Offshore Geotechnics

Technical Committee 209Offshore Geotechnics

Comité technique 209Géotechnique marine

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General Report of TC209 Offshore Geotechnics

Rapport Général du TC209 Géotechnique Offshore

Jewell R.A. Fugro GeoConsulting

ABSTRACT: This general report introduces the discussion session organized by ISSMGE Technical Committee 209 (TC209) “Offshore Geotechnics”. The main topics include offshore wind projects, pipelines and seabed structures, seabed soils, coastal andnearshore work, and pile foundations.

RÉSUMÉ : Ce rapport général introduit la séance de discussion organisée par le Comité Technique 209 (TC209) “Géotechnique Offshore” de la SIMSG. Les principaux thèmes abordés sont les projets d’éoliennes offshore, les pipelines et structures sous-marines, les sols sous-marins, les travaux côtiers et nearshore et les fondations sur pieux

KEYWORDS: offshore, caisson, piles, pipes, cyclic load, stability diagram, lateral load, tests, numerical analysis, bearing capacity

1 INTRODUCTION.

The organizers of the 18th International Conference “Challenges and Innovations in Geotechnics” have implemented important changes to the conference format. One is the inclusion of Offshore Geotechnics at this main ISSMGE forum. Second is the focus given to the technical committees.

This general report covers the session organized by the ISSMGE Technical Committee 209 (TC209) “Offshore Geotechnics” chaired by Philippe Jeanjean. Participation by TC209 at this 18th conference includes the 2nd ISSMGE McClelland Lecture by Mark Randolph, this discussion session and a workshop on recent research and development on piles under cyclic loading.

The main difference in offshore geotechnics arises from the conditions and environment offshore. There is a stark contrast in access for site investigation, soil sampling, field testing, installation and observation. Activities offshore often require new tools. Soft soil conditions at seabed level are encountered in deepwater, frequently with high carbonate content, unusual mineralogy or biogenic activity. Combined and cyclic loading usually dominate design, whether caused by waves and currents acting on structures or by repeated expansion and contraction of pipelines on the seabed. Large displacement is a feature of the installation and operation of seabed pipelines. Many of the above features of offshore geotechnics are discussed in papers to this session.

This general report has been organized into five main subject areas: Offshore Wind; Pipelines and Seabed Structures; Seabed Soils; Coastal and Near Shore work; Pile Foundations.

Since papers on the cyclic loading of piles will be presented and discussed at the TC209 workshop, these are highlighted in this general report but will not be presented during the discussion session. The main focus for presentations will be Offshore Wind and large displacement as encountered with Offshore Pipelines.

Only a limited selection of papers will be presented at the discussion session. All the papers are in the proceedings and many will be presented at the poster session. Participants are strongly encouraged to attend the TC209 workshop where the cyclic behavior and design of piles will be presented and debated based around papers to this session.

2 OFFSHORE WIND.

2.1 Site investigation

Project development, engineering design and project construction are three main phases for offshore wind farms. A major challenge is to minimize geotechnical risk for foundation engineering. In current practice, geotechnical risk is addressed mainly during the engineering design phase P2 in Table 1. Ben-Hazzine and Griffiths (2013) suggest that risk management may be improved through more extensive geophysical survey and preliminary site investigation during the project development phase P1. The authors highlight various sources of geotechnical risk such as inherent soil variability, measurement errors and “transformation errors” caused by simple empirical interpretation of data. Table 1. Timing of Geotechnical Work for Offshore Windfarms ( Ben-Hazzine and Griffiths, 2013)

Muir Wood and Knight (2013) use experience from 15 offshore wind farm projects in the UK to illustrate the manner in which geotechnical risk was managed and to define categories of poor, mediocre and good practice (or vice versa).

Poor practice includes appointment of the foundation design team after site investigation is completed, insufficient planning and poor interpretation of geophysical and geotechnical surveys. Mediocre practice often involves incorrect scope for geophysical and geotechnical surveys causing extra cost and increased risk. On projects with good practice the foundation design team was appointed at the start of the project and specified the site investigation work. The ground models were

Rapport général du TC209Géotechnique marine

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Proceedings of the 18th International Conference on Soil Mechanics and Geotechnical Engineering, Paris 2013

developed and refined over multiple phases of geotechnical and geophysical investigation. The authors conclude that a formal approach to risk management with staged investigations and early appointment of the foundation design team is the best practice.

2.2 Wind turbine performance

A wind energy project involves several disciplines within civil engineering. The complex interaction between the turbine and the supporting structure governs the dynamic behaviour. Analysis is required of the structural modal parameters that influence fatigue of a wind turbine. Indeed, continuous monitoring of modal frequencies and damping ratios during the operational life of a wind turbine can provide early warning of the onset of structural damage.

Damgaard et al. (2013) assess instrumentation data from 30 offshore wind turbines in the North Sea. The measured modal frequency and damping ratio are found to vary with time. However, the observed magnitude and pattern of change with time might result from scour erosion or backfilling around the monopile structure. This highlights that the dynamic interaction between an operating turbine and the supporting structure may be affected by changes with time of the local seabed level, something that could be reduced by scour protection.

2.3 Foundation systems

The power capacity of a wind turbine, typically in the range 2MW to 5MW, but increasing toward 10MW in future, determines the required height above mean sea level and the maximum horizontal and vertical loads to be supported. Deeper water increases the moment arm for both wind and wave loading and often signifies larger design storms. There are many novel features to the required foundation engineering as discussed in papers to this session, Table 2.

T able 2. Papers on Foundation Systems for Offshore Wind.

SINGLE LOAD

CYCLIC LOAD

Shallow Foundation

Arroyo et al A

Monopile Roesen et al E.1g; A Caisson Kim et al E.c; A Versteele et al A Hybrid (pile/caisson)

Arshi et al E.1g; A

E = Experimental (E.1g lab floor; E.c centrifuge); A = Analysis In brief résumé, shallow foundations and monopiles

generally provide efficient support for wind turbines. However, these foundations are less effective when the moment load to be resisted increases due to larger turbines and/or deeper water. One approach is to improve the capacity by combining a monopile and shallow foundation, Figure 1. Alternatively, a caisson can support combined vertical, horizontal and moment loading at seabed level. Three or four caissons may be combined to support a structure where greater load carrying capacity is required, Figure 2.

The issues for foundation engineering discussed in the papers include: (1) the impact of the loading path and loading direction on safety factor and use of a single failure envelope; (2) improved performance by combining a shallow foundation and monopile; (3) the incremental displacement of cyclically loaded monopiles; and, (4) assessment of simultaneous pore water pressure generation and dissipation for caissons under storm loading.

(Arshi et al, 2013) Figure 1. Hybrid Monopile and Shallow Foundation.

(Versteele et al, 2013) Figure 2. Monopod and Multipod Foundations.

2.3.1 Bearing capacity analysis The assessment of bearing capacity for offshore wind turbine foundations differs from onshore practice in several respects. The issues include: (a) separate correction factors in analysis for shape, depth, load inclination and eccentricity that are cumbersome and prone to calibration error; (b) use of separate partial factors on loads and resistances as in DNV-OS-J101 when the difference between favourable and detrimental effects can be subtle; and (c) simultaneous application of two major horizontal loads from wind and wave acting in separate directions. Arroyo et al highlight these issues and question the suitability of the conventional design approach for offshore wind foundations (Arroyo et al 2013).

A more satisfactory framework for capacity checks would be through failure-envelopes as detailed in the paper. Arroyo et al examine a synthetic design example to illustrate their case using the geometry of a Thornton Bank GBS and a set of derived loading parameters, Figure 3. The complex interaction between horizontal and moment loading, and the impact of different directions for wind and wave loading are illustrated by the authors using Figure 4 and the results in Table 3.

It is conventional to increase both detrimental loads at the same time, so that the load increment causing the limit to be reached is in the same direction as the reference combined load. The authors consider the cases where only wind or wave loading is increased. Such analysis might be used to assess the impact of any error in the assessment of those loads. Because of the greater influence on moment of wind loading, the analysis shows that an underestimate 21% (in this case) in the wind load would be sufficient to cause failure compared with an underestimate 50% for the wave load, Table 3.

Arroyo et al (2013) conclude that failure envelopes offer a powerful framework for analysis of shallow foundation capacity; the approach is particularly well suited for offshore wind structures that require refined design in the face of considerable uncertainty.

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Figure 3. Thornton Bank GBS (after Peire et al 2009).

(Arroyo et al 2013) Figure 4. Incremental loading paths to failure.

Table 3. Results of analysis on incremental load to failure.

2.3.2 Hybrid foundations When considered as a monopile design, the addition of a shallow footing at seabed level may be thought of as adding “fixity” to the monopile “head” thereby generating improved resistance and stiffness to lateral loading, Figure 1. A simplified design analysis would assess the limiting moment capacity of the shallow foundation acting alone and include the equal and opposite moment resistance to the analysis of the monopile. The shallow foundation not only increases load carrying capacity but also reduces the bending moment supported by the monopile, by about 25% in the example cited by Arshi et al (2013).

Experimental modelling of these hybrid systems at 1g and in the centrifuge are described together with numerical and analytical work. The significance of the geometric ratio of footing to pile diameter, and pile length to pile diameter is demonstrated and a form for design charts proposed, Figure 5. Some tests on caisson/monopile combinations are noted and indicate additional benefit due to the lateral load resistance of

the caisson. Centrifuge tests are currently underway to define better the benefit of caisson versus simple shallow foundation in such hybrid foundation systems.

(Arshi et al 2013) Figure 5. Moment resistance chart for hybrid foundations.

2.3.3 Lateral displacement due to cyclic loading

The focus above is the limit resistance of offshore wind tower foundations subject to a single application of combined loading. In practice, the structures are subject to several episodes of extreme loading caused by major storms and a great number of cycles of low amplitude loading from normal wind and wave conditions. The latter source of repeated loading may cause fatigue or serviceability problems (Roesen et al. 2013).

The authors report a series of 1g laboratory tests on monopiles in sand subject to one-way cyclic loading over more than 50,000 cycles. One limitation in these 1g tests is a more rigid pile compared with a typical prototype, but the trend of results should be similar. The cyclic loading is described by two non-dimensional ratios: the maximum moment compared with the static moment capacity ζb = Mmax/Mstatic in the range 1 > ζb > 0; and the ratio of the minimum to the maximum moment ζc = Mmin/Mmax which has a value 1 for a static test, 0 for one-way loading (the case examined by Roesen et al) and -1 for two way cyclic loading.

The pile displacement is measured by the rotation θ at the soil surface. The results of a static load test and the measured displacement in one-way cyclic load tests (ζc = 0) with load intensity ζb = 0.2 to 0.4 are shown on Figure 6.

(Roesen et al 2013) Figure 6. Static and cyclic one-way loading tests on model monopiles.

The incremental rotation due to one-way cyclic loading may

be quantified with respect to the rotation caused by the first, single loading ΔθN = θN – θ1 . Tests with different loading intensity may then be compared through the non-dimensional form ΔθN/θ1 as shown on Figure 7.

The test data are compared with a simple power law (ΔθN/ θ1) = aNb, where a and b are empirical constants found from testing. The power law seems to provide a reasonable asymptotic limit for the data after about 1000 or more load repetitions, Figure 7. The value of the constant a is found to vary almost linearly with the applied load magnitude ζb . The slope of accumulating displacement with repeated loading, the constant b, appears to be a function of the combination of

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monopile and sand parameters, presumably including the installation process. The rate of accumulation appears not to depend markedly on the applied load magnitude. The analysis in §6.2 suggests that the data might also fit with a natural logarithm through to low numbers of cycles (ΔθN/ θ1) = c.ln(N).

(Roesen et al 2013) Figure 7. Monopile rotation versus number of cycles.

2.3.4 Caisson foundations A centrifuge test of a caisson in sand is reported by Kim et al. (2013). The caisson response to single combined load to failure is measured and numerical analysis applied. The test details are provided elsewhere and it is not clear whether the 1/70th scale caisson was installed by suction during the centrifuge test or before testing. Soil material from the proposed site is used to model a planned prototype caisson foundation. The measured response of the modelled prototype 15.5m diameter 10.5m long caisson is shown on Figure 8 in terms of applied moment versus rotation.

The authors report a parametric analysis using FLAC to show the significant influence of the assumed elastic modulus and cohesive strength parameters assumed for the soil.

(Kim et al 2013) Figure 8. Measured moment-rotation for prototype caisson.

The offshore design standard DNV-OS-J101 (DNV, 2011) requires structures to resist a 50 year design storm considering both peak loads and the entire history of cyclic loading. It is this latter requirement that is tackled by Versteele et al (2013) for the case of caisson foundations in sand.

Because a full analysis of cyclic loading of caissons in sand is not practically feasible with current numerical methods, the authors develop an analysis to provide insight into the competing processes of excess pore water pressure generation and dissipation during the design storm. The analysis breaks the storm into several packages of cyclic loading (magnitude, number of cycles and time). The excess pore water pressure generated at each point in the soil by the package of cyclic loading is computed analytically and input into the numerical analysis. The dissipation and redistribution of pore water pressure during the time period is computed numerically. The

process is repeated for the next package of cyclic loading, and so on.

The analysis for pore pressure generation uses two relations for the sand material. First the measured cyclic shear strength versus number of cycles to liquefaction, Nl, from cyclic laboratory tests; second an empirical formula linking pore water pressure generation to number of load cycles, Figure 9. Liquefaction occurs at u/umax = 1 when N/Nl = 1.

(Versteele et al 2013) Figure 9. Generation of excess pore water pressure.

The results of a 3D analysis for a caisson foundation are

reported to illustrate the method, Figure 10. The 20m diameter by 10m length caisson is subjected to a 6 hour design storm of 2160 waves. These are split for analysis into 5 individual load packages. The direction of wind and wave loading is assumed to be aligned. The results illustrate the asymmetric nature of pore water pressure generation that has potential consequences for possible differential settlement and tilting of the caisson.

(Versteele et al 2013) Figure 10. Example of excess pore water pressure below a caisson.

Versteele et al (2013) conclude that the model is useful in

predicting areas beneath the caisson prone to the development of excess pore pressure. However, the analysis does not predict liquefaction behaviour or compute settlement, nor does it allow for load redistribution in the caisson due to the changing effective resistance in the soil during the design storm. There is further development work to be done.

3 PIPELINES AND LARGE DISPLACEMENT.

A challenging feature for offshore pipelines is the large displacement that can occur during installation and service. Large displacement is particularly extreme for laying pipe on a soft seabed. Large displacement also results from multiple cycles of heat expansion and contraction of the operating pipeline. This requires engineering design to avoid localized

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pipe distortion and over-stress irrespective of the sea bed soil type. A less discussed cause of large displacement is where a pipeline crosses a seismic fault; here it is movement of the ground with respect to the pipe that causes gross distortion. Fault crossings occur both onshore and offshore.

The geotechnical analysis for large displacement requires suitable tools and numerical models and both have developed significantly in recent years. A variety of large displacement numerical methods with 3D capability are commercially available for design purposes. Similarly, constitutive models for soft clay that account for competing strain rate and strain softening effects, and competing pore pressure generation and dissipation, are available for designers. Specific numerical elements to model the large displacement interaction between a pipe and the surrounding soil are also currently under development for practical application in design (SAFEBUCK JIP). However, the constitutive and numerical modeling for large displacement in dense sand is less well advanced.

Pipe buckling and pipe walking is usually assumed to occur between fixed seabed structures. There may be scope for permitting the seabed structures to move horizontally to help accommodate axial pipe displacement.

Several of these topics are described in four papers to the discussion session.

3.1 Dynamic embedment of offshore pipelines

Dutta et al (2013) examine pipe laying and dynamic embedment using Coupled Eulerian Legrangian (CEL) methods available in ABAQUS software. Progressive degradation of undrained shear strength with plastic shear strain is included using the model of Einav and Randolph (2005). Similar analysis by Wang et al (2010) used remeshing and small strain (RITSS analysis).

The simplified problem is illustrated in Figure 11. A pipe is penetrated monotonically into a soft clay sea bed under self-weight (submerged weight of pipe). The pipe is then cycled laterally by a displacement u/D = ± 0.05, in the x direction of Figure 11, under constant self-weight vertical load. This causes additional pipe penetration.

(Dutta et al 2013) Figure 11. Pipe penetration of a seabed.

The analysis uses the same dimensions, soil parameters and

loading sequence as the first stage of two pairs of centrifuge tests by Cheuk and White (2008) on a light and heavier pipe. The progressive pipe penetration and magnitude of horizontal resistance caused by cyclic lateral displacement is computed. The penetration of the pipe with cycles of lateral displacement is shown in Figure 12 for one pair of pipe tests.

Current practice is to estimate separately the embedment due to pipe laying and due to dynamic effects. The initial embedment on pipe laying involves temporary overload at the touch down point. Typically the pipe weight is increased by a “lay factor” and the initial pipe penetration under monotonic loading is computed for this higher load. The effect of small amplitude cyclic lateral motion is incorporated using a “dynamic embedment factor” that multiplies up the initial monotonic pipe lay embedment to determine a final estimated pipe embedment. The centrifuge tests and analysis here did not

incorporate initial overloading of the pipe, but the forty cycles of lateral loading resulted in a dynamic embedment factor of the order 4 to 5, within the range often assumed in practice.

(Dutta et al 2013) Figure 12. Static and dynamic pipe penetration of a seabed.

