAY SHEN, Ph.D., P.E., S.E.Associate Professor

Department of Civil and Architectural Engineering, Illinois Institute of Technology , Chicago

ABOLHASSAN ASTANEH-ASL, Ph.D., P.E.Professor

Department of Civil and Environmental Engineering, University of California, Berkeley

DAVID B. McCALLEN, Ph.D.Director

Center for Complex Distributed Systems, Lawrence Livermore National Laboratory, Livermore

____________________________________________

TABLE OF CONTENTS

ABSTRACT / Page 1

ACKNOWLEDGMENTS / Page 2

TABLE OF CONTENTS / Page 3

NOTATIONS / Page 4

CHAPTER 1. INTRODUCTION / Page 5

CHAPTER 2. USE AND BEHAVIOR OF FRAMES WITH DEEP COLUMNS / Page 8

CHAPTER 3. ANALYSIS OF CYCLIC BEHAVIOR OF DEEP COLUMN CONNECTIONS / PAGE 17

CHAPTER 4. CONCLUSIONS / Page 33

REFERENCES/Page 36

ABOUT THE AUTHORS / Page 38

LIST OF PUBLISHED STEEL TIPS REPORTS / Page 39


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_________________________________________________________________________

Notations_________________________________________________________________________

In preparing the following notations, whenever possible, the definitions are taken from

various references as indicated inside the parentheses whenever applicable.

bf Width of flange.

E Modulus of elasticity.

Fy Specified minimum yield stress of the type of steel to be used, ksi. As used in the LRFD

Specification, "yield stress" denotes either the minimum specified yield point (for those

steels that have a yield point) or the specified yield strength (for those steels that do not

have yield point). (AISC, 1997).Fyw Specified minimum yield stress of the web.

h Depth of web.

J Torsion constant, cross section property.

in. Inch, 1 inch= 25.4mm.

Ix Moment of inertia about x-axis.

Iy Moment of inertia about y-axis.

ksi Kilo-pounds per square inches, 1 ksi=6,895 kilo-Pascal.

rx Radius of gyration about x-axis.

ry Radius of gyration about y-axis.

Sx Section modulus about x-axis.

Sy Section modulus about y-axis.tf Thickness of flange.

tw Thickness web.

Zx Plastic modulus about x-axis.

Zy Plastic modulus about y-axis.

p Limiting slenderness parameter for a compact element. (AISC, 1997).

r Limiting slenderness parameter for a non-compact element. (AISC, 1997).

f Equals bf/2tf for flange.

w Equals h/tw for web.

c Twisting of column.


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1.Introduction

1.1. Introduction

Moment-resisting frames are one of the frequently used lateral load resisting systems in

many steel building structures. During the 1994 Northridge earthquake, a large number of welded

steel moment frames developed cracks in their beam-to-column welds at or near joints. Although,

none of the damaged structures developed any partial collapse or even injuries, the structural

engineering and steel construction community undertook an extensive effort to study the

phenomenon and mitigate it. In the aftermath of the 1994 Northridge earthquake and during

1994-2000 periods, a comprehensive research and technology development project was

undertaken by SAC Steel Joint Venture (FEMA-350, 2001) primarily funded by the Federal

Emergency Management Agency to address this problem. The main goal of the project,

sometimes denoted as simply the SAC project, was to develop technologies for design,

construction, inspection, evaluation and retrofit of the moment frames subjected to seismic

effects.

As part of the SAC Project, a large number of cyclic tests of beam-to-column connections

of moment frames were conducted. The aim was to establish the actual behavior of existing as

well as the improved beam-to-column moment connections. Most of these tests were done on

specimens where the columns were W14 sections with a maximum depth of column being about

14-16 inches. When the studies were completed, SAC Project produced a set of reports (FEMA-

35, 2001) on various aspects of the problem and its solutions. One of the important items in the

FEMA reports was the introduction of pre-qualified moment connections. The pre-qualified

connections have specific ranges of material properties and geometry, which are based on tested

connections. It is expected that if properties of a designed connection fall within these ranges, the

designed connection will behave in a manner similar to those tested within the SAC Program.


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Almost all the pre-qualified connections in SAC reports have a W14 column traditionally

used in many structures. However, in todays design offices, structural engineers in many projects

find it more economical to use columns that are deeper than the W14 sections. In recent years, it

has been recognized that there is a strong economic incentive for the design engineer to use deep

columns to satisfy increasingly more stringent drift limitations. Using W14 columns to satisfy driftlimitations specified by the codes often results in unnecessarily heavy columns. Structural

engineers have, from time to time, used deeper columns for some steel building projects, when

they had resources to carry out the physical tests of project-based connections. The deep

columns would be more extensively used for moderate-rise to high-rise buildings if the time

consuming and costly physical tests could be avoided. So far, limited research has been done

regarding the behavior and design of a beam-to-column connection with deep columns. Two

reports (Gilton et al, 2000) and (Ricles et al., 2000) include the results of cyclic testing of a few

beam-to-column connection specimens where the column was a deep wide flange section.

Therefore, there is a need for information on the performance of beam-to-column momentconnections with deep columns. A deep column in this context is a column with a depth of

greater than 21 inches.

1.2. Background on This Study

After the 1994 Northridge earthquake, extensive studies were conducted to improve the

performance of the steel moment-resisting frame when subjected to strong ground motions. Since

then, the Reduced Beam Section (RBS), where a portion of the beam flange is removed in order

to force the plastic hinge in the beam away from the column face, has become one of thefrequently used welded moment connections. Researchers have studied the behavior of the RBS

connections when connected to W14 columns (FEMA-350, 2001), and have found that the

connections with RBS have larger cyclic rotational ductility than the same connections without

RBS. This type of beam-to-column connection assembly has been pre-qualified by FEMA-350

for seismic design of moment-resisting frames along with a number of other configurations of

welded and bolted connections

In 2000, a report by Gilton et al. (2000) presented the results of cyclic tests of three RBS

moment connections where deep columns were used. The authors have reported twisting of the

deep columns. Although the twisting of deep column in their tests appears to have been observedduring the late stages of loading and after rotations in excess of 0.03 radian, the authors have

expressed concern about twisting of the deep columns and have formulated and proposed

limitations on the geometry of the column cross section to prevent the observed twisting of deep

columns. A review of the report by Gilton et al (2000) indicates that the lateral movement of

RBS hinge and the resulting twisting of deep columns in their tests may have been due to


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unrealistic boundary conditions and lack of bracing normally provided to top flange by the floor

beams.