As shown previously by Wang et al (2010), the analysis provides insight into the size of the zone of highly sheared and softened soil around the pipe and the shape of the berms formed by pipe penetration. The results in Figure 13 are for the heavier pipe and show dynamic pipe penetration and monotonic pipe penetration to the same depth (increased vertical load). The comparison is striking. Dynamic embedment causes more extensive plastic strain softening in the soil, coloured red, and wider and flatter berms than generated by monotonic pipe penetration. The latter could be important for the analysis of initial lateral breakout of the pipe. Dynamic embedment affects the magnitude of pipe penetration, the zone of soil remoulding and the shape of the berms formed.

(Dutta et al 2013) Figure 13. Dynamic and monotonic pipe penetration of a soft seabed.

3.2 Pipeline fault crossing

Damage is caused to pipelines that cross a seismic fault that subsequently displaces. Rupture of oil or gas pipelines can cause fire and environmental risk. For critical pipelines, the magnitude and direction of localized fault displacement should be assessed and appropriate engineering implemented to avoid pipe rupture due to ground movement.

Moradi et al completed centrifuge tests on buried steel pipe subject to an upward thrust fault at 30⁰ to the vertical in the direction of the pipe. A fault displacement 70mm was applied across an 8mm diameter buried pipe with 0.4mm wall thickness tested in a centrifuge at 40g. In one test the pipe is simply buried in the compacted sand. A low density and light weight loose backfill was used in the second test. The axial and bending strain in the pipe was measured in both tests. The light backfill allowed the pipe to buckle and displace over a greater length considerably reducing the damage to the pipe. The pipe embedded in the sand suffered more localized deformation and damage, as illustrated by the photos post testing, Figure 14.

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(Moradi et al 2013) Figure 14. Pipe response to shear fault displacement in a centrifuge.

3.3 Seabed structures that displace horizontally

The sea bed in deep water is generally soft and often requires large shallow foundations to support seabed facilities. If some movement could be tolerated the size could be reduced. Further, if the structure connects with a pipeline subject to walking or other axial force, there may be merit in allowing the structure to slide horizontally to help relieve concentrated load.

Bretelle and Wallerand (2013) examine the design for a shallow foundation that displaces horizontally in a cyclic fashion, as might be caused by repeated pipe expansion and contraction. The influence of soil softening, foundation settlement and potential change in stiffness with time is examined through relatively straightforward analysis. The authors conclude that shallow foundations designed to displace horizontally could be useful for subsea pipeline networks.

3.4 Large displacement in dense sand

While numerical analysis for large deformation is increasingly amenable for engineering design, a relatively simple constitutive model for dense sand that provides stable large deformation analysis is still subject to study. Li et al (2013) propose a Critical State Mohr Coulomb (CSMC) model: deformation up to peak strength is elastic and thereafter dense sand dilates (including non-associated flow) and reduces in strength to the critical state angle of friction. The concept of the state parameter defined by Been and Jeffries (1985) is used.

A key objective is analysis for punch through of a spudcan footing in dense sand overlying soft clay. Li et al (2013) have not reached that target. However, development of the model starkly highlights non-uniform deformation and preferential shear band formation in dense sand post peak that makes data acquisition (lab tests) and model calibration such a challenge.

In analysis for bearing capacity of a circular plate on uniform sand, the authors found that the elastic stiffness of the sand influences bearing capacity by as much as 50% over the realistic range, reminiscent of rigidity index in penetration problems. Stiffness was found to have greater impact than dilation angle. The analysis for bearing capacity is described in terms of a combined bearing capacity factor Nqγ that applies across the range from Nq alone to Nγ. The proposed formula for Nqγ includes soil stiffness and dilation angle along with peak friction angle, foundation size, soil unit weight and surcharge.

4 SEABED SOILS.

The three papers on soil properties cover diverse topics. Ho et al (2013) describe undrained cyclic triaxial compression tests on isotropically consolidated Singapore Marine Clay. The focus is the behavior of the clay when it is sheared monotonically to failure after cycling. The tests show that when the current mean effective stress in the sample reduces below half the original preconsolidation pressure, p�/pc� ≤ 0.5, due to cyclic loading, some increase in mean effective stress commences at higher stress ratio in each cycle. At mean effective stress p�/pc� ≥ 0.6 (first few cycles) the mean effective stress of the clay always reduces. This behavior is similar to normally versus over consolidated clay. The final effective stress path for monotonic triaxial compression to failure after cycling similarly depends on the mean effective stress p�/pc� at the end of

cycling. However, the shear strength is found to be largely independent of the previous number of load cycles and the strain amplitude. Kim and Safdar (2013) report cyclic direct simple shear tests on compacted silty sand to define the limiting cyclic stress ratio versus number of cycles for two initial void ratios. Tyldesley et al (2013) describe site investigation to define parameters for wind farm foundation design on a deep deposit of carbonate clayey silt till in Ontario Canada. This onshore site investigation demonstrates the use of insitu tests and shear wave velocity measurement, interpreted together with laboratory tests, to assemble knowledge on soil strength and stiffness properties.

5 COASTAL AND NEARSHORE WORK.

There are three papers on diverse topics. Madrid et al (2013) describe site investigation, cyclic laboratory tests and numerical analysis for the stability of a caisson breakwater in about 20m to 25m depth of water. The caissons are founded on a rubble mound infilling a large zone where the deep underlying soft clay soil was removed, Figure 15.

CAISSON

RUBBLE MOUND

172.27

-26.00

RIP RAP 300kg

SEAWARD SIDE

RIP RAP 4 tonCONCRETE BLOCK

RIP RAP 300kg

0.00

-21.00-23.00

-13.75-15.00-13.00

-22.00

-18.00

0.00

+11.00

+3.00

(Madrid et al 2013) Figure 15. Caisson breakwater and stability analysis for wave impact.

There is much detail in the paper on soil testing and soil

properties, loading cases for various phases of project construction and hydrodynamic testing to determine dynamic uplift. A good description is provided on the way cyclic loading and shear strength reduction were treated for design.

Relic footprints from earlier jack-up activity can occur next to the location for new shallow foundations. Ballard and Charue (2013) describe a study on a circular zone of remoulded soft clay (Sr = 2) with a diameter equal to the size of the square mudmat and with soft clay thickness of half that size. The limiting envelope for combined moment and horizontal resistance is computed for a range of applied vertical load (V/Vult), and a range of distance between the mudmat and the remoulded zone/footprint that causes the moment and horizontal resistance to be reduced, as well as Vult. 2D and 3D analyses show very substantial benefit from the 3D geometry in this case.

A detailed design and project record for installation of large diameter, buried HDPE pipes in a nearshore environment prone to seismic loading is described by Bellezza et al (2013). Details for the case history and the various code requirements considered in design are documented. Initial measurements are provided on the vertical deflection of the installed pipes.

6 PILE FOUNDATIONS.

A lack of code guidance on capacity, stiffness and displacement for cyclically loaded piles is being addressed by collaborative research including the original GOPAL study and the current SOLCYP project, supplemented by individual research work. Several papers to this session report on SOLCYP results from instrumented field tests, calibration chamber and centrifuge

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tests. SOLCYP will be presented and discussed at the TC209 workshop and recorded for publication. Therefore only some key aspects are described below to avoid duplication.

There are four axial load magnitudes: the mean Qmean and the half-amplitude of the cyclic load Qcyclic define the maximum Qmax = Qmean + Qcyclic and the minimum Qmax = Qmean - Qcyclic pile loads. These loads are typically referenced to the ultimate pile capacity in tension QT or compression QUC. The ultimate capacity and the capacity under cyclic load is determined at a limiting displacement (0.1D or less) or due to an increasing rate of displacement; either continuing displacement after a static load increment or the cyclical displacement rate (mm/cycle).

6.1 Stability diagram: cyclic axial loading

The stability diagram is a non-dimensional map for cyclic pile behavior. The diagram in Figure 16 is for axial tension tests on model driven piles in dense sand in a calibration chamber (Silva et al, 2013). A similar diagram is found for the equivalent field test data (Rimoy et al, 2013). The chart defines the region of stable cyclic load combinations for a number of load cycles to be resisted. One way loading (Qmean>Qcyclic) is more stable. The dashed line shows the limit of valid load combinations.

Puech et al (2013) provide the equivalent stability diagrams for cyclic compression loading of bored piles in dense sand from field and centrifuge tests. The field tests by Benzaria et al (2013a) are shown on Figure 17. The results from centrifuge model tests are very similar. Note that only one-way loading was tested and the data should not be extrapolated beyond.

While bored piles in stiff clay have lower ultimate resistance compared with driven piles, cyclic compression tests on bored piles in over-consolidated clay indicate a much larger range of stable load combinations compared with piles in dense sand (Benzaria et al, 2013b).

Information on deformation can be included on the stability diagram as shown by Rimoy et al (2013) for field tests under cyclic axial tension on driven piles in sand, Figure 18. The data on the accumulation of displacement with cycling shows mostly stable behavior that suddenly degrades near the limiting number of cycles, a rather “brittle” behavior under tensile load.

A consistent observation for driven piles in dense sand is that stable cycling increases the ultimate capacity when tested subsequently. This is attributed to densification with some relaxation in lateral effective stress around the pile, as measured in the exceptional data of Silva et al. (2013). Conversely, strong cyclic loading (as would lead to instability) acts to reduce the ultimate axial pile capacity that can be mobilized subsequently.

Data on large axial pile tests in silt on a 4.2m long test pile are also interpreted in a stability diagram by Chen et al (2013).

6.2 Cyclic lateral loading

Rosquoët et al (2013) report data on lateral displacement for model driven piles in sand tested under one-way loading at 40g in a centrifuge. As in §2.3.3, but for displacement rather than rotation, the lateral displacement yN compared with the first load displacement y1 [ΔyN = yN – y1] is linked with the number of cycles. A logarithmic form ΔyN/ y1 = c.ln(N) fits the data well where c varies with the amplitude of cyclic load and the maximum lateral load (equivalent to 2Qcyclic/ Qmax in the axial terminology above, written as DF/F by Rosquoët et al). Based on their test data the authors suggest c = 0.1(DF/F)0.35 as a general fit, but sand and pile properties might also be important.

(Tsuha et al (2012) calibration chamber tests) Figure 16: Stability diagram: tension tests on model driven piles in sand. (Puech et al, 2013) Figure 17: Stability diagram: field compression on bored piles in sand.

(Rimoy et al, 2013 field tension tests on driven piles in sand) Figure 18: Stability diagram: accumulated displacement. The natural logarithm form has the advantage of fitting the data through to low numbers of cycles (recall §2.3.3). Rosquoët et al (2013) also note that the maximum moment in the pile does not increase significantly with lateral cyclic loading. Finally, the work to extend p-y analysis for laterally loaded piles failed to capture the measured behavior beyond the first few cycles.

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6.3 Extension of t-z analysis to cyclic loading

Burlon et al (2013) extend static t-z analysis for piles in sand to the case of tensile and compressive cyclic load and compare the results with centrifuge test data. The analysis turns out not yet to be practical as four new parameters are introduced that require measurement in cyclic pile tests. Even then, the fit with the test data is good only for relatively few load cycles.

6.4 Plugging of open-ended displacement piles

Laboratory tests on sand to measure pile plugging, using PIV observations, are described by Lueking and Kempfert (2013). The fully plugged limit IFR=0 was not achieved in the tests. The results of 2D Plaxis analysis are reported to investigate the mechanism of plugging. Based on this study, the authors propose two largely empirical calculation methods for the analysis of end-bearing for open-ended, partially plugged piles.

6.5 Cyclic pressuremeter tests

Reiffsteck et al (2013) report new work on the application of Ménard and self-boring pressuremeter tests to measure the change in soil properties with cyclic loading and, potentially, liquefaction resistance. Data are reported for several sites where cyclic pile tests were completed (SOLCYP). The authors emphasise the importance of a high quality borehole and the need for at least 50 cycles of repeated load to characterize the shear modulus. Soil horizons susceptible to liquefaction could be identified by large volume expansion. The acquisition of pore water pressure data during the test would greatly improve the test control (data on drainage/rate effects) and interpretation.

6.6 Osterberg cell testing

The general reporter is not aware of recent cyclic pile tests using O-Cell technology (Osterberg, 1989). A two level O-Cell test arrangement, for example, permits end bearing to be eliminated from cyclic axial compression tests on piles.

7 CONCLUSIONS.

As demonstrated by papers to this session, the practical challenges of offshore geotechnics actively drive forward the development of soil mechanics and geotechnical engineering. This is partly due to more extreme loading and deformation than usually encountered onshore. The fruits of this research and development are of great value for the overall understanding and pactice of geotechnical engineering. That is why offshore geotechnics should remain part of this key ISSMGE conference.

8 ACKNOWLEDGEMENTS

The author would thank TC209 Chairman Philippe Jeanjean for the invitation to prepare this General Report, and colleagues J.C. Ballard, P. Peralta and V. Whenham for valuable support. 9 REFERENCES

Arroyo M., Abadias D., Alcoverro J. and Gens, A. 2013. Shallow foundations for offshore wind towers. Proc. 18th ICSMGE, Paris.

Arshi H.S., Stone K.J.L., Vaziri M., Newson T.A., El-Marassi, M., Taylor R.N. and Goodey R.J. 2013. Modelling of monopile-footing foundation system for offshore structures in cohesionless soils. Proc. 18th ICSMGE, Paris.

Ballard J.C. and Charue N. 2013. Influence of jack-up footprints on mudmat stability - How beneficial are 3D effects? Proc. 18th ICSMGE, Paris.

Been, K. and Jefferies, M.G. (1985). A state parameter for sands. Géotechnique, Vol. 35(2), pp. 99-112.

Bellezza I., Mazzieri F., Pasqualini E., D’Alberto D. and Caccavo C. 2013. Design and installation of buried large diameter HDPE pipelines in a coastal area. Proc. 18th ICSMGE, Paris.

Ben-Hassine J. and Griffiths D.V. 2013. Geotechnical Exploration for Wind Energy Projects. Proc. 18th ICSMGE, Paris.

Benzaria O., Puech A. and Le Kouby A. 2013a. Essais cycliques axiaux sur des pieux forés dans des sables denses. Proc. 18th ICSMGE.

Benzaria O., Puech A. and Le Kouby A. 2013b. Essais cycliques axiaux sur des pieux forés dans l’argile surconsolidée des Flandres. Proc.18th ICSMGE, Paris.

Bretelle S., Wallerand R. Fondations Superficielles Glissantes pour l’Offshore Profond – Méthodologie de Dimensionnement. Proc. 18th ICSMGE, Paris.

Burlon S., Thorel L. and Mroueh H. 2013. Proposition d’une loi t-z cyclique au moyen d’expérimentations en centrifugeuse. Proc. 18th ICSMGE, Paris.

Chen R.P., Ren Y., Zhu B. and Chen Y.M. 2013. Deformation behavior of single pile in silt under long-term cyclic axial loading. Proc. 18th ICSMGE, Paris.

Cheuk, Y.C. and White, J.D. (2008). Centrifuge modelling of pipe penetration due to dynamic lay effects. Proc. Int. Conf. on Offshore Mechanics and Arctic Engineering, Portugal. OMAE2008-57923.

Damgaard M., Andersen J.K.F., Ibsen L.B. and Andersen L.V. 2013. Time-Varying Dynamic Properties of Offshore Wind Turbines Evaluated by Modal Testing. Proc. 18th ICSMGE, Paris.

DNV-OS-J101 (2011). Design of Offshore Wind Turbine Structures. Det Norske Veritas (DNV) Offshore Standard, September 2011.

Dutta S., Hawlader B. and Phillips R. 2013. Numerical investigation of dynamic embedment of offhore pipelines. Proc. 18th ICSMGE.

Einav, I. and Randolph, F.M. (2005). Combining upper bound and strain path methods for evaluating penetration resistance. Int. J. Numer.Meth. Engng., Vol. 63, pp. 1991-2016.

Ho J., Goh S.H. and Lee F.H. 2013. Post Cyclic Behaviour of Singapore Marine Clay. Proc.18th ICSMGE, Paris.

Kim D.J., Youn J.U., Yee S.H., Choi J., Choo Y.W., Kim S., Kim J.H., Kim D.S. and Lee J.S. 2013. Centrifuge test and numerical modelling for a suction bucket monopod foundation. Proc. 18th ICSMGE, Paris.

Kim J.M. and Safdar M. 2013. Behaviour of marine silty sand subjected to long term cyclic loading. Proc. 18th ICSMGE, Paris.

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Osterberg, J.O. 1989. New Device for Load Testing Driven Piles and Drilled Shafts Separates Friction and End Bearing. Proc. Int. Conf. on Piling and Deep Foundations, London, A.A. Balkema, p. 421.

Peire, K., Nonneman, H. & Bosschem E. (2009) Gravity Base Foundations for the Thornton Bank Offshore Wind Farm. Terra et Aqua, No. 115, pp. 19–29

Puech A., Benzaria O., Thorel L., Garnier J., Foray P., Silva M. and Jardine R. 2013. Diagrammes de stabilité cyclique de pieux dans les sables. Proceedings 18th ICSMGE, Paris.

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Rimoy S., Jardine R. and Standing J. 2013. Displacement response to axial cyclic loading of driven piles in sand. Proc. 18th ICSMGE.

Roesen H.R., Ibsen L.B. and Andersen L.V. 2013. Experimental Testing of Monopiles in Sand Subjected to One-Way Long-Term Cyclic Lateral Loading. Proc. 18th ICSMGE, Paris.

Rosquoët F., Thorel L., Garnier J. and Chenaf N. 2013. Pieu sous charge latérale : Développement de lois de dégradation pour prendre en compte l’effet des cycles. Proc. 18th ICSMGE, Paris.

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Tyldesley M., Newson T., Boone S. and Carriveau R. 2013. Characterization of the geotechnical properties of a carbonate clayey silt till for a shallow wind turbine foundation. Proc. 18th ICSMGE.

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Shallow foundations for offshore wind towers

Fondations superficielles pour des installations éoliennes maritimes

Arroyo M., Abadías D., Alcoverrro J., Gens A. Dep. of Geotechnical Engineering and Geosciences, Technical University of Catalonia, Barcelona, Spain

ABSTRACT: Direct foundations are present in about 25% of the installed offshore wind power towers. The peculiarities of this typeof structure are well known: high dynamic sensitivity, complex couplings between environmental actions, machine operation and structural response, complex installation and maintenance, difficult site investigation. There is a clear need for optimized foundationdesign tools that would enable cost reduction and a more detailed assessment of the risk of every installation. One such tool is likely to be the systematic use of failure envelopes for capacity checks. The paper explores the benefits of such an approach with various realistic design examples.

RÉSUMÉ : Les fondations superficielles interviennent dans la réalisation de 25% des structures éoliennes maritimes. Les particularités de ce type de structures sont bien connues: haute sensibilité dynamique, couplages complexes entre les actions environnementales, le fonctionnement de la machine et la réponse structurelle, installation et maintenance difficiles, investigation géotechniques onéreuses. Un besoin évident d'optimisation des outils de conception est nécessaire pour permettre la réduction des coûts et une évaluation plus détaillée du risque de chaque installation. Le recours systématique à des enveloppes de rupture pour les justifications de la capacité portante des fondations peut bien être un tel outil. Ce papier explore les avantages d'une telle approche avec divers exemples de conception réalistes.

KEYWORDS: direct foundation, capacity, offshore, energy, wind farms

1 INTRODUCTION

Offshore wind is an increasingly large contributor to the energy production mix of several European countries, particularly those bordering the North and Baltic seas. An exponential increase in installations is currently anticipated in this region. It is reasonable to expect that other regions of the world will follow suit.

Offshore wind turbines (OWT) are generally larger than those installed on land, with 3 to 5 MW of nominal capacity being now the norm, but with turbines of up to 10 Mw coming soon to the market. Rotor diameters of more than 100 m and nacelle locations 80 m above mean sea level are common. The result is a relatively lightweight and slender structure, supporting a rotating machine finely tuned to maximize power production while minimizing structural loading.

While initial OWT installations took place near shore (< 10 km) at locations with relatively shallow water depths (< 20 m), current developments are clearly located offshore (10 -100 km from the nearer coast) with water depths of 20-50 m being typical. Several floating support concepts are now being developed; however, commercial installations are still always supported by some kind of fixed structure. For these, the foundation of choice would depend in any case on the particular site conditions, construction equipment availability and, to a certain extent, local traditions.

To this date pile foundations have been largely dominant, mostly as single large (4-6 m diameter) monopile installations, and lately also as smaller (1-2 m) piles for jackets and tripods. However, examining the industry databases (e.g. Burton et al 2011) it appears that at the end of 2011 about 25% of the installed power was supported by direct foundations or gravity base substructures (GBS). Most of these GBS installations took place in relatively shallow waters, but there are some examples already at larger distances from the coastline and in deeper waters. Perhaps the most significant is the Thornton Bank I

project, 27 km offshore Zeebrugge in Belgium, where 6 OWT of 5 Mw were installed in water depths of 20-30 m. The foundation design for this installation was described by Peire et al (2009) and its outline is reproduced here in Figure 1. These are large (44 m height; 23.5 m base diameter) concrete shells, floated into place and later ballasted with a mixture of sand and olivine with the base at 4 m below the original seafloor level. The geotechnical profile at the site comprises medium and high density sands and stiff tertiary clays.

2 DESIGN ISSUES FOR DIRECT OWT FOUNDATIONS

There are several specific standards dealing with OWT. Perhaps the highest ranked is IEC 61400-3 (2009) which, from the point of view of structural design, establishes design cases and site ambient load specification procedures, introduces a safety format and gives broad indications about structural design procedures. However, detailed specification of structural and foundation design procedures is deliberately referred to other documents, like the ISO 1990X offshore standard series or DNV-OS-J101 (2010).

As might be expected, the indications given by such standards are, on the one hand, firmly based in conventional design practice when being specific, and somewhat elusive with problems that lack a clear conventional solution. An example of the later is the consideration of fatigue or foundation failure under cyclic loading. An example of the former is the consideration of foundation bearing capacity which, for shallow foundations, follows a conventional superposition and correction procedure not very different from those outlined by Brinch-Hansen (1970) or Vesic (1975).

When designing foundations for OWT, there will be of course issues of geotechnical capacity under extreme loads. However the design drivers might be sometimes related to other considerations, such as dynamic characteristics of the whole

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structure (Van der Temple and Molenaar, 2002) or displacement limits imposed by operating constraints (e.g. foundation tilting limits of 0.25° - 0.5° are sometimes quoted). However, even if we narrow our focus to bearing capacity considerations there are reasonable grounds to question the suitability of the conventional design approach.

Figure 1. Thornton Bank GBS (Peire et al, 2009).

Indeed there are several aspects of the traditional approach to bearing capacity that are poorly suited to deal with OWT. Firstly, using separate corrections for shape, depth, load inclination, load eccentricity is cumbersome and prone to calibration error if the effects that are being corrected for are not truly independent. This is perhaps the reason behind the large scatter between inclination factor formulations (Siefert & Bay-Gress; 2000); that uncertainty is particularly undesirable for structures, like OWT, that are mostly designed to sustain horizontal loads.

Secondly, the traditional approach to bearing capacity quickly leads to conundrums when the security format (as is the case for most modern codes, like DNV-OS-J101) is based on separate partial factors for loads and resistances. As discussed in detail by Lesny (2007) the same action might have a detrimental or favourable effect depending on which other actions are being simultaneously considered. Also it is fairly evident that a traditional bearing capacity check is far from eliminating the most likely path towards failure.

Finally, it is very difficult to generalize the traditional approach to cases when two major horizontal loads (wind, wave) are acting in separate planes. All these problems are best dealt with if the traditional approach to capacity checks is replaced by a failure-envelope based one.

3 FAILURE ENVELOPES

3.1 Concept

Failure envelopes were introduced (Butterfield & Ticof, 1979) as an alternative to classical bearing capacity analyses. They were based on the concept of interaction diagram, which was applied to the system of loads acting on the foundation. Most developments to date –but not all-, refer to the case in which that system can be reduced to loads acting within a plane (V, H, M) –where M represents the moment acting within the plane, M normalised by a characteristic foundation dimension, M/B.

Failure envelopes are implicit in the traditional approach to bearing capacity. However, it was clearly appreciated from the beginning that an explicit failure envelope was useful to link previously separate checks on different foundation failure modes (e.g. sliding and bearing capacity) into a coherent view. Failure envelopes offered advantages also from the experimental viewpoint, because they provide a clearer framework for experimentation, even suggesting new, more efficient, procedures (like “swipe” tests).

Failure envelopes are also attractive because they can fit well with generalized force-displacement foundation models (“macroelements”; Nova and Montrasio, 1991) that are used to compute foundation displacements and represent an economical solution to non-linear soil-structure interaction studies. Finally, failure envelopes are interesting because they enable a more coherent approach to foundation safety.

3.2 Safety considerations

Already Georgiadis (1985) clearly identified as one major advantage of failure envelopes that they allow a very natural consideration of the influence of different loading paths. To do that, it is important to distinguish between the reference design load state and incremental loading paths (Figure 2).

M

H

(H, M)

(H, M)(Hr, Mr)

Figure 2. Schematic load envelope illustrating a reference design load and one incremental load path

Any load system (V, H, M) shall remain within the failure envelope. It is however convenient to establish a non-dimensional safety measure. To do so a simple approach is, for any incremental loading direction, to obtain the crossing point with the failure envelope (Vr, Hr, Mr) and then define a generalized safety factor, SF, as

V,H, V,H, 1 rMSF M (1)

( , , )r r r r r rV V V H H H M M M (2) It is thus made explicit the fact that safety is not only

dependent on the initial design situation but also on the incremental loading path. This definition includes, as a particular case, the traditional safety factors against bearing capacity (the incremental load direction and the reference

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design load are collinear) or sliding (incremental loading direction collinear with the Horizontal component of the reference design load). Another particular case included is that of “plastic overturn”, a prescribed check for breakwater design in Spanish regulations (Puertos del Estado, 2005) in which the lever arm of the horizontal loading is maintained (i.e. the incremental load is aligned with the the Horizontal and Moment components of the reference load).

3.3 Example formulations

Figure 3 Failure envelope by Gottardi et al (1999) There are many failure envelopes in the literature. For foundations failing without drainage at the soil-foundation interface Gourvenec & Randolph (2011) offer an excellent review. For the example below a sand profile is assumed and drained conditions are reasonable. In these circumstances a convenient expression for a failure envelope is that proposed by Gottardi et al. (1999) (Figure 3)

2 2

2

0 0 0 0

( , , )

2 (4 (1 )) 0

F V H M

h m hma v v

h m h m

(3)

Where (a, h0, m0) are shape factors, empirically determined

as (-0.22, 0.12, 0.09) for quartzitic sand, and we use a non-dimensional notation in which v = V/V0, h = H/V0, m = M/(DV0) and D is the foundation diameter. The normalizing factor V0 is the maximum load (i.e. centered vertical) that the foundation can sustain. Here that maximum load is computed assuming no embedment and introducing the bearing capacity factor N from Bolton & Lau (1993) into

2

0

1

2 4

DV DN

(4)

It is worth noting that (a) it is relatively straightforward to

generalize expression (3) to more complex loading situations –e.g. Lesny 2010- although the experimental base for adjusting the parameters in those circumstances is somewhat scarce, (b) that the shape of (3) above has proven rather resilient and very similar expressions have been found to fit well other foundation test results in materials like carbonate sand or even clay (Martin & Houlsby, 2001), as long as the contact surface remains drained. Of course the choice of V0 would change according to the material and foundation shape.

4 EXAMPLE APPLICATION

To illustrate the argument we propose an example, synthetic but realistic. The case is developed using the characteristics of the gravity base substructure built at Thornton Bank (Peire et al. 2009) and the design loading specified for a Baltic windfarm development site, Kriegers Flak (Bulow et al, 2009). This reference gives some basic characteristics for the OWT superstructure (Table 1).

Table 1 Super-structure characteristics

Rated power 5 MW

Rotor diameter 126 m

Nacelle height above msl 90 m

Nacelle-rotor weight 4.1 MN

Tower weight 3 MN

The same reference also includes resultants from ambient loads for a range of depths and load hypothesis (e.g. extreme, fatigue). Using these data, Table 2 has been computed for a 30 m depth case and extreme load scenario. It appears that, in this particular case, 80% of the total horizontal thrust is due to sea action, but this load is the source of less than 20% of the overturning moment at foundation level. This might partly reflect the fact that at that particular site sea current is relatively strong, lowering the action line of sea forces.

These ambient loads should be combined with the OWT selfweight. Using the Thornton Bank design like a template for substructure shape, the relevant characteristics of that part of the OWT are those listed in Table 3. As usual with gravity base OWT, the dead weight of the substructure is significantly larger than that of the superstructure. Combining all environmental actions and structure selfweight the resultant load combination acting at the foundation level is (H, V, M) = (10.1; 44.5; 284.3) in MN and MNm. This will be the reference design load state in this example.

Tabla 2 Ambient loading parameters

Parameter / load Unit Value

Total thrust, H MN 10.1

Total overturning moment, M MNm 284.3

Wind thrust, Hw MN 2.03

Wind arm lever, bw m 120

Sea thrust, Hs MN 8.07

Sea arm lever, bs m 5

Tabla 3 Thornton Bank type substructure characteristics

Parameter / load Unit Value

Base diameter m 23,5

Concrete weight MN 30

Fill weight MN 38

Buoyant volume m3 2965

From that reference state we probe the failure surface

alongside three different incremental loading paths. One will correspond to a simultaneous and proportional increase of all ambient actions (the “plastic overturn” case). The other two hypothesis would correspond to increases of just one of the

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ambient horizontal actions, (sea, wind) while the other remains constant. These hypotheses would, for instance, naturally follow from any circumstance in which the estimates of wind and wave carry different uncertainties. Figure 4 illustrates graphically the meaning of these load directions in an idealised section of the failure envelope at constant V.

For this check we use the failure envelope of Gottardi et al described above. The soil profile below the foundation is characterised by a friction angle of 33° and submerged weight of 10 kN/m3. These values might correspond well to the characteristic values of a medium-dense sand profile, frequently encountered in North Sea locations. It is assumed that the foundation base is perfectly rough.

Figure 4 Incremental load paths in the example Table 4 Example: results

Hypothesis Hr (MN) Hr / Hi ΔH (%)

Sea 14.1 1.4 50

Both 11.6 1.15 15

Wind 10.5 1.04 21

Some relevant results from the computation are presented in table 4. For each incremental loading path a failure point is identified in the envelope, with values (Hr, Mr). In the table the value Hr corresponding to each loading path is reported in the first column. In the second column this value is normalized by the reference state horizontal load. This corresponds to the generalized safety factor defined above, which, only for the hypothesis in which both loads are simultaneously increasing, coincides with the “plastic overturn” safety factor of ROM 0.5-05. As a reference the value required for that safety factor in breakwaters is commonly above 1.3 (Puertos del Estado; 2005).

For the other two load hypothesis in which only one environmental action is increased no similar reference exists to judge on the computed safety factor. For these cases it is perhaps more meaningful the number in the third column of Table 4, where the difference between failure and reference thrust is expressed as a percent of the reference ambient load that is increasing. In the case computed, a 21% error in the reference estimate of wind thrust would result in foundation failure, whereas it would be necessary a 50% underestimate of the hydrodynamic thrust to fail the foundation.

The previous computations have always been made under the hypothesis of increased thrust and constant lever arm. This can be interpreted as action magnitude uncertainty. Alternative hypothesis dealing with lever arm uncertainty can be equally set up with relative ease. Note, finally, that most geotechnical uncertainty can be lumped in the V0 estimate to achieve a relatively straightforward approach to reliability evaluation.

5 CONCLUSION

Failure envelopes offer a powerful framework to analyze shallow foundation capacity problems. They seem particularly suitable for offshore wind towers, where refined design in the face of large load uncertainties is likely to be a frequent situation.

6 ACKNOWLEDGEMENTS

The research on direct foundations for offshore wind towers described in this paper was partly founded by the company ACCIONA ENERGY within the framework of the CENIT-AZIMUT project supported by the Spanish Ministry of Science.

7 REFERENCES (TNR 8)

Brinch-Hansen J. (1970). A revised and extended formula for bearing capacity. Danish Geotechnical Institute Bulletin, n° 28, 5-11

Bülow, L., Jorgensen, L. and Gravessen, H. (2009) Kriegers Flak Offshore Wind Farm. Basic data for conceptual design of foundations. March 2009, Vattenfall Vindkraft AB

Burton, T., Sharpe, D., Jenkins, N. and Bossanyi, E. (2011) Wind Energy Handbook, 2nd Edition, John Wiley & Sons, Chichester, UK

Butterfield R., Ticof J. (1979). The use of physical models in design (discussion). 7th European Conference on Soil Mechanics and Foundation Engineering, Brighton, UK, Vol.4, 259-2

Georgiadis, M (1985) Load-path dependent stability of shallow footings, Soils & Foundations, 25,1, 84-88

Gottardi, S., Houlsby, G.T. y Butterfield, R. (1999) Plastic response of circular footings on sand under general planar loading, Géotechnique 49, No. 4, 453-469

Gourvenec, S. y Randolph, M. (2011) Offshore geotechnical engineering, Spon Press, New York

DNV (2010) DNV-OS-J101 Design of offshore wind turbine structures IEC 61400-3 (2009) International standard Wind turbines –Part 3:

Design requirements for offshore wind turbines. International Electrotechnical Comission

Lesny, K. (2010) Foundations for offshore wind turbines : tools for planning and design, VGE, Essen

Lesny, K. (2007) Design approaches of Eurocode 7 and their effect on the safety of shallow foundations, ICASP10, Applications of statistics and probability in Civil Engineering, Taylor & Francis

Martin, C., M., Houlsby, G. T. (2001) - Combined loading of spudcan foundations on clay: numerical modeling. Géotechnique, 51, No. 8, pp. 687 – 699

Nova R. y Montrasio L. (1991). Settlements of shallow foundations on sand. Géotechnique, vol.41(2), 243-256.