To investigate this, non-linear cyclic behavior of RBS moment connections with W14 and

deep columns were studied and the results are summarized here. The analyses began with

building the model of a beam-to-column sub-assemblage that had been physically tested (Gilton

et. al., 2000). After the results of a tested specimen was well simulated by a finite element model,

a group of more realistic beam-to-column sub-assemblages with other deep column configurations

were analyzed, and the results were evaluated. The results confirmed that indeed column twisting

in Gilton et al. (2000) tests might have occurred primarily because of the way the specimen was

tested. In these tests, there was no flange bracing which normally is provided to the top flange of

the beam by the floors in actual buildings.

The authors hope the information presented here can be useful in better understanding the

actual behavior of moment connections with deep columns in buildings. In addition, we hope the

information can assist future researchers in planning their test set-up to test moment connections

with deep columns in a realistic and proper manner.

1.3. Objectives of this Report

The main objectives of this Steel Technical Information and Product Services (Steel TIPS)

report are:

1. To review the use of frames with deep columns (Section 2).

2. To conduct pushover and inelastic time history analyses of frames with W14 as well as

deep columns and compare their seismic behavior (Section 2).

3. To conduct a critical review of the results of a few cyclic tests available at this time on

deep columns. The deep columns are defined as columns with a depth of 21 inches or

greater, particularly columns with 24, 27, 30 and 33 inch depths (Section 3).

4. Using realistic models of the connections with deep columns, to conduct simulated cyclic

tests of these connections and compare the results of computer analyses to actual test

results to ensure that the computer analyses predict the actual test results well (Section 3).

5. To conduct more analyses of moment connections with different beam and deep columns

sections and with floors being present or not (Section 3).

6. To formulate tentative recommendations for the use of deep columns in moment frames.

Such recommendations can be verified by selective, well-planned and correctly executed

testing (Section 4).


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2. USE AND BEHAVIOROF FRAMES WITH

DEEP COLUMNS

2.1. Introduction

In most cases of design of moment frames, drift limitations, and not strength, govern the

design. One of the efficient ways of reducing the drift of a moment frame is to increase the

bending and shear stiffness of its columns. Using deeper cross sections than the W14s

traditionally used in many moment frames will accomplish this. The following text provides a

discussion of the issues related to the use of deep columns.

2.2. Issues Related to the Use of Deep Columns

2.2.a. Stiffness of the Moment Frame

Deep columns with W21 to W30 sections provide larger moment of inertia for the same

weight compared to traditional W14 column sections. For example, the weight/ft of a W27

section will be less than of the weight/ft of a W14 section with comparable moment of inertia.

Relatively large bending stiffness of the deep columns results in increasing the global stiffness of

the moment frame, which in turn results in reducing the drift and damage.

2.2.b. Strength

In moment frames subjected to relatively large lateral forces, bending strength of thecolumns is one of the important parameters. Deep columns provide larger plastic moment capacity

than the equivalent W14s, making it possible to more easily meet the strong column-weak beam

design requirements. For example, the weight/ft of a W27 section will be less than 70% of the

weight of a W14 section having the same plastic moment capacity. In using deep columns with

relatively small weak axis moments of inertia, one has to check the possibility of lateral torsional

buckling of the deep column, especially for tall floors.


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yww FEth /45.2/

ywyp FErL /7.1=

ywy FErL /7.1/

According to AISC Specification (AISC, 2001), if un-braced length of compression flange

of a beam in bending is less than the Lpgiven by the following Equation 2.2, lateral-torsional

buckling is not expected to occur before the beam reaches its plastic moment capacity.

If L Lp the beam is compact for lateral-torsional buckling, where:

(AISC-LRFD Manual, 2001, P. 16.1-33) (2.1)

By rearranging the above equation we can obtain a limit for L/ry of the column, Equation

2.2, that below this limit lateral-torsional buckling is not expected and need not be checked.

(2.2)

For A36, Grade 50 and Grade 65 steel, the above limit of L/ryis equal to 48, 41 and 36

respectively.

2.2.c. Panel Zone Issues

Deep column sections have deeper webs than the W14 columns and provide more web

area than the W14 for the same weight. This means that shear strength and stiffness of the panel

zone in a deep column is greater than the corresponding values in a W14 column with the same

weight. The larger shear strength of the panel zone in deep columns can help reduce the need for

doubler plates. The larger shear stiffness of the panel zone in deep columns can help reduce panel

zone distortions. As a result, the contribution of panel zone distortions to the story drift can be

smaller when deep columns are used. In deep columns, where the web is relatively slender, shear

buckling of panel zone should be investigated. Shear buckling of web can be avoided by limiting

the h/twof the column web to the following value from the AISC Specification (AISC, 2001).

(AISC-LRFD Manual, 2001, P. 16.1-35) (2.3)

If h/tw of the column web satisfies the above equation, it is expected that the column web

can reach shear yielding before buckling. The term on the right side of the Equation 2.3 above

for A36, grade 50 and Grade 65 steel (Fy=36, 50 and 65 ksi) is equal to 69, 59 and 52


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respectively. A check on currently available rolled shapes indicate that all rolled wide flange

shapes tabulated in the first part of the current AISC-LRFD Manual (AISC, 2001) have h/tw less

than 59 therefore satisfy the limit of Equation 2.3 above for A36 and grade 50 steels. For grade

65 steel, with the exception of a few sections, almost all rolled shapes have h/tw less than 52

satisfying the limit of Equation 2.3.

2.2.d. Local Buckling

As far as local buckling is concerned, deep columns have a disadvantage compared to

W14 columns. In general, b/t ratio of flanges and h/twof webs of deep columns are larger than the

W14s with the same weight. However, most deep column sections with grade 50 steel have

compact webs and flanges and can be used in high seismic areas.

2.3. Comparison of Behavior of a Frame with W14 and Deep Columns

In order to identify benefits and limitations of using deep columns in moment frames, a

limited comparative study was done. In the study, a typical building was selected and was

designed using W14 columns. Then, the same building was designed using W27 columns. Both

frames had the same girders. The results of analyses of these two frames indicated that in all

respects, the frames behaved similarly. However, the weight of the frame with W27 columns was

considerably less. Of course, one should not generalize the outcome of this one case of

comparison, but as an example, it sheds some light on seismic behavior of similar frames with

W14 and W27 columns. In addition, it shows the extent of saving in the weight of columns for

this building if one uses deep columns.

2.3.a. Building Used in the Comparative Studies

The building selected for the comparative study was a 10-story perimeter frame building.