Peire, K., Nonneman, H. & Bosschem E. (2009) Gravity Base Foundations for the Thornton Bank Offshore Wind Farm. Terra et Aqua, N. 115, pp. 19–29

Puertos del Estado (2005) ROM 0.5-05 Recomendaciones Geotécnicas para Obras Marítimas y Portuarias, http://www.puertos.es/es/programa_rom/rom_05_05.html

Sieffert, J.G., y Bay-Gress, Ch. (2000). Comparison of the European bearing capacity calculation methods for shallow foundations; Geotechnical Engineering, Institution of Civil Engineers, Vol. 143, pp. 65-74

Van der Tempel, J. y Molenaar, D.P. (2002) Wind turbine structural dynamics – a review of the principles for modern power generation, onshore and offshore, Wind Engineering, 26,4, 211-220

Vesic, A. S. (1975) Bearing capacity of shallow foundations, Ch. 3 in Winterkorn H.F. & Fang H.Y., Foundation Engineering Handbook, Van Nostrand Reinhold

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Modelling of monopile-footing foundation system for offshore structures in cohesionless soils

Modélisation d’un système de fondation superficielle isolé pour sur les structures maritimes dans les sols pulvérulents

Arshi H.S., Stone K.J.L. University of Brighton, UK

Vaziri M. Ramboll UK Limited, UK

Newson T.A., El-Marassi M. University of Western Ontario, Canada

Taylor R.N., Goodey R.J. City University London, UK

ABSTRACT: While monopiles have proven to be an economically sound foundation solution for wind turbines, especially inrelatively shallow water, their installation in deeper water and in hard ground may require a more complex foundation design in orderto satisfy the loading conditions. One approach is that foundation systems are developed which combine several foundation elements to create a ‘hybrid’ system. In this way it is possible to develop a foundation system which is more efficient for the combination ofvertical and lateral loads associated with wind turbines while maintaining the efficiency and simplicity of the design. Previous studies have reported the results of single gravity tests of the hybrid system where the benefits of adding the footing to the pile are illustrated.This paper presents experimental results on the performance of skirted and unskirted monopile-footings. A simplified design approachbased on conventional lateral pile analysis is presented.

RÉSUMÉ : Alors que les fondations de type monopile se sont révélées être une solution économiquement viable pour les fondationsd’éoliennes, en particulier dans les eaux relativement peu profondes, leur installation dans des eaux plus profondes et dans un sol durpeut exiger une conception plus complexe afin de satisfaire les conditions de chargement. Une approche possible est que les systèmes de fondations développés combinent plusieurs éléments de fondation pour créer un système hybride. De cette manière, il est possiblede développer un système de fondation plus efficace vis à vis des charges verticales et latérales associées aux éoliennes, tout enmaintenant une conception efficace et simple. Des études antérieures sous gravité simple ont montré l’efficacité d’un système hybride en combinant une semelle et un pieu . Cet article présente des résultats expérimentaux sur la performance de systèmes avec et sans pieu pour des semelles. Une approche de conception simplifiée basée sur l'analyse classique d’un pieu sous charge latérale est présentée.

KEYWORDS: Hybrid monopile footing, offshore piles, laterally loaded piles, wind turbine foundations

1 INTRODUCTION

Due to the needs of on-going developments in the oil and energy sector, the design of offshore foundations is constantly evolving. In the hydrocarbon extraction sector, exploration and development is moving in to ever deeper water resulting in ever more challenging geotechnical conditions. Similarly the expansion of the offshore wind sector involves the development of deepwater sites, together with requirements for heavier high capacity turbines. Conventional offshore foundations are not always economical or practical for this new generation of turbines, and there remains a requirement to develop foundation solutions which can better satisfy future developments in the offshore wind sector.

The foundations of a typical offshore wind turbine are subjected to combined loading conditions consisting of the self-weight of the structure (V), relatively high horizontal loads (H) and large bending moments (M). The preferred foundation system to date has been the monopile, which has the advantage that it can be employed in a variety of different soil conditions. However, a disadvantage in the use of monopiles in deep water sites is that the system can be overly compliant. For sites with intermediate water depths, it may be possible to stiffen the lateral response of the monopile at the mudline.

Figure 1. Schematic illustration of the prototype hybrid system.

One such approach to increase the lateral resistance of a monopile is the ‘hybrid’ monopile-footing system. As schematically represented in Figure 1, this foundation system

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comprises of a circular footing attached to the monopile at the mudline. A 2-D analogy of this system is that of a retaining wall with a stabilising base (Powrie and Daly, 2007). The role of the footing is to provide a degree of rotational restraint at the pile head, leading to an improvement in the lateral resistance of the pile. It has also been shown that the use of a relatively thick pile cap leads to an increase in the lateral resistance through the development of passive soil wedges (Mokwa, 1999), in a similar way to the behaviour of skirted foundations (Bransby and Randolph, 1998).

Analysis of the hybrid system would involve both lateral pile analysis and bearing capacity analysis. The lateral response of piles is well reported in the literature and various methods of analysis have been proposed by numerous researchers, such as Matlock and Reese (1960), Broms (1954), Poulos (1971), Reese et al. (1974), Randolph (1981), Duncan et al (1994) and Zhang et al. (2005). Where the plate diameter is relatively small, the system is similar to a single capped pile, for which methods have been developed for analysing the influence of the pile and pile cap under axial loading (Poulos and Randolph, 1983), and the effect of the pile cap on the lateral performance of single piles has also been investigated by others (Kim et al., 1979), (Mokwa and Duncan, 2001: 2003), (Maharaj, 2003).

The bearing capacity problem has also been investigated under different loading conditions relevant to offshore foundations, see for example references Houlsby and Puzrin (1999), and Gourvenec and Randolph (2003).

2. EXPERIMENTAL INVESTIGATIONS

The potential performance of the hybrid system was investigated in single gravity studies (Stone et al. (2007)) and is illustrated in Figure 2. These studies suggested that the additional rotation restraint provided by the footing can result in a stiffer lateral response of the pile and greater ultimate lateral load. The degree of restraint at the pile head was dependent on the size of the footing, the initial contact between the soil and the footing and the stiffness of the soil beneath the footing. Observations of heaved and displaced soil in front of the edge of the footing also suggested that a degree of passive soil resistance is likely to be generated under the lateral movement and rotation of the footing.

Figure 2. Lateral load response of the hybrid system (after Stone et al.2007).

Arshi (2011), and Arshi and Stone (2012) reported the results of a comprehensive series of single gravity testes carried out on the foundation system where the elements affecting the overall performance of the foundation system was investigated in depth. It was reported that the size for the footing has a direct effect on the overall lateral load bearing capacity of the foundation system. Furthermore it was reported that the ratio between the vertical and horizontal load has a significant effect on the lateral performance of the foundation system where larger vertical loads tend to improve the lateral load bearing capacity of the hybrid system. The connection between the

footing and the pile was also investigated where it was suggested that the hybrid foundation system tends to be more effective if vertical movements are allowed at the pile-footing connection. This movement allows the footing to act independently from the pile where the positive contact between the footing and the soil underneath is solely controlled by the vertical load acting on the footing.

Table 1. Notations for skirted hybrid foundations system

ID Footingsize (mm)

Skirtlength (mm)

Deadload(N)

Footing to pile connection

P.W0 - - 0 -P.F80.W1.FR 80 100 SlippingP.F80.S1.W1.FR 80 100 SlippingP.F80.S2.W1.FR 80 100 SlippingP.F80.S3.W1.FR 80 100 Slipping

More recent single gravity tests are presented in Figure 3 where skirts with different lengths have been added to the footing. The tests were conducted in sand and the results indicate that the presence of the skirts has a relatively significant contribution on the lateral load capacity of the system. The results show that adding the skirts to the footing and increasing the skirt length tends to increase the lateral load bearing capacity of the foundation system by about 50% in comparison to a non-skirted hybrid system. It is also apparent that footings with very short skirts do not tend to show any ‘apparent’ additional advantage to that without the skirt. This could be due to the fact that the stresses around the skirt induced by the soil are very small at 1g. Further studies in the centrifuge are in the taking place to investigate the effect of the skirts and the results will be reported soon.

Figure 3. Load vs. deflection plot for the hybrid system with skirts.

Stone et al (2011) reported the results of a series of centrifuge tests in sand. The results of the combined vertical and lateral loading tests are best represented through plots of lateral load versus lateral displacement. Figure 4 shows a plot of the lateral load versus lateral displacement for the monopile-footing (HL 1) and single pile (PL 1) with a vertical load of 600N at 50 g. It is apparent from this plot that the initial lateral stiffness of the monopile-footing and pile are similar for the first 1–1.5mm of lateral displacement. However the monopile-footing continues to exhibit a stiffer response than the single pile as the lateral displacement increases. Further analyses of these data provided information on the redistribution of bending moment in the pile due to the plate.

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Figure 4. Load deflection graph for centrifuge tests carried out on the hybrid system (after Stone et al. 2011).

In Figure 5, the bold lines represent the bending moments at 5% and 20% of the maximum deflection for the pile only case and the dashed lines show the behaviour of the hybrid system. The results show that adding the footing to the pile reduces the bending moment at any given deflection, and as a result increases the moment capacity of the system at any given applied lateral load. The results indicate about 25% improvement in the bending moment for at both deflections.

Figure 5. Bending moment distribution along the pile length for the hybrid system.

3. ANALYSIS

Whilst some advanced numerical modelling of monopiled footings has been undertaken (El-Marassi et al. 2008; Stone et al. 2010; Arshi et al. 2011; Arshi and Stone 2012), the method presented here utilises conventional lateral pile analysis methodology where the hybrid system is idealised to a lateral pile with a resisting moment applied at the mudline. The resisting moment capacity provided by the footings were estimated analytically using conventional bearing capacity theory and applied at the mudline acting in the oposite dirtection to the loading. This approach only considers the ultimate condition of the system and does not allow the moment

developed by the footing to be generated as a function of the footing rotation.

The results generated by this approach are illsutetared in Figure 6 where it is shown how different pile to footing diamater increses the moment capacityof the piles, where this variation lies between a fully free and a fully fixed pile.

The dashed lines in Figure 6 show the ultimate moment capacities of the hybrid system. Although this method successfully leads to obtaining the ultimate load bearing capacity of the hybdegree of rigidity (D.O.R 75%, 50% and 25% showing the the ultimate capacity of the system when %75, 50% and 25% of the ultimate moment at pile head is applied to the free headed pile) of the system are shown as a benchmark for comparing how differenet pile to footing diameters relate to the fully fixed moment. As apparent in Figure 6, increasng the sie of the footing tends to increse the lateral load bearing capacity. As the footing size increases, it gets close to the fully fixed head condrion. This also indicates that there for a given pile diamater and length, there ought to be a footing size afterwhcih increseing the footing size further will not enhance the lateral load bearing capacity of the foundation system.

Figure 6. Moment vs. rotation plot for the hybrid system with different pile to footing ratios.

In addition to this, design charts have been developed which relate the pile embedment length to pile and footing diameters. Numerous design charts have been developed covering a wide range of pile diameters, pile lengths, footing diameters and normalized moment capacities an example of which is shown in Figure 8 where the L/D ratios vary from 1 to 10 and the footing to pile diameter ratios varies from 0 to 1. The moment capacity of the hybrid system has been normalised and is shown against footing to pile diameter ratio. The lines in between represent different pile embedment depth where for a given moment capacity the designer could utilise this graph to choose the appropriate pile length as well as pile and footing diameters. It is also notable that for any given value of normalized moment capacity the designer has the option of choosing a short pile relatively large footing diameter, or long pile with relatively small footing diameter. The flexibility in this design approach is beneficial in particular designing the hybrid system in difficult soil conditions.

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Figure 7. Example of a design chart for the hybrid system developed using analytical and numerical methods.

4. DISCUSSION & CONCLUSION

It is apparent that the ultimate lateral response of a single monopile foundation can be enhanced by the presence of a footing resulting in a greater ultimate lateral capacity. This improvement was observed at both load versus deflection as well as the bending moment versus depth plots. Whilst the effect on the initial lateral stiffness may not be significant, the lateral stiffness beyond this initial movement was significantly enhanced through the presence of the footing.

The effect of adding skirts to the hybrid system has been shown to further increase the lateral performance of the hybrid system, and centrifuge tests are planned to investigate the skirted system in more detail.

A simple analytical approach using conventional lateral pile analysis methods is presented from which preliminary design charts can be generated. This approach can be developed to generate realistic design charts where the lateral capacity of the hybrid system is related to the development of bearing capacity coupled to the lateral resistance of the pile shaft.

5. REFERENCES

Arshi HS. (2011). Structural behavior and performance of skirted hybrid monopile-footing foundations for offshore oil and gas facilities. Proceedings of the Institution of Structural Engineers: Young Researchers Conference ‘11. London: IStructE Publications, 8.

Arshi HS, Stone KJL and Newson TA. (2011). Numerical modelling on the degree of rigidity at pile head for offshore monopile-footing foundation systems. 9th British Geotechnical Association Annual Conference, London.

Arshi HS and Stone KJL. (2011). An investigation of a rock socketed pile with an integral bearing plate founded over weak rock. Proceedings of the 15th European Conference of Soil Mechanics and Geotechnical Engineering. Amsterdam: Ios Pr Inc, 705 – 711.

Arshi HS. (2012). A new design solution for increasing the lateral resistance of offshore pile foundations for wind turbines located in deep-water. Proceedings of the Institution of Structural Engineers: Young Researchers Conference ‘12. London: IStructE Publications, 10.

Arshi HS and Stone KJL. (2012). Lateral resistance of hybrid monopile-footing foundations in cohesionless soils for offshore wind turbines. Proceedings of the 7th International Conference on

Offshore Site Investigation and Geotechnics. London: Society for Underwater Technology, 519 – 526.

Bransby MF and Randolph MF. (1998). Combined loading of skirted foundations. Géotechnique. 48(5), 637–655.

Broms BB. (1964). Lateral resistance of piles in cohesionless soils. ASCE Journal of the Soil Mechanics and Foundation Division.90(SM3), 123-156.

Duncan JM, Evans LT and Ooi PS. (1994). Lateral load analysis of single piles and drilled shafts. ASCE Journal of Geotechnical Engineering. 120(6), 1018-1033.

El-Marassi M, Newson T, El-Naggar H and Stone KJL. (2008). Numerical modelling of the performance of a hybrid monopiled-footing foundation. Proceedings of the 61st Canadian Geotechnical Conference, GeoEdmonton 2008. Edmonton, (Paper No. 480), 97 – 104.

Gourvenec S and Randolph M. (2003). Effect of strength non-homogeneity on the shape of failure envelopes for combined loading of strip and circular foundations on clay. Géotechnique.53(6), 575–586.

Houlsby GT and Puzrin AM. (1999). The bearing capacity of a strip footing on clay under combined loading. Proc. R. Soc. London Ser. A. 455, 893–916.

Kim JB, Singh LP and Brungraber RJ. (1979). Pile cap soil interaction from full scale lateral load tests. ASCE Journal of Geotechnical Engineering. 105(5), 643-653.

Maharaj DK. (2003). Load-deflection response of laterally loaded single pile by nonlinear finite element analysis. EJEG.

Matlock H and Reese LC. (1960). Generalized solutions for laterally loaded piles. ASCE Journal of Soil Mechanics and Foundations Division. 86(SM5), 63-91.

Mokwa RL. (1999). Investigation of the Resistance of Pile Caps to Lateral Loading. Ph.D Thesis. Virginia Polytechnic Institute, Blacksburg, Virginia.

Mokwa RL and Duncan JM. (2001). Experimental evaluation of lateral-load resistance of pile caps. ASCE Journal of Geotechnical and Geoenvironmental Engineering. 127(2), 185 - 192.

Mokwa RL and Duncan JM. (2003). Rotational restraint of pile caps during lateral loading. ASCE Journal of Geotechnical and Geoenvironmental Engineering. 129(9), 829 - 837.

Poulos HG. (1971). Behaviour of laterally loaded piles: Part I-single piles. ASCE Journal of the Soil Mechanics and Foundations Division. 97(SM5), 711-731.

Poulos HG and Randolph MF. (1983). Pile group analysis: a study of two methods. ASCE Journal of Geotechnical Engineering.109(3), 355-372.

Powrie W, and Daly MP. (2007). Centrifuge modelling of embedded retaining wall with stabilising bases. Geotechnique. 57(6), 485-497.

Randolph MF. (1981). The response of flexible piles to lateral loading. Géotechnique. 31(2), 247-259.

Reese LC, Cox WR and Koop FD. (1974). Analysis of laterally loaded piles in sand. Offshore Technology Conference. Vol. II (Paper No. 2080), 473-484.

Stone KJL, Newson TA and Sandon J. (2007). An investigation of the performance of a ‘hybrid’ monopole-footing foundation for offshore structures. Proceedings of 6th International on Offshore Site Investigation and Geotechnics. London: SUT, 391-396.

Stone KJL, Newson TA and El Marassi, M. (2010). An investigation of a monopiled-footing foundation. International Conference on Physical Modelling in Geotechnics, ICPMG2010. Rotterdam: Balkema, 829-833.

Stone KJL, Newson TA, El Marassi M, El Naggar H, Taylor RN, and Goodey RA (2011). An investigation of the use of bearing plate to enhance the bearing capacity of monopile foundations. International Conference on Frontiers in Offshore Geotechnics II - ISFOG. London: Taylor and Francis Group, 623-628.

Zhang L, Silva F and Grismala R. (2005) Ultimate lateral resistance to piles in cohesionless soils. Journal of Geotechnical and Geoenvironmental Engineering. Vol. 131(1), 78–83.

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Influence of jack-up footprints on mudmat stability – How beneficial are 3D effects?

Influence des dépressions laissées par les jack-ups sur la capacité portante des mudmats – quels sont les effets bénéfiques d’une analyse en 3D?

Ballard J.-C., Charue N. Fugro GeoConsulting Belgium

ABSTRACT: Jacket platforms are piled into the seabed but need to be supported temporarily by mudmats during installation. Theysometimes need to be located next to seabed features such as pug marks formed by previous deployments of jack-up rigs. These features may influence the bearing capacity of the mudmats. This is a 3D problem for which simplified approaches are unsatisfactory, simplified 2D plane strain simulations can lead to over-conservative results. This paper presents a project example in very soft clay for which the software package Plaxis 3D has been successfully used. The presence of a pug mark was found to degrade significantly the yield surface in the VHM load space. A comparison between 2D and 3D analyses shows that the beneficial 3D effects are substantial, especially when the pug mark is located at the corner of the mudmat. The zone of influence of the pug mark is also much more limited when the problem is modelled in 3D.

RÉSUMÉ : Les platesformes de type « Jacket » sont fondées sur pieux mais nécessitent d’être supportées temporairement pendantl’installation par des mudmats (fondations de type superficiel). Ces jackets sont parfois situées à proximité de dépressions laissées parl’installation antérieure de jack-ups. Ces dépressions peuvent influencer la capacité portante des mudmats. Il s’agit d’un problème 3Dtypique pour lequel aucune solution simplifiée n’existe. Une approche 2D (en état plan de déformation) peut même mener à desrésultats trop conservatifs. Cet article présente un exemple dans de l’argile molle pour lequel la suite de logiciels Plaxis a été utiliséeavec succès. Les conclusions sont les suivantes : la présence des dépressions modifie singulièrement la surface de rupture dansl’espace VHM. Une comparaison entre les approches 2D et 3D montre que les avantages à faire appel au 3D sont substantiels,spécialement quand la dépression est située à proximité du coin du mudmat. La zone d’influence de la dépression est aussi bien pluslimitée lorsque le problème est modélisé en 3D.

KEYWORDS: Pug mark, mudmat, stability, VHM, 2D, 3D, Finite Element Analysis, soft clay, remoulded, jack-up, mesh

1 INTRODUCTION

Jacket platforms are the most common type of offshore structure in the offshore hydrocarbons industry (Dean, 2010). They consist of open-framed steel structures made of tubular leg chords, horizontal bracing, and diagonal bracing. These structures are piled into the seabed but need to be supported temporarily by mudmats during installation. Mudmats are essentially flat stiffened metal plates attached to legs or the lower braces. In soft soils, mudmats can cover the entire surface between the legs to maximise the bearing area. They are generally subjected to combined Vertical, Horizontal and Moment (VHM) loads induced by the jacket weight, wind, waves and currents.

Jacket platforms are not always installed on a virgin seabed and are sometimes located next to features such as pug marks formed by previous deployments of jack-up rigs. A jack-up is a mobile, self-elevating offshore platform consisting of a hull and three or more retractable legs passing through the hull (McClelland et al, 1982). A unit moves onto location, sets its legs onto the seabed, and raises its hull out of the water. The legs are supported on independent foundations called spudcans. Penetration and extraction of spudcans in soft grounds create zones of remoulded soil and seabed depressions (Hossain et al, 2012). These seabed features potentially influence the bearing capacity of the mudmats and need to be accounted for in the stability verification.

Mudmats subjected to combined VHM loads and located next to a jack-up footprint is a 3D problem for which simplified approaches for analysis do not exist. Simplified 2D plane strain simulations are generally performed but they can lead to over-

conservative results. This type of problem is better analysed by means of 3D Finite Element (FE) analyses.

This paper presents a project example in very soft clay for which the software package Plaxis 3D (Plaxis, 2011) has been used successfully. The analysis allowed confidence to be established for the selected location of the mudmat with respect to a pug mark. In contrast, a simplified 2D analysis suggested that the proximity of the mudmat to the pug mark was unacceptable.

It is shown for this particular example how the presence of a pug mark degrades the yield surface in the VHM load space. 3D analyses are compared with 2D analyses to quantify the beneficial 3D effects for different pug mark locations.

2 PROBLEM GEOMETRY AND SOIL CONDITIONS

A 30 m by 30 m square mudmat is considered. The mudmat is located next to a circular pug mark of 30 m in diameter. Analyses were performed for 2 positions of the mudmat. The first position considers a pug mark located along the width of the mudmat while the second position considers a pug mark located at the corner of the mudmat, as illustrated on Figure 1. The distance d between the edges of the mudmat and the pug mark is varied in the analysis.

Soil conditions in this example consist of very soft clay. The soft deposit is considered to be 15 m thick and underlain by stiffer soils, which are not modelled. The intact undrained shear strength increases linearly with depth according to su = 4 + 0.8z (in kPa), where z is the depth below ground level in meter. This gives a strength heterogeneity = 6 where = kB/su0, k is the

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shear strength gradient, B the mudmat width and su0 the undrained shear strength at mudline.

A cylinder of soil with remoulded properties is considered to model the pug mark. This cylinder extends to the bottom of the soft layer. A 30 m diameter cylinder corresponds approximately to a 15 m diameter spudcan. The remoulded zone created by the penetration and extraction of a spudcan has indeed been observed to be of the order of 2 times the spudcan diameter (Hossain, 2012). On removal of the jack-up unit, the spudcans leave depressions at the site. The depth and configuration of the depression depends on several factors including soil strength, spudcan final penetration, amount of soil backfill during installation, etc. A seabed depression of 2m is considered in this paper. To keep the model geometry simple, a horizontal depression with vertical walls is modelled, as illustrated on Figure 1. A remoulded undrained shear strength profile sur = 2 + 0.4z (in kPa), where z is the depth in meter below original ground level, is considered within the pug mark area.

Position 1

Position 2

B = 30 m

B = 30 m

D =30 m

V

15 m Remoulded soil

2 m Intact soil

dM

H

Figure 1. Problem geometry: plan view and cross-section

3 FINITE ELEMENT MODEL

2D plane strain and 3D FE simulations were carried out using Plaxis (Plaxis, 2011). The 2D analyses only consider a pug mark along the width of the mudmat while the 3D analyses consider two positions for the pug mark: along the width and at the corner.

An example of 3D finite element mesh is shown on Figure 2. Similar mesh discretization was adopted for the different analyses. The external boundaries were set sufficiently remote so as not to intercept the different failure mechanisms.

Preliminary analyses were first performed for the base case without a pug mark and for which analytical and/or numerical solutions exist. The aim was to check for any effects due to mesh size on the accuracy of the solution. A compromise was found between the accuracy of the solution and computational time. It is estimated that the over-estimation of the true solution due to discretization errors was maximum 10% for the selected mesh, which was judged to be reasonable.

Figure 2. 3D Finite Element meshes

The soil is modelled as an isotropic elasto-perfectly plastic continuum, with failure described by the Mohr-Coulomb yield criterion. It is assumed to behave “undrained” and is characterized by a cohesion equal to the undrained shear strength su with u=0. The elastic behaviour was defined by a Poisson’s ratio =0.495, and a constant ratio of Young’s modulus to undrained shear strength E/su=300 for both undisturbed and remoulded clays.

The strength of the mudmat/clay interface is modelled using an interface factor , where the maximum shear stress at the interface max = su. The “rough” and “smooth extremes of interface strength correspond to = 1 and = 0 respectively. An intermediate roughness was assumed with = 0.5, which is a typical assumption for steel/soft clay interface. A no-tension condition allowing separation of the mudmat from the seabed was permitted at the mudmat/clay interface.

The jacket mudmat is modelled as a 30 m by 30 m rigid plain square plate. The seabed is assumed to be perfectly flat below the mudmat.

4 DESIGN PROCESS

The vertical load V from the mudmat and jacket structure is generally known and well-defined. It should typically be limited to a maximum of 50% the uniaxial vertical capacity. Then, for a given mobilisation ratio of the uniaxial vertical capacity v = V/Vult, the stability verification consists of ensuring that there is adequate factor of safety on the ‘live’ loading M and H. This can be performed by comparing design load combinations to the MH failure envelope. The higher the mobilisation of the uniaxial vertical capacity, the lower the moment and horizontal capacity, i.e. the MH failure envelope shrinks with increasing v. M and H loads are applied in the direction of the pug mark centre, namely perpendicular to the side of the mudmat or along its diagonal.

The presence of a pug mark with remoulded soil conditions in the vicinity of a mudmat has two adverse effects. First, it affects the moment and horizontal capacities for a given v. Second, it reduces Vult and therefore increases v, reducing further the moment and horizontal capacities.

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5 RESULTS AND DISCUSSION

5.1 2D Analyses

The accuracy of the 2D FE model was verified by computing the uniaxial vertical capacity and comparing with results published in the literature. An interface factor at the mudmat/soil interface = 1 was assumed for this comparison. A normalized vertical capacity Vult/suoB = 10.45 was found. This compares well with the analyses results published by Salgado (2008) who found 10.42 using ABC program (Martin, 2004). When the interface factor is reduced to = 0.5, the vertical capacity Vult/suoB reduces to 9.8.

The MH failure envelope for the case without pug mark is shown on Figure 3 assuming a vertical load V so that v = 0.4, which is equivalent to a safety factor of 2.5 on the uniaxial vertical capacity. The results are plotted in a non-dimensional way: M/suoB² versus H/suoB. This case is for an inter-distance d = 2 m (i.e. d/B = 0.07).

0

0.2

0.4

0.6

0.8

1

1.2

1.4

0 0.1 0.2 0.3 0.4 0.5 0.6

M/S

u0B

² [-]

H/Su0B [-]

v=0,4 - No pug markv=0,4 - Pug markv=0,71 - Pug markLS:GEO verification

Figure 3. Influence of pug mark on MH failure envelope

A few comparison runs were performed with Limitstate:Geo v2 (Limistate, 2009). Limitstate:Geo is designed to analyze the ultimate limit (or “collapse”) state for a wide variety of geotechnical problems. The current version is however limited to 2D plane strain analyses. The ultimate limit state is determined using the Discontinuity Layout Optimization (DLO) algorithm (Smith & Gilbert, 2007). The agreement with the FE results is found to be excellent.

The effect of the pug mark on the MH failure envelope is significant. The normalized uniaxial vertical capacity Vult/suoBis reduced to 5.5 leading to a mobilisation ratio v = 0.71 instead of 0.4 (i.e. a safety factor of 1.4 instead of 2.5). The moment capacity is reduced by 55% to 80% depending on the applied horizontal load.

The inter-distance between the mudmat and the pug mark was then progressively increased and results are presented on Figure 4. The maximum moment capacity increases progressively with the inter-distance towards the capacity obtained for the case without a pug mark. From an inter-distance d/B = 0.2, the difference in mobilisation ratio of the vertical capacity does not affect the maximum moment capacity and, at an inter-distance d/B = 0.5, the pug mark has no effect at all.

0.0

0.2

0.4

0.6

0.8

1.0

1.2

0 0.2 0.4 0.6 0.8 1

M/S

u0B

² [-]

Inter-distance d/B [-]

V=0.4Vult - 2D

V=V0 - 2D

No pugmark

Figure 4. Influence of pugmark / mudmat inter-distance on maximum moment capacity – 2D analyses

5.2 3D Analyses

The accuracy of the 3D FE model was also checked by computing the uniaxial vertical capacity for a square footing resting on soft clay, with su constant with depth and = 1, and comparing with available literature results. A normalized vertical capacity Vult/suoB² = 5.96 was found. This compares well with the results published by Gourvenec et al. (2006) who found 6.02. When the interface factor is reduced to = 0.5 and su increases with depth (as defined in Section 2), the vertical capacity Vult/suoB² is about 9.1, which is slightly lower that the 2D plane strain capacity.

Similarly to the 2D plane strain FE analyses, the MH failure envelope for the case without a pug mark assumes a vertical load V so that v = 0.4. As discussed above and shown on Figure 1, the 3D analyses consider two positions for the pug mark: along the width and at the corner of the mudmat. The results are plotted on Figures 5 and 6 for the first and second positions, respectively, using the following non-dimensional groups: M/suoB³ and H/suoB². This case is for an inter-distance d = 2 m (i.e. d/B = 0.07). The moment capacity is not affected for small mobilisation ratios of the horizontal capacity. However, when H/suoB approaches 0.5, the failure mechanism switches rapidly from a general shear mechanism to a sliding mechanism.

For the first position (along the width), the effect of the pug mark on the MH failure envelope is noticeable but not as significant as for the 2D plane strain simulations. The normalized uniaxial vertical capacity Vult/suoB² is reduced to only 8.4 leading to a mobilisation ratio v = 0.44 instead of 0.4 (i.e. a safety factor of 2.29). The moment capacity is reduced by 20% to 28% depending on the applied horizontal load. There is very little difference in the results between a vertical mobilisation factor of 0.4 and 0.44 (Figure 5).

0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

0 0.1 0.2 0.3 0.4 0.5 0.6

M/S

u0B³

[-]

H/Su0B² [-]

v=0,4 - No pug mark - 3D

v=0,44 - Pug Mark - 3D

v=0,4 - Pug Mark - 3D

²

Figure 5. Influence of pug mark on MH failure envelope – Position 1 (along width of mudmat)

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For the second position (at the corner), the normalized uniaxial vertical capacity is not affected, meaning that the safety factor for a pure vertical load remains 2.5. The moment capacity is only reduced by 1 to 3 % (Figure 6) depending on the applied horizontal load. This is a negligible difference.

0.0

0.2

0.4

0.6

0.8

1.0

1.2

1.4

0 0.1 0.2 0.3 0.4 0.5 0.6

M/S

u0B

³ [-]

H/Su0B² [-]

v=0,4 - No pug mark - 3D

v=0,4 - Pug mark - 3D

²

Figure 6. Influence of pug mark on MH failure envelope – Position 2 (at mudmat corner)

Similarly to the 2D plane strain approach, the inter-distance between the mudmat and the pug mark was progressively increased for the position of the pugmark along the width of the mudmat and the results are presented on Figure 7. The maximum moment capacity increases progressively with the inter-distance towards the capacity obtained for the case without pug mark. From an inter-distance d/B = 0.25, the pug mark has no effect anymore. This result illustrates the benefit of considering a more realistic 3D analysis when facing this kind of problem. The inter-distance required to have no influence of the pug mark in the 3D model is half the inter-distance required in the 2D plane strain model.

Finally, Figures 5 and 6 allow the comparison of the effect of the orientation of the moment and the horizontal loads on the MH envelope. The moment capacity for the loads in the direction of the corner is about 5% lower than for the case where the loads are towards the width of the mudmat. This geometrical effect is however limited.

0.0

0.2

0.4

0.6

0.8

1.0

1.2

0 0.2 0.4 0.6 0.8 1

M/S

u0B

³ [-]

Inter-distance d/B [-]

V=0.4Vult - 3DNo pugmark

Figure 7. Influence of pugmark / mudmat inter-distance on maximum moment capacity – 3D analyses

5.3 Conservativeness of 2D Analyses

Simplified 2D plane strain and more realistic 3D simulations give similar results if the pug mark is not considered.

In the case where the pug mark is located along the width of the mudmat, the MH failure envelope is degraded in both 2D and 3D analyses. However, the effect is significantly larger in the 2D analyses for which the moment capacity is reduced by

55% to 80% depending on the applied horizontal load. In the more realistic 3D analyses, the moment capacity is only reduced by 20 to 28%. When the pug mark is located at the corner of the mudmat, the 3D analyses show very little impact on the mudmat capacity.

The analyses show also that the distance of influence of the pug mark on the mudmat stability is 2 times less in 3D compared to 2D analyses. The maximum moment capacity is not affected from an inter-distance d/B = 0.25 in 3D analyses while the distance is d/B = 0.5 in 2D analyses.

1 CONCLUSION

The presence of a pug mark has been found to degrade significantly the yield surface of a square mudmat in the VHM load space. However, a comparison between simplified 2D plane strain and 3D analyses has shown that the beneficial 3D effects are substantial. If the pug mark is located along the width of the mudmat, the more realistic 3D model shows that the moment capacity is only reduced by 20 to 28% depending on the applied horizontal load. The impact of the pug mark is significantly larger when a more simplified 2D plane strain approach is followed. Moreover, in the particular example treated in this paper, it was observed that a pug mark located at the corner of the mudmat does not influence its stability. The zone of influence of the pug mark is also much more limited when the problem is modeled in 3D and the orientation of a complex VHM loading scheme can be considered in the global stability. Simplified 2D plane strain simulations can lead to over-conservative results for this particular problem.

6 ACKNOWLEDGEMENTS

The authors acknowledge the permission of Fugro GeoConsulting to publish this work and the guidance and review provided by Dr Richard Jewell.

7 REFERENCES

Dean E.T.R. 2010. Offshore geotechnical engineering – Principles and practice. Thomas Telford, London.

Gouvernec S., Randolph M.F. and Kingsnorth O. 2006. Undrained bearing capacity of square and rectangular footings. Int. J. Geomech. 6, N°3, 147-157.

Hossain M.S., Dong D., Gaudin C. and Kong V.W. 2012. Skirted spudcans and perforation drilling for installation of spudcans close to existing footprints. Proceedings of the 7th Intern.Conf. Offshore Site Investigation and Geotechnics, London.

Salgado R. 2008. The Engineering of Foundations. McGraw-Hill, New-York.

Plaxis 2011. Finite element code for soil and rock analyses, Version 2011. Plaxis BV. Delft, Netherlands.

Limitstate Ltd 2009. Geotechnical software for stability analysis, Version 2. Limitstate Ltd. Sheffield, UK.

Smitts C. and Gilbert M. 2007. Application of discontinuity layout optimization to plane plasticity problems. Proc. of the Royal Society A.

Martin C.M. 2004. User guide for ABC – Analysis of bearing capacity.Department of Engineering Science, Oxford University, Oxford.

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Design and installation of buried large diameter HDPE pipelines in a coastal area

Project et installation de tuyaux enterrés de grand diamètre en zone côtière

Bellezza I., Mazzier F., Pasqualini E. Dep. SIMAU – Università Politecnica delle Marche, Ancona, Italy

D’Alberto D., Caccavo C., Serrani C. SPS - Società Progettazione Servizi s.r.l., Ancona, Italy

ABSTRACT: The present paper deals with the main geotechnical aspects of the design and installation of two adjacent HDPE largediameter pipelines along the Adriatic Sea (Italy) coastline. The pipelines -270 m in length and 2 m in diameter - are conceived as buried collectors of polluted runoff water, to convey to sanitation prior to discharge into the sea. Considering that pipes are below thewater table uplift analysis is detailed, showing three possible approaches in static conditions, whereas in seismic conditions a methodis proposed that include the build-up of pore-water pressures during earthquake. As far as prediction of vertical deflection isconcerned, the backfill loosening due to sheet piles extraction has been modelled by assuming no compaction (dumped backfill).Despite this assumption, theoretical short term deflection represents a lower bound of measured deflections.