This building structure, using W14 columns, was almost the same as the structure of a 10-story

study building designed by the SAC Joint Venture (SAC, 1996) and provided to researchers in

1996. For these studies, the building was assumed located in seismic areas of California within a

10 km distance of a major fault. Hayward fault ground motions were the used in the nonlinear

time history analyses. SAC designed the study buildings to comply with the UBC-97 (ICBO,

1997). Figures 2.1 shows framing plan and elevation of the 10-story study structure.


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2.3.b. Design of the Building Used in the Comparative Studies

As indicated earlier, the building used in the study was adapted from one of the study

buildings that was developed and used in the SAC Joint Venture program (SAC, 1996). The ten-

story building designed by SAC for a Los Angeles site had W14 columns. The SAC-designedstructure complied with the UBC-97 and its maximum inter-story drift (for 18 feet tall ground

floor, see Figure 2.1) was 1.7%, which is less than the 2% limit given by the UBC-97 for this

structure. The frame on column line 6 of SAC structure was selected as one of our two study

frames and was denoted as W14 Study Frame. Then, we replaced the W14 columns with W27

columns while keeping the same beams and denoted this frame W27 Study Frame. Since in

moment frames, usually drift is the governing design parameter, the replacement W27 were

selected such that the frame had still a drift value less than 2% and both W14 and W27 study

frames had comparable stress level in their members. Figure 2.2 shows cross sections of the

girders and columns used in both frames. Figure 2.3 shows Demand/Capacity ratios for membersof study frames. Instead of LRFD methods, in the design we used AISC-ASD design option of

the SAP2000n software and the nominal loads. This was done to be able to compare the stresses

and deformations generated in each frame by the combined design forces at service load level and

not at factored-load levels. The use of ASD methods here is not to advocate its use in design,

which is best done using LRFD methods. To the authors, the ASD method provided a better feel

about service level (unfactored) stresses and deformations in the frames.

Figure 2.1. Plan and Elevation Views of the 10-Story Structure

18 ft

8 @ 13

12 ft

ELEVATIONPLAN

30 ft

30 ft

30 ft

30 ft

30 ft

30 ft30 ft

5 6

A

B

C

D

E

F

30 ft 30 ft 30 ft

2 3 41


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The analysis of the frame with W27 columns showed that the maximum inter-story drift in

the frame was 1.2% and for the frame with W14 columns was 1.7%. Both drift values were less

than the limit of 2% as per UBC-97 (IBC, 1997) and occurred at the 18 feet tall ground floor.

Figure 2.3 shows values from the interaction equation for the two frames, which indicates the

stress level at code service level forces to be similar in both frames and relatively low as expectedin a moment frame.

Figure 2.2. Girders and Columns of W14, and W27 Study Frames

W14 Study Frame

W27 Study Frame


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Push-over Anal yses:

In order to compare the performance of two frames, using the SAP 2000n program,

pushover analyses of the frames shown in Figure 2.2 were conducted. In the pushover analyses,

both frames were subjected to ever-increasing first mode pushover displacements. Figure 2.4

shows the push over curves. Both frames were able to reach a roof displacement of about 2.5 feet

before collapse. Figure 2.5 shows the hinges at the time of collapse. The frame with W14

columns showed soft story formation while the frame with W27 columns had more yielding in the

columns at the time of collapse. The columns in the frame with W27 columns were considerably

lighter than the columns in the frame with W14 columns.

Figure 2.4. Pushover Curves for the Frames with W14 and W27 Columns

Roof Displacement, ft.

Base

Shear,

kips.

0

1000

3000

1.0 2.0 3.0

2000

W27

W14

Roof Disp.


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Note: Indicates a plastic hinge with partial yielding

Indicates a plastic hinge with full yielding of the cross section

Figure 2.5. Hinges in the Frames Just Prior to Collapse

I nelastic Time H istory Analyses:

In order to compare the dynamic response of two frames, inelastic time history analyses of

both frames were conducted. The dead and live load as well as the mass applied to both frames

were the same as given by SAC (FEMA-350, 2001). The inelastic models of the frames shown in

Figure 2.2 were subjected to the E-W acceleration component of the Hayward Seismic Evaluation

Earthquake (SEE) generated by Bolt and Gregor (1993). Figure 2.6 shows the time history of

displacement of the first floor for the two frames. The drift values for the first floor can be

obtained by dividing displacements by 18 feet, the height of ground floor. The inter-story drift of

the frames with W27 and W14 columns were 1% and 1.2% respectively. The drift values

calculated using UBC-97 (ICBO, 1997) provisions were 1.2% and 1.7% for frames with W14 andW27 columns respectively. Plastic hinges formed in both frames at the RBS areas. However,

since the girders in both frames were the same, it was not expected that non-linear behavior of

frames would be much different.

In previous sections, it was shown that the drift values and stresses in two study frames,

one with W14 columns and the other with W27 columns, were essentially the same. However, for


17/41


this 10-story building with a 150ft by 150ft plan, the weight of the steel using W27 deep columns

was about 1.3 lbs/ft2less than the steel in the same frame but with W14 columns. According to a

leading steel fabricator, the 1.3 lbs/ft2equals to about 6-8% in total material saving based on 16-

18 psf of steel for a typical structure of this type. Of course as mentioned earlier, this 10-story

building was just an example to demonstrate that using deep columns instead of W14 can result inimprovement in lateral load resisting behavior, much better drift and damage control as well as

possible savings in the cost of construction of steel frames.

Figure 2.6. Time History of Horizontal Displacement of First Floor to

Hayward SEE Earthquake

W14 Frame

W27 Frame


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3. ANALYSIS OF

CYCLIC BEHAVIOROF DEEP COLUMN

CONNECTIONS

3.1. Introduction

This Chapter investigates, analytically, the cyclic behavior of beam-to-column connections

with deep column sections ranging from W14 to W33. A compact beam section was used for

most of parametric studies, since; almost all available wide flange sections are compact. For

comparison, a non-compact section beam was also included. Detailed nonlinear finite element

analyses were conducted to address the issues that influence the cyclic performance and design

considerations of one of the most commonly used connections pre-qualified by FEMA-350

(2001), namely the RBS connection, whit the column becoming deeper and deeper. In the

following sections, a summary of the results of these studies is presented.

3.2. Simulation of Cyclic Behavior of Tested Specimen

3.2.a Computer Model of Test Specimen

As indicated in previous chapter, two of the three specimens tested by Gilton et al. (2000)

had web doubler plates added to the column panel zone. The third specimen without the doubler

plate, assumed to more realistically represent the current design practice, was therefore selected

to be modeled and analyzed in this study. This specimen was Specimen DC-2 (Gilton et. al.,2000). A nonlinear finite element model of this specimen was constructed with the nonlinear

finite element program, ABAQUS (ABAQUS, 2001). The specimen was a standard beam-to-

column assembly consisting of a W27194 column and a W36150 beam, both specified as A572

Gr.50 steel. A reduced beam section (RBS) was introduced to make the beam side of the

connection pre-qualified by FEMA 350 (FEMA 350, 2001). The details of the RBS, the column


19/41


stiffeners and web shear tab plate are shown in Figure 3.1. The test setup of the beam-to-column

assembly connection is shown in Figure 3.2.

Figure 3.1. Non-Linear Computer Model of the Specimen

Figure 3.2. Model of Test Set-up Used by Gilton et al. (2000)


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The computer model, denoted here as ABQ-DEEP, used fully integrated six-node

and eight-node three-dimensional solid elements (Element types C3D6 and C3D8 in ABAQUS).

A finer mesh was used in the RBS area, panel zone and shear tab plate areas. Rigid links were

used to connect the beam tip to the actual loading point (reference node), which was also

restrained to prevent out-of-plane translation (Figure 3.2). The material properties of the steel,yield strength and ultimate strength, were specified from the mill certified coupon test of the

Specimen DC-2 (see Table 3.1). Stress-strain curve for the steel was a tri-liner curve with three

segments: (a) first segment, (the elastic segment) from the origin to the yield point, (b) the second

segment from the yield point to ultimate strength point with stress equal to Fu and strain of 0.20;

and (c) the last segment, a horizontal line at stress level of Fu.

Table 3.1. Properties of Specimen DC-2 Tested by Gilton et al., (2000)

Cyclic loading pattern in the test, controlled by the displacement at the tip of the beam,

was of a standard small-to-large displacement cycles as shown in Figure 3.3. At small

displacements, the cycles were repeated four times. At larger inelastic displacements, the cycles

were repeated twice.

Figure 3.3. Loading History Used in the Test and Analysis


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3.2.b. Simulated Cyclic Behavior of Connection

When simulated cyclic loading was applied to the nonlinear model of specimen, the

specimen remained virtually elastic before 1% drift cycles, when some yielding was observed.

Though such elastic deformation cycles might be desirable for physical testing, a finite element

analysis does not record any effects of elastic cyclic loading and unloading on the assembly. Thus,

in the simulation analysis, the cyclic loading history for the analysis started from the cycles

immediately before any yielding was observed. The number of inelastic cycles appears to have a

significant influence on the post-buckling behavior in terms of strength degradation. The actual

test of specimen DC-2 indicated that strength was reduced considerably when the inelastic cycle

was repeated. Such cycle-related strength reduction became more significant when a larger

inelastic cycle was repeated, apparently due to the Bauschinger effect leading to local buckling

and low cycle fatigue phenomenon.

3.2.3. Comparison of Analytic and Experimental Results

Figure 3.4 shows the load-displacement curves from the test specimen DC-2 tested by

Gilton et al. (2000), and from the analysis discussed here. The overall cyclic responses from the

analysis and the test match reasonably well. There are some noticeable discrepancies in unloading

and reloading regions, particularly at large inelastic deformation levels. The unloading curve of

the tested specimen was highly nonlinear, significantly different from the linear unloading curve

conventionally used as analytical models of hysteretic behavior. The reloading in an opposite

direction after a full inelastic unloading made the specimen softer. The softening in unloading and

reloading appear to have been responsible for an accelerated strength reduction from its peak

value after each cycle with the same or higher level of displacement.

The deformed shapes of specimen from analysis model at 5% drift level are presented in

Figures 3.5, showing an isometric view of the buckling shape near the beam-to-column joint. The

deformed shape is similar to the final buckling shape observed in the test (Gilton et. al., 2000),

especially large deformations in the RBS area. Figure 3.6 shows top and end views of the

deformed specimen at 5% story drift.


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Figure 3.4. Load-Displacement Curve of Specimen DC-2 and Analytical Results


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Figure 3.5 Buckling Shape of the Specimen Model at 5% Story Drift

Figure 3.6. Deformed Shape of Web and Flanges at 5% Story Drift

3.3. Parametric Study of Cyclic Behavior of Deep Column Connections

Having successfully simulated the cyclic behavior of the tested specimen, the ABAQUS

model, ABQ-DEEP as the prototype, was used to model the connection assembly with various

column sizes. In the seismic design of steel moment-resisting frames based on improved

connection details summarized in recent FEMA publications (FEMA-350, 2001), there are some

concerns related to the connection strength reduction after its peak strength is reached. Slower

reduction might indicate a more stable connection performance, and vice versa. It has been


24/41


observed that strength reduction after the peak strength is reached heavily depends on the number

of inelastic cycles. The main goals of the parametric studies were:

1. To investigate whether or not there are any significant characteristics in a connection with

deep column sections that are not considered in current design practice;

2. To investigate the effects of floor slab and transverse beams in bracing the connection and

preventing lateral movement of hinge areas.

Six beam-to-column connection assemblies were studied analytically. Five of them had the

columns listed in Table 3.2, and the W36x150 beam section. The five columns were selected to

construct the connection assemblies within a practical range. The column sections were selected

based on their plastic section modulus (Zx) and moments of inertia (Ixand Iy), so that the

comparison could be made with respect to lateral movement of the hinge areas and twisting of

columns with different combinations of Zx, Ix, and Iy.

Table 3.2. Section Properties of the Studied Column Sections

In addition, the effect of lateral bracing on the connection assembly performance was also

investigated by introducing actual lateral supports from transverse beams and the concrete with

metal deck floor that exists in almost all steel framed buildings. To study bracing effects of the

floor slab, in some analytical cases, the beam was laterally braced along the beam top flange

outside the RBS. Two different boundary condition cases were considered: (1) Unbraced case

where the beam had no lateral restraints similar to specimens tested by Gilton et al (2000); and (2)


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Braced case where the beam was laterally restrained in its panel zone and top flange except in

the RBS region. For comparison, an additional beam-to-column connection with a non-compact

beam section, W30x90, and a W27x194 column, was also included in this study. The cyclic

analyses applied a maximum displacement of 6% story drift ratio in the same manner as

conducting a physical test per FEMA-350 (2001). The following sections will present a summaryof the analytical results together with discussions of various issues.