RÉSUMÉ : Cet article décrit les principales problématiques géotechniques du projet et de l’installation de deux tuyaux enterrésadjacents de grand diamètre sur le littoral adriatique italien. Les tuyaux – longs de 270 m - ont la fonction de collecteurs enterrés pourl’eau de ruissellement polluée, pour permettre sa dépuration avant du déchargement dans la mer. En considérant que les tuyaux setrouvent au-dessous du niveau de la nappe d’eau on a analysé le problème du possible soulèvement en conditions statiques utilisanttrois différentes méthodes. En conditions sismiques on a proposé une méthode qui considère le développement de pressionsinterstitielles positives excessives durant le tremblement de terre. L’ovalisation du tuyau a été calculée par une méthode de littératureen considérant un remblai sans compactage pour tenir compte de l’extraction des palplanches utilisées pendant l’excavation de latranchée. Les valeurs calculées de l’ovalisation initiale représente un limite inferieur de l’ovalisation mesurée.

Keywords : uplift, pipe of large diameter, deflection

1 INTRODUCTION

Urban and infrastructural development often involves vulnerable areas such as coastlines. To prevent pollution of the sea from runoff water of a nearby urban area and crowded roads a system of buried collectors are to be built along a stretch of the Italian shoreline of the Adriatic Sea. In such a way, the collected runoff water will be conveyed to sanitation before discharging into the sea. A preliminary hydraulic study allowed to define different drainage basins, and for each basin an adequate collector is required. This paper deals with the design and execution of the first part of the system, concerning a collected water volume of about 1300 m3, for which the designers opted to realise the collectors by two adjacent buried pipelines of 2000 mm in internal diameter and 270 m long.

ADRIATIC SEA

COLLECTORRAILWAY

PIEZOMETER

Figure 1. Plan view of the working area

The design and execution of geotechnical works in coastal area must in general face regulatory requirements, environmental and aesthetic concerns, public attention.

As far as engineering problems are concerned, the designer must take into account the objective difficulties connected with the critical location (e.g., underwater excavation, tidal and storm waves, risk of uplift). For the specific case, additional constraints are represented by the closeness of the working area to an important railway (Figure 1), and consequent limited accessibility for materials and machinery. Moreover, local authorities required to minimize the working area, avoiding the occupancy of the beach for the overall length of the collectors through the entire duration of the works. To comply with this requirement, a staged execution was envisaged.

2 OUTLINE OF THE DESIGN

2.1 Pipe material

Corrugated HDPE pipes were selected (Table 1). HDPE offer ingeneral significant advantages in terms of costs, corrosion resistance, ease of handling and jointing over more traditional materials such spheroidal cast iron. The selected HDPE pipes are manufactured in modules 6.0 m long (less than other materials, e.g., cast iron) which allowed solving the problem linked to area accessibility. In particular with the use of 6 m long modules it is possible to reach the beach area passing through a narrow railway underpass. For longer pipe modules, a more expensive marine transportation would be requested.

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Table 1. HDPE Pipe properties

Parameter Weight for unit length (kN/m) 2

External diameter (mm) 2240

Internal diameter (mm) 2000

Moment of Inertia (mm4/mm) 45899

Young modulus short term(MPa) 1185

Young modulus long term (MPa) 288

2.2 Underwater excavation

The soil stratigraphy is essentially composed by a sandy layer 3.5-4.0 m thick overlying a cohesive bed. An open standpipe piezometer installed close to the working area (Fig.1) indicated that the groundwater table is located 1.0-1.5 m below the ground level. Fig.2 shows a typical CPT profile with the characteristic values of geotechnical parameters obtained by laboratory and in situ tests.

Considering the large pipe diameter, the bedding layer and a minimum soil cover to counteract buoyancy (as detailed later), an excavation depth of at least 4.70 m was necessary. Moreover, a minimum inclination of 0.5% to the horizontal is required to ensure gravity flow. This results in an excavation depth ranging form 4.7 m to 6.0 m from ground surface.

Various techniques were considered for the excavation. Unsupported trench with inclined sidewalls was excluded due to excessive breadth to ensure stability and the need for continuous dewatering by well-points. Other equally suitable technologies, (e.g., soil freezing), were incompatible with the budget.

The selected solution consisted in a 6.1 m wide trench supported by strutted sheet piles, embedded in the impervious clay layer. The total length of the sheet piles varied between 8 m and 10 m as depending on excavation depth. Sheet piling allowed retaining the vertical trench walls, minimizing seepage into the trench and protecting the working area from tidal and storm waves (the top of sheet piles was +1 m above g.l., Fig. 2).

To comply with the requirement of minimizing occupation of the area, the installation of the collectors (270 m) was realized in four distinct segments (i.e. the excavation in a zone starts only after the work in the previous zone is completed). For the first segment, sheet piles were preliminarily installed to enclose a rectangular excavation zone, creating a continuous barrier to groundwater along the entire perimeter. For the subsequent segments, the presence of the installed pipes prevented to create rectangular hydraulic barrier by sheet piles only. Therefore, cast-in-place concrete waterproof screens were designed around the protruding edge of the pipes to block seepage due to extraction of sheet piles from the adjacent completed segment.

2.3 Pipe uplift

During the service phase the pipelines are expected to be only rarely filled by runoff water but permanently submerged by groundwater and then subjected the buoyancy. Consequently, the design shall be checked against failure by uplift.

L=8-10 m

B = 6.10 m

Hexc>4.7 m

SHEET PILES

STRUT

SILTY CLAY = 20 kN/m3

cu= 80 kN/m2

k =1*10-8 cm/s

SILTY SAND = 20.3 kN/m3

=40° k =1*10-3 cm/s

0 5 10 15 20

qc (MPa)

Figure 2. Simplified sketch of the excavation geometry with geotechnical soil characterization and a typical cone resistance profile.

According to Italian Building Code (NTC, 2008), as well as Eurocode 7 (2004), for any mass potentially subjected to the failure mechanism, the following inequality must be satisfied:

ddd RGV (1)

where Vd is the design destabilizing action acting upwards (obtained by a partial factor 1 = 1.1 in static conditions), Gd is the design stabilizing permanent action including the weight of the mass subjected to uplift (obtained by a partial factor2 = 0.9 in static conditions) and Rd is the design soil resistance by friction along the vertical contours of the assumed block.

Considering the closeness of pipes to the sea (Fig. 1) it can not be excluded that in the future a portion of the soil above the pipe can be eroded. To confer protection against erosion a cast-in-place concrete slab (6.05 m wide and variable thickness) is realized above the pipes, as illustrated in Figure 3. This solution allows also to increase the average unit weight of the material above the pipes and enlarge the size of the block subjected to uplift. Finally, it represents a protection against accidental damage due to anthropic activities and the superficial sand layer enables to continue the recreational use of the beach.

In the application of Eq. (1) different approaches can be adopted to calculate the term Vd and in Gd. Eurocode 7 indicates a total stress analysis for uplift problems (EC7, 2004 §10.2). According to this approach, Vd is the upward resultant of pore water pressure acting on the lower boundary of the assumed block. Consistently, Gd includes the total weight of the soil block above the pipes. However, following this approach, the resultant of pore water pressure acting downwards is multiplied by a partial safety factor (1 =1.1) different to that applied to the vertical upward resultant 2 = 0.9). This results in a violation of the “single source principle” enunciated by Eurocode 7 (EC7, §2.4.2). According to this principle, when destabilising and stabilising permanent actions come from a single source, “a single partial factor may be applied to the sum of these actions or to the sum of their effects”. Based on the above considerations, in the second approach the destabilizing action is assumed to be the buoyancy force on the two submerged pipes (i.e. the weight of the water displaced by the pipes Ww).Consistently, Gd includes the submerged weight of the block above the pipes. Finally, a third approach can be used in which the destabilizing action is assumed to be the resultant buoyant force of the submerged pipes, i.e. the algebraic sum of weight of displaced water Ww and weight of pipes Wp (WSSC, 2008). This latter approach implies that the check against failure by uplift is automatically verified when Ww < Wp .

The three approaches described previously are applied assuming the simplified sliding surface shown in Fig. 3, which implies a failure mechanism involving pipes, slab and soil above and between the pipes as well (hatched zone in Fig.3). The results were obtained for the worst-case scenario of complete erosion of the superficial sand layer (h3 = 0 in Fig 3) and minimum cover thickness above the pipes (h1 + h2 = 0.5 m, s = 0.6 m). The unit weight of concrete and saturated soil were 23.5 kN/m3 and 20.3 kN/m3 respectively.

The Rd term was calculated as the sum of the friction forces along the vertical planes on each side of the assumed block (BC, B’C’, DE, D’E’)

2221

3

2

3)(tan'tan' sRhhsKsKR e

ks

BCsd

(2)

where BC is the interface friction angle between concrete and the sandy soil (BC = 30°), is the shear resistance angle of the granular backfill, 3 = 1.25 is the partial safety factor applied to the shear strength parameters. The angle k after backfilling is assumed to be 40°- 42°, but a reduced value of 38° is assumed in eq. (2) owing to potential loosening induced by sheet pile

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extraction. Ks is the lateral earth pressure coefficient assumed conservatively equal to 0.5 by neglecting the effect of compaction.

Results of static analysis of uplift failure (Table 3) indicate that, as expected, the second and third approaches are less conservative in static conditions; however they appear to better represent the physical reality (i.e. the destabilizing force on a fully submerged pipe is independent of the depth below the water table).

In seismic conditions, Italian Building Code (NTC, 2008 §7.11.1) requires that Eq. (1) shall be checked using 1 =2 =1,which result in uniqueness of the approach in seismic analysis(i.e. the difference between Gd and Vd is the same using a total stress analysis or an effective stress analysis).

Using a pseudo-static approach the vertical inertial force FVacting on pipes, soil, concrete is assumed upwards and proportional to relevant weights by the seismic coefficient kv(=0.046) defined in Italian Building Code (NTC, §7.11.3.5.2).

It is well recognized that in the presence of earthquake, a build-up of pore water pressures can occur with respect to static conditions. A simplified procedure to account for this phenomenon is the introduction of the pore-pressure coefficient ru = u/’v0 assumed constant with depth (e.g. Ebeling and Morrison, 1992; Kramer, 1996). Accordingly, the unit weight of water and submerged soil are given by:

wsatuwuwwe rr ' (3)

uwsatue rr 11'' (4)

For ru = 1 soil liquefaction occurs which implies that (a) upward action acting on pipes is proportional to we (= sat)instead of w. and (b) the submerged weight of the soil block and soil resistance Rd vanish.

In seismic analysis the coefficient ru shall be selected on the basis on seismic input (magnitude and maximum acceleration), as well as soil characteristics. In the analyzed case the presence of a coarse backfill (gravel) around the pipes (Fig.3) is expected to strongly limit the build-up of pore water pressure. Hence, the seismic analysis of uplift are performed assuming ru = 0 and ru= 0.1. Results of seismic analysis shown in Table 3 indicate that also for ru = 0.1 the inequality (2) is satisfied.

2.4 Pipe deflection

Flexible conduits fail by excessive deflection rather than by rupture of the pipe wall. It is necessary, therefore, to estimate the deflection of this type of conduit and to establish limits of allowable deflection for the proposed installation.

Table 3. Uplift analysis Static conditions Seismic conditionsAction

(kN/m) appr.1 appr.2 appr.3 ru = 0.1 ru = 1

Vd 174.1 86.6 82.3 86.6 159.9 Wpipes 4.0 4.0 - 4.0 4.0

Wsoil 87.6 44.4 44.4 40.0 0Wslab 85.3 49.0 49.0 45.3 11.6 Fv 0 0 0 -8.1 -8.1 Gd 159.2 87.7 84.1 81.2 7.5 Rd 15.6 15.6 15.6 13.4 0

Rd+Gd 174.8 103.3 99.7 94.6 7.5

h3

H s

B

d

A’

C

B’

E

D’

F

h1

D

E’

G G’

clay

GRAVELLY SOIL

Hexc

Concrete slab

Native sand

F’

h0

h2

A

B

C’

De

Hw

bedding

haunching

Figure 3. Details of pipe installation

For flexible conduits the vertical deflection y mainly depends on the actual load acting on the pipe and the stiffness of the backfill at the side of the pipes whereas the contribute of pipe stiffness is generally small (i.e. Rogers et al. 1995). In the specific case, the presence of two pipes as well as of the concrete slab makes the analysis more complex than the classical solutions available in the literature.

2.4.1 Load on pipes According to Young and Trott (1984), the pair of pipes can

be considered as equivalent to a single pipe of overall width D’where D’ = 2De+d. The load on D’ is calculated by taking the lesser of the two values obtained by the complete ditch condition (P1) and the positive projection condition (P2). This load is taken as being shared equally by the two pipes.

The load acting on a pipe (stiffer than side fill) in a trench with a partially submerged homogeneous backfill (see Fig.3) is given by Bulson (1985):

BHB

BHBP ww 2exp1

22exp1

2

22

1

(5)

where is a coefficient ranging from 0.11 to 0.19 depending on soil type, is the total unit weight of backfill. In the analysed case actual backfill unhomogeneity is accounted for by assuming a weighted average unit weight (av).

For a positive projection conduit the P2 value depends on the relative settlement between the soil prism above the pipe and the adjacent soil, which determines positive or negative arching. In the specific case, the presence of the slab prevents the occurrence of complete ditch or projection conditions. Moreover, considering that the ratio H/D’ is very small, it is reasonable to neglect arching. Therefore, the value of P2 is assumed to be the weight of overlying prism of width D’.

Obviously, the maximum deflection is expected to occur at the section with the maximum cover (3.30 m) with the lowest groundwater level (-1.60 m below g.l.). With reference to Figure 3, for h1 = 1 m, s = 0.30 m, h2 =1.7 m, h3 = 0.3 m, Hw=1.7 m av = 20.85 kN/m3 = 0.19, P1 and P2 are calculated as 281 kN/m and 264 kN/m, respectively. Following the suggestion of Young and Trott (1985), the load acting on a single pipe (P) is taken as 132 kN/m.

2.4.2 Backfill A large part of ability of flexible pipes to support vertical

load must be derived from the passive pressures induced as the sides move outward against the soil. Therefore, any attempt to analyse the behaviour of the flexible conduits must take into account the soil at the sides as an integral part of the structure, since such a large proportion of the total supporting strength is attributable to the side material.

Considering that compactive effort is restricted by the geometry of the trench and the difficult in compacting underneath the pipe in the haunch zone (Fig.3), as well as the

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sensitivity of installed flexible pipe to compaction of material around it, a clean gravelly soil has been selected as structural backfill (i.e. the part of backfill that extends from the base of the bedding to a maximum of 30 cm above the pipe, as shown in Fig.3). This coarse-grained soil is preferred over native silty sand for easy of compaction, high earth pressure response and stability when saturated and confined. The same material – well compacted - has been used also as bedding soil (Fig.3).

2.4.3 Calculation of pipe deflection The pipe deflection is predicted by the method of Spangler

(1941) or Iowa formula, although it is well recognized that this method contains some debatable assumptions.

'E.REJPKDy L

06103

(6)

where y = vertical deflection of pipe (m); P = vertical load on the pipe (MN/m); EJ = flexural pipe stiffness (MNm2/m); R= mean radius of the pipe (m); DL= time-lag factor (-); K =bedding constant (K = 0.1 for bedding angle = 60°, see Fig.3). E’ = horizontal modulus of soil reaction (MPa).

Considering the absence of vehicular loading and the prevalent recreational use of the site, live loads have been neglected. In design the sheet pile extraction is accounted for using the value of E’ relevant to a dumped backfill (Table 4). In long term analysis a time lag factor of 1.5 and a reduced pipe stiffness are considered (see Table 1). With the above assumptions short term and long-term deflections are calculated as 10.1 cm and 20.5 cm , respectively.

Numerous authors have reported that pipes have been distorted by 10-20% and still continue to perform adequately. Therefore the theoretical deflections have been considered acceptable, but a monitoring activity was planned during installation.

3 COMPARISON OF ACTUAL AND THEORETICAL DEFLECTIONS

The large diameter of pipelines allowed accessibility and direct measurement of vertical diameter at prescribed positions during the various stages of installation (structural backfilling, slab, final backfilling and sheet piles extraction). The trend of measured vertical deflection versus time is not monotonic, showing an initial small deflection, followed by a slight decrease, a sharp increase and a final stabilisation. The observed trend can be ascribed to the variation of the acting loads (backfill height and groundwater level) and the different lateral support offered by the soil before and after the extraction of sheet piles. Therefore, for the comparison between actual and theoretical deflections, only the stabilised values are considered because they better represent the service conditions of the pipes, with the groundwater level certainly above the crown of the pipes.

With reference to a pipe stretch 45 m long the vertical deflections were measured in sections spaced 3 m apart. Final (stabilized) deflections are shown in Fig. 4. In spite of a quite uniform cover height the measured deflections vary considerably along the pipeline with a minimum of 7 cm to a maximum of 15 cm. This behaviour can be attributed mainly to inherent differences in compacting the soil beside the pipes and variable effect of sheet pile extraction. Moreover, variability in stiffness of native soil can influence the overall response owing to the closeness of pipes to trench sides.