3.3.1. Overall Cyclic Behavior of Deep Column Connections

Figures 3.7, 3.8, and 3.9 show the cyclic behavior of the connection assemblies with

W30x191, W33x169, and W201x201, respectively. The cyclic loops of the connections

demonstrated that the connections with deeper columns were stable. With lateral bracing (the

solid-blue lines in the figures), the connections did not have any significant strength reductionbefore the 4% drift ratio. Under the cyclic loading, the strength degradation occurred upon the

Figure 3.7. Cyclic Behavior of the Connection with W30x191 Column and

W36x150 Beam.


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load reversal in both positive and negative deformation regions after the plastic hinge formed in

the RBS region at about 3% drift ratio, mainly due to inelastic local web and flange buckling.

Without lateral bracing (the dashed-red lines in Figures 3.7, 3.8, and 3.9), the connections

experienced column twisting and beam lateral torsional buckling after 4% drift ratio,

demonstrating a larger strength reduction than those with lateral bracing.

It seems apparent that the lateral supports to the beam flange under compression improved

the inelastic behavior of connections with deep columns. In particular, the post-buckling strength

degradation was reduced considerably by lateral supports provided by the floor, as shown in

Figures 3.7, 3.8, and 3.9. The lateral supports to the beam prevented lateral movement of plastic

hinge area and extended the deformation prior to the onset of strength degradation. The local

buckling of the flanges and web was mainly responsible for a slow degradation in strength at a

later deformation stage for the braced connections. A larger strength degradation under negative

bending moment, when the beam top flange was in tension, in the above figures indicates thatextra lateral supports to the bottom flange can help to enhance inelastic cyclic behavior. Note that

all cases involved a compact beam section, W36x150 with Fy=50 ksi. If any non-compact beam

section were used, the strength degradation would have been more significant, as discussed later.

Figure 3.8. Cyclic Behavior of the Connection with W33x169 Column and W36x150 Beam


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3.3.2. Effect of Column Size/Depth

Figure 3.10(a) shows the plan views of deformed RBS connections, with no floor

slab and transverse beams present, at a relatively large story drift ratio of 6%. The large

story drift was selected to show the deformations at very late stages of cyclic behavior and

at drift values much beyond what can be expected in major seismic event. The figure

shows RBS connections with deep columns where no lateral bracing was provided in

order to reveal the effect of the column size on the lateral stability of the connection

assembly. The larger lateral torsional deformation of the beam was observed when the

column was weaker in out-of-plane stiffness. For example, there was no lateral torsional

buckling of the same beam when the column was changed to a W14x426.

It seems that in this case, due to lack of floor slab and transverse beams, the deep

column was the only element responsible to resist the torque applied to it by the beam.

Being subjected to such twisting effects, the deep columns with no floor underwent

twisting as shown in Figure 3.10(a) for four study cases. The values of c given in Figure

3.10 are approximate values of column twisting alone in degrees.

Figure 3.9. Cyclic Behavior of the Connection with W33x201 Column and W36x150 Beam.


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Figure 3.10. Lateral Deformation of the RBS Area and Column Deformations for:

(a) Connections with no Floor Slab and Transverse Beam; and

(b) Connections with Floor Slabs and Transverse Beams

Figure 3.10(b) shows the same four connections as in Figure 3.10(a) but this time the

connections have floor slab attached to the top flange of the beam at shear stud locations and a

transverse beam is attached to the panel zone of the column. As the figure indicates, by having the

floor slab and transverse beam, the column twisting was negligible.

As can be seen in Table 3.2, the torsional stiffness and weak-axis flexural stiffness of a

W14 sections are greater than the corresponding values for deeper columns with comparable

strong axis flexural stiffness. When a beam-column connection specimen is tested with no slab

and transverse beam, there is no lateral restraint to prevent lateral movement of the highly yielded

and locally buckled RBS hinge as shown in Figure 3.10(a). When the hinge area, not attached to

the floor, moves laterally, it can apply large enough moment to bare column to twist it as shown

in Figure 3.11.

W27x194 Column W30x191 Column W33x169 Column W33x201 Column

Note: All Beams: W36x150)

W27x194 Column W30x191 Column W33x169 Column W33x201 Column

Note: (All Beams: W36x150) ,

(a)

(b)

c 0.0 c 0.0c 0.0

c 0.0

c 1.5 c .2.5 c 3.0 c 2.5


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Figure 3.11. Torque acting on Column due to Lateral Movement of RBS

We believe that the lack of floor slab in Gilton et als (2000) tests is the main reason for

development of column twisting in their tests. Had the floor slab been present, as is the case in

almost all buildings, or at least the restraining effects of floor slab been represented by bracing in

the test set-up, most likely the twisting of columns would have been minor and non-consequential.

It is strongly recommended that in future tests of beam-column connections particularly RBS

connections with deep columns, the restraining effects of the floor be represented either by having

the actual floor cast with the specimen or by attaching to top flange appropriate bracing

mechanisms to represent the floors.

3.3.3. Effect of Beam Section Compactness

It is necessary to use a compact beam section in the earthquake-resistant moment frame to

ensure a stable cyclic performance during a strong earthquake. The limit of b f/2tfratio for a

compact flange, p, is equal to 52 (Fy). In practice, most wide flange sections are compact

sections. In this study, all previous discussions have been based on a compact beam section,

W36x150 (f= bf/2tf= 6.4; f/p=0.87). In this section, a non-compact section, W30x90 (bf/2tf=

8.5; f/p=1.16), was selected to compare the behavior of the deep-column connection assembly

with compact and non-compact beam sections. For definitions of terms, see Notations in Page

4. Figure 3.12 shows the cyclic response of the assembly with W30x90 beam and W27x194

column. The strength reduction rates are 35% and 50% at 4% and 5% story drift levels,

respectively, which are twice as much as those observed from previous analyses based on

W36x150 beam. An early local buckling of the flanges, as well as the lateral torsional buckling

might be responsible for such accelerated strength degradation. Figure 3.13 and 3.14 present the

buckling shape of the assembly at 5% story drift level. It is apparent that the buckling of the

flange is much more extensive with a non-compact flange than the compact one. However, even

in the case with a non-compact beam, after considerable local buckling and distortion of the RBS

hinge, the column did not develop twisting.

Torque=(Flange Force)x( Eccentricity.)


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Figure 3.12. Load-displacement curve of the assembly with W30x90 beam and W27x194 column


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Figure 3.14 Buckling shape of the assembly with W30x90 beam and W27x194 column

(the top flange view).

Note: 1 kN= 0.225 kips, 1mm=0.0394 inch.

Figure 3.15. Cyclic Behavior of Connection with W14x426 Column and W36x150 Beam


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3.3.4. Lateral Stability of the Connection with W14 Column

For comparison with deep-column connections, an RBS connection assembly with

W14x426 column and W36x150 beam was used. Four cases were investigated. The first case,

named as ABQ-Fu, involved no RBS. Other three cases involved RBSs with different

eccentricities and flange reduction rates. The eccentricity is measured from the column flange face

to the near end of the RBS, and the flange reduction rate is the ratio of the cut flange area of the

smallest RBS to the original flange area. Figure 3.14 shows analytical and experimental responses

of the assembly with ABQ-e1 RBS. There was practically no strength reduction visible from the

load-displacement curve. The deformed shapes of the four cases are given in Figure 3.15. There is

no lateral torsional buckling in all but one case. The case with a large eccentricity RBS suffered

lateral torsional buckling primarily due to a distant RBS from the column. In none of the cases,

there was any torsion or weak-axis flexural deformation visible in the column.

Figure 3.16. Deformed Shapes of Connections with W14x425 Column and W36x150 Beam: (a)

No RBS; (b) Small eccentricity and moderate flange reduction RBS; (c) Large

eccentricity and moderate flange reduction RBS; and (d) Moderate eccentricity and

large flange reduction RBS


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In order to compare behavior of connections with W14 and deep columns, a connection

with W14x426 also was analyzed. The beam at this connection was the same as the others, a

W36x150. Figure 3.15 shows cyclic moment-rotation behavior of this connection established by

non-linear finite element analysis. The connection was analyzed with and without the bracing

provided by floor slab. In addition, a third case was also analyzed where the beam did not havethe RBS. The analyses indicated that in this case, presence or absence of floor slab did not make

much difference. The RBS area of the beam did not move laterally and the column did not show

tendency to twist as shown in Figure 3.16.

It appears that in this case, the W14 column alone, because of its large stiffness in torsion

and lateral bending, was able to brace the RBS hinge and prevent its lateral movement. This may

be the reason why in more than 100 tests of connections conducted within the SAC Program, and

almost all were without the slab, very few specimens showed tendency for column twisting. As a

result, the SAC tests using W14 columns, by default, ended up being valid tests even though there

was no floor to brace the beam. Simply put, the column alone provided the bracing. However, incase of connections with deep columns, the columns were not able to provide the bracing that the

floor normally provides. As a result, the RBS area of these specimens moved in lateral direction

causing twisting of column making these tests somewhat unrealistic and the results questionable.

Based on studies summarized in previous sections, it can be concluded that the twisting of

the deep columns during the tests conducted by Gilton et al (2000) most likely was the result of

the way the tests were done rather than a realistic behavioral phenomenon. The test specimens did

not have the lateral bracing provided by the floors that exists in almost all steel structures. Had

Gilton, Chi and Uang (Gilton et al, 2000) done the tests with correct boundary conditions and

representative bracings, the results would have been realistic representation of actual condition inthe field and most likely the twisting of deep columns would have been negligible and non-

consequential to the behavior and design. This was clearly the case with tests done by Ricles,

Mau, Lu and Fisher (Ricles et al, 2000), where the boundary conditions in the test set-up were

correctly presented. No twisting of deep columns were reported for deep column specimens

tested by Ricles et al.

Currently, a series of cyclic tests on RBS moment connections with deep columns is in

progress at Lehigh University by Professor Ricles and his research team. The results of such

tests, expected to be done properly as the earlier tests at Lehigh (Ricles et al, 2000) and the

design recommendations stemming from such results, will be a valuable addition to the field.


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4.CONCLUSIONS

4.1. Introduction

Based on the results of non-linear analyses of steel moment frames with RBS connections

and with W14 through W33 columns, the following conclusions were reached. The conclusions

herein should not be used or relied upon for any specific application without competent

professional examination and verification of its accuracy, suitability, and applicability by a licensed

professional engineer, designer or architect. As indicated in the Disclaimer section, anyone

making use of the information herein assumes all liability arising from such use.

4.2. Conclusions

1. Based on the observed performance of the frames with deep columns and the behavior of

their connections, there were no considerable reasons found to suggest preventing the use

of deep column sections in any moment frame including special moment frames.

2. The inelastic analyses of connections with deep columns indicated that the study

connections should be able to provide the required strength and especially the rotational

ductility in excess of those required by FEMA-350 (2001) for pre-qualified connections.

Figure 4.1 shows the FEMA requirement for minimum moment-rotation envelope curve(curve OYF) as well as representative envelop curve for connections with deep column

studied herein (curve OYA). As the figure indicates, the connections with deep column

clearly satisfy the FEMA requirement.


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3. In reference to deep columns, FEMA-350 (2001), Page 2-23, states: The pre-qualified

connections should only be used with W12 and W14 column sections. According to

FEMA-350, this statement is based on the results of only two tests of deep column

specimens that were done at the time of development of FEMA reports. In these two

tests, the deep columns showed a tendency to twist. A critical review of the test set-up, as

discussed in previous sections, revealed that most likely such column twisting would not

have occurred had the test set-up and the specimens been realistic representative of actual

buildings. The specimens had no transverse beams connected to the panel zone of the

columns and had no floor slabs. Almost all moment frame steel structures have floors(typically steel deck/concrete slab) and transverse beams, which provide significant lateral

bracing. This investigation indicated that presence of the floor was enough to provide

necessary bracing and to eliminate or to reduce the column twisting to insignificant and

non-consequential levels.

4. The cyclic behavior of RBS connections with deep columns was found to be similar to the

behavior of the same connection with W14 columns. Our studies indicated that there is no

difference in bracing requirement for RBS connections with W14 and deep columns of up

to W33 when there is a floor slab at least on one side of the beam.

5. By using deep columns, in a moment frame, the drift limits can be met with less steel

tonnage compared to W14 column sections. This is due to considerably large moment of

inertia of deep sections for the same weight per foot as a comparable W14 column.

Figure 4.1. Comparison of the M-Curve of Connections with Deep Columns to

the M-Required by FEMA for Special Moment Connections


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6. An added advantage of using deep column is a potential for saving in the cost of material

and construction. In the 10-story study frames, the weight of the steel using W27 deep

columns was about 1.3 lbs/ft2less than the steel in the same frame but with W14 columns.