Considering that measurements refer to a design cover height ranging from 2.15 m to 2.37 m, the vertical deflection is calculated by (6) for an average cover height H = 2.26 m. The load on pipe (P = 104 kN/m) is calculated following the suggestion of Young and Trott (1984) discussed previously. The groundwater level was assumed at 1.6 m below the ground

surface (Hw = 0.66 m) based on measurement in the nearby piezometer. As shown in Fig. 4, the theoretical deflection calculated by Spangler method is lower than the actual average deflection. This can be ascribed to effect of sheet pile extraction which results in a loosening of backfill and a probable increase of the load on pipes to due to negative arching.

Table 4. Values of E’ for a clean coarse-grained soil (Howard, 1977)

Degree of compaction dumped slight moderate

E’(MPa) 1.4 6.9 13.8

4 CONCLUSIONS

In the present paper some aspects of design and installation of two adjacent large diameter pipelines along the Adriatic Sea coastline in Italy are described.

Uplift analysis is detailed, showing three possible approaches which lead to different results in static conditions, whereas in seismic condition a unified approach is proposed that account for build-up of pore-water pressures.

As far as prediction of vertical deflection is concerned, in the analyzed case the backfill loosening due to sheet piles extraction has been modelled by assuming no compaction (dumped backfill). Despite this assumption, theoretical short term deflection represents a lower bound of measured deflections.

4

6

8

10

12

14

16

0 10 20 30 40 50 6

vert

ical

def

lect

ion

(cm

)

0

sea-side pipe

railw ay-side pipepredicted by (6)

average of measured deflections

Figure 4. Comparison between measured and predicted short–term deflections

5 REFERENCES

Bulson P.S. (1985) “Buried Structures. Static and Dynamic strength”.Chapman & Hall ed.. London.

Ebeling, R.M., Morrison, E.E. (1992). “The seismic design of waterfront retaining structures”. Technical report ITL-92-11.Washington, DC. US Army Corps of Engineers.

EC7 (2004). Eurocode 7: Geotechnical design – Part 1: General Rules. European Committee for Standardization.

Howard A.K. (1977) “Modulus of soil reaction values for buried flexible pipeline”. J. of Geotech. Eng. Div. ASCE 103 (1), 33-43.

Kramer, S.L. (1996). Geotechnical Earthquake Engineering. Pearson Education Inc. New Jersey.

NTC (2008). Norme Tecniche per le Costruzioni. D.M. 14/01/2008. (in Italian).

Rogers C.D., Fleming P.R., Loeppky M.W. and Faragher E. (1995) “Structural performance of profile-wall drainage pipe – stiffness requirements contrasted with results of Laboratory and Field Test”. Transportation Research Record, 1514, 83-92.

Spangler M. G. (1941) The structural design of flexible pipe culverts”. Bul. 153. Iowa Engineering Experiment Station, Ames, Iowa.

Young O.C., Trott J.J. (1984). Buried rigid pipes: structural design of pipelines. Elsevier Applied Science Publishers. London-New York.

WSSC (2008) Pipeline Design Manual. Part Three - Common Design Guidelines. Washington Suburban Sanitary Commission. www.wsscwater.com/home/jsp/misc/siteMap.faces

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Geotechnical Exploration for Wind Energy Projects

Compagnes géotechniques destinées aux parcs éoliens

Ben-Hassine J. Renewable Energy Systems Americas Inc., Broomfield, Colorado, USA

Griffiths D.V. Colorado School of Mines, Golden, Colorado, USA

ABSTRACT: Wind energy projects are often fast-paced and cover large terrains. Such conditions result in increased geotechnicalrisks and require specially adapted geotechnical exploration and data analysis techniques that are designed to manage risks at different stages of project development. Use of geophysical methods, in addition to the traditional subsurface exploration methods, is generallyrequired to collect design critical data. During the early stages of project development, using quick qualitative geophysical methods can prove advantageous for finalization of wind farm layout and preliminary foundation design. However, as project plans progress, amore thorough geotechnical investigation is required. At all stages of a project, an understanding of the available geotechnical tools, along with their associated risks and cost implications is essential to minimize the likelihood of design changes that result in cost overruns. This paper presents geotechnical exploration methods used at different stages of project development and discusses key geotechnical parameters for wind turbine foundation design, available geotechnical tools, and the degree of confidence associatedwith these tools. The paper makes an attempt to present an exploration approach that is optimized for efficiency and risk.

RÉSUMÉ : Les projets d'énergie éolienne sont souvent réalisés dans un contexte d'exécution rapide et couvrent des terrains degrandes envergures. Ces conditions présentent des risques géotechniques accrus et nécessitent des compagnes d'exploration géotechnique et des techniques d'analyse de données spécialement bien adaptées pour gérer les risques à différentes étapes du projet.Le recours à des techniques géophysiques en plus des méthodes d'exploration traditionnelle est généralement requis pour obtenir les données critiques. Durant les premières étapes du projet, le recours à des méthodes géophysiques qualitatives et rapides peut s'avérerplus avantageux pour établir '' la faisabilité du projet, '' le plan d'implantation du projet et la conception préliminaire des fondations. Toutefois, dans les étapes plus avancées, une étude géotechnique plus poussée doit être réalisée. A toute étape, une connaissance adéquate des méthodes géotechniques disponibles, des risques et coûts qui leurs sont associés est essentielle pour minimiser l’éventualité de changements à la conception résultant en dépassement de coûts. Cet article est un essai de présenter une approched'exploration optimisant l'efficacité et le risque.

KEYWORDS: geotechnical exploration, risk management, wind energy, efficiency.

1 INTRODUCTION

The period leading up to an operational wind energy plant starts several years before construction and can be divided into three overlapping phases: project development, engineering design, and project construction (Figure 1). During the development phase, various risk types and sources are evaluated and decisions are made to maintain, modify, or abandon the project. During the engineering design phase, decisions are made to refine the design while maintaining acceptable levels of risk. Any subsequent changes to the design typically result in additional cost. This paper focuses on geotechnical risks, particularly how such risks are being addressed currently and how this process may be improved. The objective is to assess risks and catch flaws as early as possible in the project development phase while there are still opportunities to make changes before significant development funds are spent. As in all large expenditure projects, early decisions have the greatest impact on financial performance. The motivation of this paper is to minimize the cost of civil infrastructure related to wind energy projects (turbine foundations, access roads and facilities such as the substation and the operation and maintenance building) through a rational redistribution of the geotechnical exploration effort. It has been estimated that civil infrastructure accounts for 4 to 10% of the total wind energy project cost. Given the thin profit margins of wind energy projects, a 2% saving can make the difference between whether a project goes ahead or is shelved.

Figure 1. Overlapping phases leading up to a wind energy plant.

1 PROJECT REALISATION PHASES

All three project phases (development, design, and construction) involve some level of geotechnical risk assessment and management, with most of this effort currently focused at the engineering design phase. Current and proposed activities related to geotechnical risk assessment for each phase are described below.

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1.1 Project Development Phase

During the project development phase, the developer is usually focused on wind resource assessment, land agreements, power purchase agreements, and identifying potential investors. The geotechnical aspect is secondary and is typically limited to site visits and reporting of surficial characteristics such as terrain topography, accessibility, proximity to bodies of water, etc. The initial environmental permitting effort presents an opportunity to identify geotechnical conditions that carry cost implications as most environmental permitting efforts include an evaluation of geo-environmental conditions. However, the development phase of a wind project rarely includes geotechnical field investigation activities. However, the development phase is the most opportune time to identify significant geotechnical risks. The findings of a preliminary geotechnical investigation conducted during the development phase rarely render the project non-pursuable. However, preliminary geotechnical investigations are critical to proper planning and allocation of risks to the appropriate stakeholders. The achievement of the benefits of this proposed shift can be formalized through techniques such as geostatistics, Bayesian updating, statistical inference and neural networks (Christian et al. 2006 and Lin and Hung 2011). An immediate benefit of a more holistic development phase exploration is to focus the detailed exploration effort on the critical issues or portions of the project area. In addition to desk studies, recommended development phase exploration techniques include: Geophysical surveys using seismic methods such as Multi-Analysis of Surface Waves (MASW) at all proposed turbine locations, possibly excluding locations where rock is at the surface. An MASW survey provides depth profiles of shear and compression wave velocities. The information is used to gain an insight into site stratigraphy and to estimate elastic moduli needed to verify foundation stiffness requirements. The MASW survey, conducted at the project development phase, helps to identify soft locations or locations with potential difficulties as an aid to micrositing of turbines. This exercise lessens the likelihood of needing very large foundations or performing costly ground modifications at soft sites. MASW surveys are also quick and relatively low cost, making them the ideal qualitative tool that is suited for the development phase. Preliminary geotechnical exploration borings using a limited number of traditional SPT, SCPT, CPT, or DMT borings spread over different portions of the project area. Exploration pits may also be used along planned access road alignments. Information obtained through site visits and review of available published information and online maps can be used to decide on the locations of the preliminary borings so that the captured range of variability is as wide as possible. Information from the preliminary exploration is used to assess the type and range of variability of site geomaterials, to identify potential foundation types and to plan the full investigation. For example, if a gravity base (shallow) foundation is deemed feasible, an effort should made at the project development stage to assess the depth range of the stationary groundwater table in order to decide if buoyancy will be a design consideration. If soft materials are encountered requiring consideration of deep foundations, the depth of borings during the full investigation can be adjusted. The preliminary geotechnical exploration should also include electrical and thermal resistivity testing as this input is critical to sizing the electrical collection system which is associated with a significant share of project cost.

1.2 Engineering Design Phase

During this phase, a full geotechnical investigation must be carried out to finalize the design. The full geotechnical study is designed to complete the investigation and to fill the gaps remaining after the preliminary exploration. The full investigation should confirm and refine the assessment of the

risks identified during the preliminary investigation and should assess any additional risks that may be uncovered. At a minimum, standard practice includes at least one exploratory boring at every turbine location extending to a depth of interest not less than the largest base dimension of the structure (DNV/Risø 2002, GL 2010). For a typical shallow foundation used to support wind turbines, the explored depth is 1 to 1.5 times the foundation diameter. Common current practice is to perform geophysical testing during the full investigation phase at a limited subset (approximately 10%) of turbine locations. In the proposed redistribution of effort, a more extensive geophysical survey is recommended at the development phase. A non-exhaustive list of risks that should be assessed as early as possible during the development and preliminary design phases (but prior to the final design phase) is shown in Table 1. Geotechnical exploration activities help in identifying these risks but are not the sole resource.

Table 1. Non-exhaustive list of potential wind farm geotechnical and eo-environmental risks (in no particular order). g

No. Risk Identification tools

1 High groundwater Drilling, excavation pits, monitoring wells and permeability testing.

2 Flooding, storm surge, tsunami Records, maps

3 Shallow bedrock / blasting Visual, drilling, MASW

4 Slope stability & landslides Visual, geologic study

5 Mine subsidence Records, LiDAR, maps

6 Coal seams Drilling, records

7 Karst subsidence, caves & voids Records, drilling, LiDAR, maps, type of underlying rock, groundwater regime

8 Shrink/swell (expansive) soils Laboratory testing

9 Frost heave Records, climatic data

10 Permafrost Records, climatic data

11 Freeze-thaw Climatic records

12 Collapsible soils Laboratory testing

13 Excessive consolidation / tilt Laboratory testing

14 Aggressive environments: high sulfates, high salinity, corrosion

Laboratory testing

15 Alkali-Silica Reaction (ASR) Testing, local information

16 Peat bogs and soft grounds Visual, drilling, MASW

17 High seismicity / liquefaction Exploration, design codes

18 Hurricanes Records, design codes

19 Volcanic activity Records, geologic study

20 Scarcity of gravel / road base Visual, local information

21 Buried pipelines & infrastructure Records

22 Forest logging roads Drilling, excavation pits

23 Drifting sands Visual, local information

24 High soil electrical resistivity Field and lab testing

25 High soil thermal resistivity Field and lab testing

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1.3 Project Construction Phase

During the construction phase, geotechnical activity is typically limited to quality assurance testing which serves to confirm and ensure that the design assumptions remain valid. This is the phase where risks missed during the earlier phases may become apparent with the potential for project cost overruns.

Rarely would geotechnical input in this phase result in cost savings. However, value engineering where the balance of plant (BOP) contractor is provided an opportunity to redesign is becoming more popular. Value engineering often occurs shortly before construction or as the BOP contractor is mobilizing to construct the project. Ironically, the likely reason for value engineering is the tendency of the original designer to err on the conservative side because of compressed schedules and/or lack of substantive geotechnical basis of design at the end of the development phase, creating opportunities for the BOP contractor to cut costs at the last minute.

1.4 Summary of Current and Proposed Practice

Table 2 shows a summary of current and proposed practice. The essence of the proposed redistribution of the geotechnical exploration effort is to advance the geophysical survey and the preliminary investigation to the development phase (P1). Details of the geotechnical activities for the proposed redistribution are shown in Table 3.

Table 2. Common and proposed geotechnical effort.

Common Proposed

Phase P1 P2 P3 P1 P2 P3

Desk study X X

Geophysical survey X X

Preliminary investigation X X

Full investigation X X

Assurance & validation X X

Phases: P1 = Development, P2 = Design, P3 = Construction

Table 3. Wind farm realization phases and proposed geotechnical ctivities.a

Phase Proposed minimum geotechnical activities

Development

Desk study: o Often required for permitting but

can be useful in planning preliminary investigation

Geophysical Survey o All turbine locations except

possibly sites where rock is at the surface

o Useful for micrositing Preliminary Investigation

o Drilling at a subset of turbine locations distributed strategically to capture maximum variability

o Excavation pits along potential access road alignment

o Electrical and thermal resistivity testing

o Limited laboratory testing

Design

Full Investigation o Drilling at all turbine locations o Extensive laboratory testing o Fill all gaps to form design basis

Construction

Construction QA/QC o Confirm validity of design

assumptions o Ensure compliance with design

requirements

2 SOURCES OF UNCERTAINITY

Wind energy projects differ from most traditional projects in that they cover large terrains. Wind turbines are typically placed 5 to 10 rotor diameters apart to optimize energy extraction (Denholm et al. 2009). Nowadays, typical rotor diameter for large wind turbine generators is around 120 meters, signifying turbine spacing of 0.5 to 1 kilometer just for energy extraction efficiency. Therefore, wind turbines are too far apart to consider any relationship between ground conditions from one turbine location to another. This is separate from regional or larger scale characteristics which may be applicable to the project area or portions of it, such as those related to different geologic settings or terrains. Turbine structures themselves are also unique due to the nature of loading they impart to foundations and supporting soils in terms of type, magnitude and variation. Thus, in addition to increased uncertainty due to essentially independent conditions at turbine locations, these projects also require parameters unique to these structures such as those needed to ensure adequate foundation stiffness.

Generally, there are three main sources of uncertainty in a geotechnical design property: i) inherent soil variability, ii) measurement error, and iii) transformation error (see Baecher and Christian 2003, Phoon and Kulhawy 1999). Often, a design parameter is not measured directly in-situ or in a laboratory test but is calculated based on other measured properties. Two of the above sources (inherent variability and measurement error) are associated with the measured property. The third source is associated with uncertainty in the selected transformation model, i.e., the empirical or theoretical relationship used to calculate the design property from the measured properties. Point estimates, as well as spatial variability of various shear strength, mechanical and index properties, are available in the literature (e.g., Lee et al. 1983). This information can be used to select the test methods that result in lowest variability depending on the soil type. In this section, uncertainty sources are discussed in more detail as they relate to wind energy projects.

2.1 Uncertainty Due to Inherent Soil Variability

Inherent soil variability is related to the natural geologic processes that produced the soil and should not include the influence of deterministic trends (e.g., trends due to depth), mixing of soils from different geologic units, or measurement errors. In the case of wind projects, inherent variability should be considered at each individual turbine location.

Another source of uncertainty is related to spatial variability extending vertically and horizontally to dimensions of influence. Uncertainty related to spatial variability is affected by the scale of fluctuation or correlation distance which is an important statistical parameter loosely defined as the distance within which the values of a given parameter are significantly correlated (Fenton and Griffiths 2008). Due to the often layered nature of soils, the correlation distance is typically shorter in the vertical direction than in the horizontal direction. Engineering design practice, including that within the wind energy industry, considers single (or point) variables to represent properties of an

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entire soil mass. Thus, in designing a shallow foundation for a wind turbine, for example, traditional practice assumes an infinite horizontal correlation length where a single value is assumed for the soil in each layer. Furthermore, while focus is on variation in the vertical direction, geotechnical exploration rarely goes beyond one boring at the center of the foundation unless there is strong reason to believe conditions are non-uniform in the lateral directions, such as in cavitose terrain. Thus, knowledge of the vertical spatial variation is often limited to the line of the boring. On the other hand, knowledge in the horizontal direction is limited to the observation and verification of the exposed foundation bearing surface. This is very limited information but standard practice. This is also why at least two forms of exploration should be carried out at each turbine location: a traditional boring and a seismic survey (MASW).