According to a leading steel fabricator, the 1.3 lbs/ft2equals to about 6-8% in total

material saving based on 16-18 psf of steel for a typical structure of this type. Of courseas mentioned earlier, this 10-story building was just an example to demonstrate that using

deep columns instead of W14 not only can result in increasing lateral load resisting

strength, decreasing drift, and reducing the cost. In other cases, the amount of saving may

vary but most likely still there will be some economic gain in using deep columns.

7. The specimens without floor bracings, Figure 4.2(a), tested by Gilton, Chi and Uang

(2000), cannot be considered representative of the actual structures. Design procedures

and recommendations based on such test results cannot be justified. Future testing of the

connections with deep columns need to be done such that the bracing effects provided by

the floors and transverse beam(s) are represented. An example is shown in Figure 4.2(b). .

Figure 4.2. (a) Unrealistic Test Set-up used by Gilton, Chi, Uang (2000) and (b) Realistic Set-up

(a) (b)


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________________________________________________________________________

References________________________________________________________________________

ABAQUS (2001), User ManualI, II and III, Version 6.2, Hibbitt, Karlsson & Sorensen, Inc.,

Providence, RI.

AISC (1998),Load and Resistance Factor Design Specification for Structural Steel Buildings,

American Institute of Steel Construction, Chicago, IL.

AISC (2002), Seismic Provisions for Structural Steel Buildings, (in review at this writing),

American Institute of Steel Construction Inc., Chicago, IL.

Astaneh-Asl, A. (1995), Seismic Behavior and Design of Bolted Steel Moment-Resisting

Frames, SteelTIPS, Structural Steel Educational Council, Moraga, CA.

(This report can be downloaded free from www.aisc.orgweb site.)

Bolt, B. and Gregor, N., (1993), Synthesized Strong Ground Motion for the Seismic Condition

Assessment of the Eastern Portion of the San Francisco Bay Bridge, Report No.

UCB/EERC-93/12, University of California, Berkeley, CA.

FEMA-350 (2001), Seismic Design Criteria for Steel Moment-Frame Structures,Report,

Federal Emergency Management Agency, MD.

(This report can be downloaded free from www.fema.govweb site.)

Flynn, L., (2000), Letter to the Editor,Modern Steel Construction, American Institute of Steel

Construction, November, Chicago, IL.

Gilton, C., Chi, B. and Uang, C. M. (2000), Cyclic Response of RBS Moment Connections:

Weak-Axis Configuration and Deep Column Effects, Report No. SSRP-2000/03,

Structural Systems Research Project, Department of Structural Engineering, University of

California, San Diego, La Jolla, CA.

ICBO (1997), Uniform Building Code, International Conference of Building Officials, Whittier,

CA.


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Kitjasateanphun T. (2001), Seismic performances of Reduced Beam Section Frames, Ph.D.

Thesis, Department of Civil and Architectural Engineering, Illinois Institute of

Technology, Chicago, IL.

Moore, K.S., Malley, J.O., and Engelhardt, M.D., (1999), Design of Reduced Beam Section(RBS) Moment Frame Connections, Steel TIPS, Structural Steel Educational Council,

Moraga, CA.

(This report can be downloaded free from www.aisc.orgweb site.)

Ricles, J.M., Mao, C., Lu, L-W and Fisher, J.W., (2000) Development and Evaluation of

Improved Details for Ductile Welded Unreinforced Flange Connections,ATLSS Report

No. 00-04, ATLSS Engineering Research Center , Lehigh University, Bethlehem, PA.

SAC, (1996), Northridge Model Buildings,Internal Report for SAC Researchers, SAC JointVenture, Sacramento.

SCI, (2000), Structural and Earthquake Engineering Software, SAP-2000 Software, Computers

and Structures, Berkeley.

SDI (1989), LRFDDesign manual for Composite Beams and Girders with Steel Deck,No.

LRFD1, The Steel Deck Institute.

SEAOC, (1999), Recommended Lateral Force Requirements and Commentary, Seventh Ed.,

Structural Engineers Association of California, Sacramento, CA.


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About the authors.

Jay Shen, Ph.D., P.E., S.E. is an associate

professor of structural engineering at the

Illinois Institute of Technology, Chicago. He

is a registered Professional Engineer and

Structural Engineer. Dr. Shen's expertise and

research interests are in the areas of inelastic

behavior of steel structures and earthquakeengineering. Topics of research interests

include: cyclic behavior and design of

earthquake-resistant steel structures, dynamic

analysis, seismic retrofit of bridges, and

computer-integrated analysis and design of

steel structures. His current research

includes nonlinear finite element analysis of

steel bridges subject to strong ground

motions, and seismic study of special moment

frames and semi-rigid steel frames. Since

1994 Northridge California earthquake, he

has conducted considerable research on

inelastic behavior of moment connections,

particularly pre-Northridge welded and the

post-Northridge RBS steel moment

connections.

He has also been involved in providing

consulting and advice to the industry on

behavior and design of steel structu res.

He can be reached at:

Ji e-Hua Jay Shen, Ph.D., P.E., S.E.Department of Civi l and Arch. Engin eeri ng,

I ll inoi s I nstitute of Technology,

3201 South D earborne Street,

Chi cago, IL , 60616.

Phone: (312) 567-5860, Fax: (312) 567-3579.


Abolhassan Astaneh-Asl, Ph.D., P.E., is a

professor of structural engineering at the

University of California, Berkeley.

He is the winner of the 1998 AISC, T.R.

Higgins Award.

Since 1982, he has been involved in

teaching, research and design of steel

structures. In recent years, he hasconducted several major projects on

seismic design and retrofit of steel long

span bridges and tall buildings. Since1995,

he has also been studying behavior of steel

structures subjected to blast loads and has

been involved in testing and further

development of a cable- based mechanism

to prevent progressive collapse of steel

structures. The original concept of the

system was suggested by Dr. Joseph

Penzien in 1996 and in the aftermath of

terrorist attack on Murrah building in

Oklahoma City.

Since September 11, 2001, he has been

heavily involved in conducting research,

funded by the National Science

Foundation, on the collapse of the World

Trade Center due to terrorist attack.


Abolhassan Astaneh-Asl, Ph.D., P.E.,

781 Davis Hall , Uni versity of Calif ornia,

Berk eley, CA 94720-1710

Phone: (510) 642 4528, Fax: (510) 643 5258

Home off ice Phone and Fax: (925) 946-0903

Cell Phone for U rgent Calls: (925) 699-3902

E-mail: [email protected]

David McCallen, Ph.D, is the Director of

the Engineering Technology Center for

Complex Distributed Systems at LLNL. The

Center is responsible for developing

Engineering's capabilities in agile

distributed sensor networks for data

gathering and advanced techniques forcombining simulation and sensing for

enhanced characterization of complex

systems. McCallen has a Ph.D. in Structural

Engineering and Structural Mechanics from

the University of California at Davis. His

expertise is in the area of structural

dynamics and the response of structures to

extreme events. He has collaborated at the

University of California (UC), Berkeley as a

Visiting Research Engineer and has

performed recent work in earthquake

simulations with a multidisciplinary team of

earth scientists and engineers from the UC

Berkeley and LLNL.

He has also led LLNL projects for the

California Department of Transportation

that studied the seismic response of key

California transportation structures.


David B. McCall en, Ph.D.,

Center for Complex Distri buted Systems,

Lawrence L ivermore Nati onal Laboratory,

7000 East Avenue, MS L -151,

L ivermore, CA 94550.

Phone: (925) 423-1219



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List of Published Steel TIPS Reports*------------------------------------------------------------------------------------------------------------------------------------------

June 02: Use of Deep Columns in Special Steel Moment Frames, by Jay Shen, Abolhassan Astaneh-Asl and

David McCallen.

May 02: Seismic Behavior and Design of Composite Steel Plate Shear Walls, by Abolhassan Astaneh-Asl.Sept. 01: Notes on Design of Steel Parking Structures Including Seismic Effects, by Lanny J. Flynn, and

Abolhassan Astaneh-Asl.

Jun '01: Metal Roof Construction on Large Warehouses or Distribution Centers, by John L. Mayo.

Mar. '01: Large Seismic Steel Beam-to-Column Connections, by Egor P. Popov and Shakhzod M.Takhirov.

Jan 01: Seismic Behavior and Design of Steel Shear Walls, by Abolhassan Astaneh-Asl.

Oct. '99: Welded Moment Frame Connections with Minimal Residual Stress, by Alvaro L. Collin and James J.

Putkey.

Aug. '99: Design of Reduced Beam Section (RBS) Moment Frame Connections, by Kevin S. Moore, James O.

Malley and Michael D. Engelhardt.

Jul. '99: Practical Design and Detailing of Steel Column Base Plates, by William C. Honeck & Derek Westphal.

Dec. '98: Seismic Behavior and Design of Gusset Plates, by Abolhassan Astaneh-Asl.

Mar. '98: Compatibility of Mixed Weld Metal, by Alvaro L. Collin & James J. Putkey.

Aug. '97: Dynamic Tension Tests of Simulated Moment Resisting Frame Weld Joints, by Eric J. Kaufmann.

Apr. '97: Seismic Design of Steel Column-Tree Moment-Resisting Frames, by Abolhassan Astaneh-Asl.

Jan. '97: Reference Guide for Structural Steel Welding Practices.

Dec. '96: Seismic Design Practice for Eccentrically Braced Frames (Based on the 1994 UBC), by Roy Becker &

Michael Ishler.

Nov. '95: Seismic Design of Special Concentrically Braced Steel Frames, by Roy Becker.

Jul. '95: Seismic Design of Bolted Steel Moment-Resisting Frames, by Abolhassan Astaneh-Asl.

Apr. '95: Structural Details to Increase Ductility of Connections, by Omer W. Blodgett.

Dec. '94: Use of Steel in the Seismic Retrofit of Historic Oakland City Hall, by William Honeck & Mason Walters.

Dec '93: Common Steel Erection Problems and Suggested Solutions, by James J. Putkey.

Oct. '93: Heavy Structural Shapes in Tension Applications.

Mar. '93: Structural Steel Construction in the '90s, by F. Robert Preece & Alvaro L. Collin.

Aug. '92: Value Engineering and Steel Economy, by David T. Ricker.Oct. '92: Economical Use of Cambered Steel Beams.

Jul. '92: Slotted Bolted Connection Energy Dissipaters, by Carl E. Grigorian, Tzong-Shuoh Yang & Egor P.

Popov.

Jun. '92: What Design Engineers Can Do to Reduce Fabrication Costs, by Bill Dyker & John D. Smith.

Apr. '92: Designing for Cost Efficient Fabrication, by W.A. Thornton.

Jan. '92: Steel Deck Construction.

Sep. '91: Design Practice to Prevent Floor Vibrations, by Farzad Naeim.

Mar. '91: LRFD-Composite Beam Design with Metal Deck, by Ron Vogel.

Dec. '90: Design of Single Plate Shear Connections, by Abolhassan Astaneh-Asl, Steven M. Call and Kurt M.

McMullin.

Nov. '90: Design of Small Base Plates for Wide Flange Columns, by W.A. Thornton.

May '89: The Economies of LRFD in Composite Floor Beams, by Mark C. Zahn.

Jan. '87: Composite Beam Design with Metal Deck.Feb. '86: UN Fire Protected Exposed Steel Parking Structures.

Sep. '85: Fireproofing Open-Web Joists & Girders.

Nov. '76: Steel High-Rise Building Fire.

The Steel TIPS are available at AISC website: www.aisc.organd can be downloaded free for

personal use courtesy of the California Field Iron Workers Administrative Trust and the AISC.


41/41

P.O. Box 6190

Moraga, CA 94570

Tel. (925) 631-1313

Fax. (925) 631-1112

Fred Boettler, Administrator

Steel TIPS may be viewed and downloaded at www.aisc.org

S P O N S O R SAdams & Smith Four Star Erectors Plas-Tal Manufacturing Co.

Bannister Steel, Inc. Gayle Manufacturing Reno Iron Works

Baresel Corp The Herrick Corporation SME Industries

Bethlehem Steel Corporation Hoertig Iron Works Schollenbarger-Borello, Inc.

Bickerton Industries, Inc Junior Steel Company Strocal Inc.

Bostrum Bergen. Martin Iron Works Inc. Templeton Steel Fabrication

California Erectors McLean Steel Inc. Trade Arbed

Eagle Iron Construction Nelson Stud Welding Co. Verco Manufacturing, Inc

Eandi Metal Works Oregon Steel Mills Vulcraft Sales Corp.

Western Steel & Metals, Inc.

Funding for this publication was provided by the California Field Iron Workers Administrative Trust.

STRUCTURAL STEEL EDUCATIONAL COUNCIL

Steel

The local structural steel industry (above sponsors) stands ready to assist you in determining the most economical

solution for your products. Our assistance can range from budget prices and estimated tonnage to cost comparisons,

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Special Steel Moment Frame