2.2 Uncertainty Caused by Measurement Error

Measurement uncertainty is related to the equipment being used, in-situ or laboratory test procedures, and random data scatter. Naturally, measurement error is different for different test procedures. Reported measurement error data have been summarized for various laboratory and field tests by various investigators (e.g. Phoon and Kulhawy 1999). It is worthwhile to note that the highest variability attributed to in-situ test measurement error is that corresponding to the Standard Penetration Test (SPT). The error introduced by sample size is sometimes considered as a measurement error. Normally, the greater the number of data points or sample size, the smaller the error. However, beyond a rather low number of samples, it is more important to capture the full range of variability than to obtain more data points. There are numerous simplified rules to estimate standard deviation and variability based on the range and number of samples (Tippett 1925, Withiam et al. 1997, Whitman 2000 and Foye et al. 2006). For this reason, the effort to capture the full range of variability as early as possible is very important to the early assessment of risks.

2.3 Uncertainty Caused by Transformation Error

Transformation or model errors are introduced when test measurements are used to calculate the desired design properties using empirical or theoretical relationships. The sources of the error include the fitting errors in the case of empirical equations and the simplification/idealization errors in the case of theoretical relationships. The transformation errors for several design properties (undrained shear strength, effective stress friction angle, Young’s modulus, horizontal stress coefficient, etc.) have been compiled (e.g. Phoon and Kulhawy 1999) for various laboratory and in-situ test methods. Noteworthy remarks from these compilations include:

Higher variability (as expressed in higher coefficients of variation) result for sand properties obtained though correlations with SPT blow counts, especially if “universal” empirical relationships are used; i.e., relationship not calibrated to a specific geology. Hence, “local knowledge” seems to be important for interpretation of SPT results.

Higher variability is typically obtained for sand properties than for clay properties.

3 CONCLUSIONS

Wind energy projects are almost always developed and built under compressed schedules where project realization phases overlap. They also cover large terrains that involve wide variability of geotechnical and geo-environmental conditions. For these reasons, geotechnical risks must be addressed as early as possible during the development phase to avoid overlooking fatal hazards that can shelve or financially devastate the project. This paper proposes to conduct extensive, low cost and quick

geophysical surveys during the development phase to help with turbine micrositing and to gain an insight into the variability of the entire project area. The paper lists potential hazards that should be assessed and discusses sources of geotechnical uncertainty and how they relate to wind energy projects.

4 ACKNOWLEDGEMENTS

The author would like to acknowledge his employer, Renewable Energy Systems Americas Inc, for support in the preparation and presentation of this paper.

5 REFERENCES

Christian, J.T. and Baecher, G.B. 2006. The meaning(s) of statistical inference in geotechnical practice, Proceedings of the 2006 GeoCongress, ASCE, Atlanta, GA, 26 Feb. – 1 Mar., 2006.

Lin, C.P. and Hung, Y.C. 2011. Parameter estimation and uncertainty analysis incorporating engineering judgment and Bayesian inversion, Proceedings of Georisk 2011, ASCE, Atlanta, GA, Jun. 26-28, 2011.

DNV/Risø 2002. Guidelines for the Design of Wind Turbines. 2nd

edition. Copenhagen. GL (Germanischer Lloyd) 2010. Guidelines for the Certification of

Wind Turbines. Edition 2010. Denholm P., Hand M., Jackson M. and Ong S. 2009. Land-Use

Requirements of Modern Wind Power Plants in the United States. National Renewable Energy Laboratory. Technical Report4NREL/TP-6A2-45834.

Baecher G.B. and Christian J.T. 2003 Reliability and Statistics in Geotechnical Engineering, John Wiley & Sons, New York.

Phoon K. K., Kulhawy H. 1999a. characterization of geotechnical variability. Canadian Geotechnical Journal 36, 615-624.Lee I.K., White W. and Ingles, O.G. 1983 Geotechnical Engineering, Pitman, Boston.

Fenton G.A. and Griffiths D.V. 2008. Risk Assessment in Geotechnical Engineering. John Wiley & Sons, New York.Phoon K.K., Kulhawy H. 1999. Evaluation of Geotechnical Property Variability. Canadian Geotechnical Journal 36, 625-639. Tippett, L.H.C. 1925. On the extreme individuals and the range of

samples taken from a normal population. Biometrika 17(3/4), 364-387.

Withiam, J.L., et al. 1997. Load and Resistance Factor Design (LRFD) for Highway Bridge Substructures, Federal Highway Administration, Washington, D.C.

Whitman, R.V. 2000. Organizing and evaluating uncertainty in geotechnical engineering, J. Geotech. Geoenvviron. Eng., ASCE126(7), 583-593.

Foye, K.C. Salgado, R. and Scott, B. 2006. Assessment of variable uncertainty for reliability-based design of foundations, J. Geotech. Geoenviron. Eng., ASCE 132(9), 1197-1207.

2323

Essais cycliques axiaux sur des pieux forés dans des sables denses

Cyclic axial load tests on bored piles in dense sands

Benzaria O. Fugro GeoConsulting, Nanterre, France – IFSTTAR, Paris, France

Puech A. Fugro GeoConsulting, Nanterre, France

Le Kouby A. IFSTTAR, Paris, France

RÉSUMÉ : Dans le cadre du projet national SOLCYP, cinq pieux forés instrumentés ont été installés dans les sables denses desFlandres. Deux pieux de 8 mètres de fiche et 420 mm de diamètre ont été soumis à des séries d’essais de chargements statiques etcycliques axiaux en compression. Cette communication présente les résultats les plus significatifs des essais statiques conventionnels et des essais cycliques en compression.

ABSTRACT: As part of the national project SOLCYP, five bored piles were installed in dense Flanders sands. Two 8m long, 420mmdiameter instrumented piles were submitted to extensive series of static and cyclic load tests in compression. This paper presents keyresults from conventional static tests and cyclic one-way tests in compression.

MOTS-CLÉS : SOLCYP, pieux forés, sables denses, chargements cycliques.

KEYWORDS : SOLCYP, bored piles, dense sands, cyclic loading.

1 INTRODUCTION

Le projet national SOLCYP (Puech et al., 2012) a pour objectif principal le développement d’une méthodologie pour le dimensionnement des pieux soumis à des chargements cycliques. Dans ce cadre, des essais sur pieux réels ont été conduits sur deux sites expérimentaux du Nord de la France: le site de Merville constitué d’argile des Flandres et le site de Loon-Plage constitué de sables denses. Sur le site de Loon-Plage ont été installés cinq pieux forés et deux pieux métalliques battus.

Les résultats obtenus sur les pieux battus et forés de Merville ont été partiellement publiés (Benzaria et al., 2012, 2013 ; Puech et Benzaria, 2013). Cette communication est centrée sur les résultats des essais statiques et cycliques en compression exécutés sur les pieux forés sur le site de Loon-Plage.

2 SABLE DES FLANDRES

Le site expérimental se situe sur la commune de Loon-Plage (59) près de Dunkerque dans le Nord de la France. Il se caractérise par une couverture de remblais récents (0-0,6m) et d’argile sableuse (0,6-2,2m) sous laquelle on rencontre la formation de sable des Flandres. La nappe phréatique au moment des essais se situait à environ 2m sous le niveau du terrain naturel.

Une campagne spécifique d'investigations a été réalisée au droit du plot d'essais comportant 4 essais au piézocône (CPTu), 2 essais au pressiomètre Ménard (PMT), 3 carottages continus et une série d'essais de laboratoire sur carottes (Figures 1 et 2). Le sable est un sable siliceux très fin (D50 voisin de 0,15mm) et mal gradué (coefficient d’uniformité CU=0,98). La formation est latéralement homogène et se caractérise par des valeurs de résistance au cône qn croissant de 5 à 40MPa vers 8 m de profondeur pour se stabiliser ensuite entre 30 et 50MPa jusque vers 11,5m. Entre 12 et 16,5m se trouve une couche d’argile molle qui fait brutalement chuter les caractéristiques mécaniques. L’interprétation des CPT par la méthode de Jamiolkowski et al. (2003) (Figure 2) conduit à un indice de densité ID compris entre 0,7 et 0,9 (sable dense à très dense).

Une série d’essais triaxiaux monotones a donné un angle de frottement interne Øcv voisin de 31° en bon accord avec les valeurs trouvées par Kuwano sur le sable de Dunkerque (Kuwano, 1999 ; Jardine et Standing, 2000)

Figure 1. Profils stratigraphiques et pressiométriques à Loon-Plage

3 INSTALLATION ET CHARGEMENT DES PIEUX

Les deux pieux, F4 et F5, sont géométriquement identiques (D=420mm, fiche 8m). Ils ont été exécutés par l’entreprise Botte Fondations à l’aide d’une tarière à axe creux vissée dans le sol sans extraction notable de matériau (Figure 3a) puis extraite sans dévissage tandis que le béton est injecté simultanément par l'axe creux. Les pieux sont équipés d’un train d’extensomètres amovibles de type LCPC introduits dans

2324


un tube de réservation positionné entre les armatures (Figure 3b).

Figure 2.Profils pénétrométriques à Loon-Plage : résistance au cône qc

et indice de densité ID.

Les pieux ont été testés trois mois environ après leur mise en place. Les programmes de chargement comportaient des essais statiques de référence à paliers d’une heure selon la norme NF P 94-150, des essais de chargement rapides (réduction des paliers à 3mn) et des essais de chargements cycliques axiaux de type répété. Une description plus précise des modes de chargement est indiquée dans Benzaria et al. (2012).

La caractérisation complète d’un chargement cyclique suppose la définition des paramètres suivants :

Qm: valeur moyenne de la charge sous chargement cyclique, Qc : demi-amplitude du chargement cyclique, Qmax: charge maximale (Qmax = Qm + Qc)N : nombre de cycles (les essais ont été conduits à la rupture

ou à grand nombre de cycles N>1000) f : fréquence des cycles (en général 0,5Hz) Qu : capacité statique ultime selon le mode considéré.

L’essai est dit répété si Qc< Qm et alterné si Qc>Qm.

Figure 3.a) réalisation d’un pieu foré à l’aide d’une tarière à axe creux b) schéma d’instrumentation d’un pieu foré à l’aide d’extensomètres amovibles de type LCPC

Le dispositif de chargement des pieux est similaire à celui utilisé pour les pieux de Merville et décrit dans Benzaria et al. (2012).

Figure 4. Dispositif d’essais de chargement utilisé à Loon-Plage

4 ESSAIS STATIQUES

La Figure 5 montre la courbe charge-déplacement en tête obtenue pour l’essai statique de référence sur le pieu F4, vierge de tout chargement, ainsi que la courbe de mobilisation de l’effort en pointe et la courbe de fluage (représentant la vitesse de déplacement du pieu lors de chaque palier de chargement).

On note : - une rupture de type ductile, - une entrée du pieu en grands déplacements pour un

déplacement en tête de l’ordre de 5% du diamètre. La charge de rupture Quc conventionnelle pour 10% de déplacement de la tête (42mm) peut être estimée à environ 1 100kN,

- une charge de fluage QF vers 850kN soit QF/Quc # 0,77, - une mobilisation de l’effort de pointe retardée puis quasi

linéaire jusqu’à 8% de déplacement relatif de la tête laissant préjuger d’une croissance de l’effort de pointe au-delà de la valeur conventionnelle à 0,1D.

Qu

= 11

00 k

N

à

Z0 =

42

mm

Figure 5. Essai statique de référence sur le pieu F4. Courbes de charge-déplacement en tête, de mobilisation de l’effort de pointe et de fluage.

a) b) Les mesures extensométriques ont permis de déterminer la distribution des charges le long du pieu (par éléments de 1m) ainsi que les courbes locales de mobilisation du frottement latéral selon la procédure décrite dans Benzaria et al. (2012). Les courbes locales de transfert de charges (dites aussi courbes t-z) sont données sur la Figure 6 pour les seuls niveaux correspondant au sable des Flandres. La mobilisation du frottement est très progressive : le déplacement local à la

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rupture est de l’ordre de 3 à 5% du diamètre du pieu. Le caractère ductile de la rupture est confirmé. Les frottements sont modestes (f < 45kPa).

Figure 6. Courbes de mobilisation des frottements locaux obtenues lors de l’essai statique de référence sur le pieu F4.

5 ESSAIS CYCLIQUES

5.1 Essais cycliques sur pieu vierge

La Figure 7 représente la séquence d’essais de chargements réalisés sur le pieu F5 vierge de tout chargement. Leurs caractéristiques sont données dans le Tableau 1

Tableau 1 : Caractéristiques des chargements appliqués au pieu F5 C : cyclique Quc : 1100 kN (pieu F1)

L’essai CC1 a été réalisé avec un taux de chargement relativement modeste (Qmax/Quc = 0,63). Il a été arrêté prématurément en raison de l’instabilité d’un pieu de réaction mais il est manifeste que la rupture était imminente après seulement 14 cycles. La rupture est définie pour un déplacement cyclique permanent de la tête du pieu de 3% du diamètre soit 12mm.

L’essai CC2, enchaîné avec un taux nettement plus faible (Qmax/Quc = 0,35), a pu être conduit à 5000 cycles mais pour un déplacement additionnel permanent de plus de 16mm.

L’essai CC3 (Qmax/Quc = 0,54) a provoqué un déplacement supplémentaire de 6mm pour seulement 280 cycles.

CC1

CC3

Figure 7. Courbes charges-déplacements en tête obtenues lors des essais de chargements réalisés sur le pieu F5.

Il est à noter que la vitesse d’évolution des déplacements en tête pour un nombre de cycles donné augmente avec le taux de chargement. Cependant pour chacun des essais cette vitesse décroît avec le nombre de cycles (Figure 8). Cette observation peut paraître contradictoire avec le constat de rupture.

CC2

Figure 8. Courbes déplacement en tête en fonction du nombre de cycles obtenues lors des essais de chargements réalisés sur le pieu F5.

Une observation fine des phénomènes générés par les chargements cycliques sur un pieu dans un sable permet de mieux comprendre cet apparent paradoxe. Les cycles provoquent une succession de petits glissements relatifs sol-pieu dont le cumul détermine le déplacement global. La vitesse de déplacement est fonction de l’amplitude et du niveau de chargement mais ces paramètres conditionnent également l’évolution du frottement qui peut se détériorer (cycles sévères) ou s’améliorer (petits cycles). Sur un pieu sollicité en traction, la vitesse initiale se modifie pour conduire vers la rupture de plus en plus rapide ou vers la stabilisation (Tsuha et al., 2012). La rupture peut être indifféremment définie de manière conventionnelle (par exemple 0,1D) ou en terme de déplacement acceptable. Sur un pieu en compression, le déplacement s’accompagne d’une mobilisation progressive de l’effort de pointe de sorte que dans tous les cas le pieu tend vers la stabilisation. Il en résulte que le critère de rupture doit être défini en terme de déplacement acceptable et non de manière conventionnelle. Le critère peut être alors atteint aussi bien en phase de vitesse de déplacement croissante que décroissante. Dans le premier cas, le nombre de cycles amenant à la rupture est faible. Dans le deuxième cas il peut être très élevé. On peut alors introduire la notion de zones instables et métastables (e.g. Puech et al, 2013).

TestF5-

Type Date f(Hz)

Qm /Quc

Qc /Quc

N

Installation 25/11/11

CC1 C 08/03/12 0,5 0,36 0,27 14

CC2 C 08/03/12 0,5 0,27 0,09 5000

CC3 C 08/03/12 0,5 0,36 0,18 280

CC1

CC3

CC2

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5.2 Effet de l’histoire des chargements

La Figure 9 montre l’histoire des chargements du pieu F4, détaillée dans le Tableau 2.

Sous chargement statique la rupture est ductile et la mobilisation du frottement nécessite des déplacements globaux et locaux importants entre 3 et 5% du diamètre du pieu.

L’essai CC1 a été exécuté après que le pieu ait été amené à la rupture (essai S1). Malgré son faible taux de chargement (Qmax/Quc = 0,31) le critère de rupture est atteint avant 2000 cycles. La capacité post cyclique (essai R1) ne semble toutefois pas affectée par cette série de cycles de faible amplitude.

Le comportement cyclique est très fortement dépendant de l’histoire des chargements. D’une manière générale, les pieux forés apparaissent très sensibles aux chargements répétés. Des déplacements importants peuvent être générés avant de pouvoir bénéficier d’une capacité en pointe suffisante pour stabiliser le pieu. La rupture doit donc être définie par un critère en déplacement et non de manière conventionnelle.

Une interprétation plus approfondie des résultats reste à proposer après traitement complet des autres données recueillies (essais alternés et essais en traction) et en liaison avec le comportement mécanique du sable des Flandres.

Tableau 2 : Caractéristiques des chargements appliqués au pieu F4 S : statique conventionnel R : statique rapide C : cyclique

TestF4-

Type Date2012

f(Hz)

Qm /Quc

Qc /Quc

N


CS1 S 01/03/12

CC1 C 02/03/12 0,5 0,18 0,13 1819

CR1 R 02/03/12

CC2 C 02/03/12 0,5 0,36 0,18 200

CC3 C 02/03/12 0,5 0,36 0,27 200

CC4 C 02/03/12 0,5 0,47 0,25 100

CC5 C 02/03/12 0,5 0,45 0,36 200

CR2 R 02/03/12

CR3 R 27/03/12

7 REMERCIEMENTS

Les résultats présentés dans cette communication ont été acquis dans le cadre du Projet National Français SOLCYP. SOLCYP est un projet de recherche sur le comportement des pieux soumis à des sollicitations cycliques, regroupant 12 entreprises et bureaux d’études du génie civil et 6 organismes universitaires et de recherche. Il est piloté par l’IREX et financé par les partenaires, l’Agence Nationale de la Recherche, le Ministère de l’Ecologie, du Développement Durable et de l’Energie et la Fédération Nationale des Travaux Publics. Les auteurs remercient les partenaires du projet d’avoir autorisé la publication de ces données.

8 REFERENCES

AFNOR.1999. NF P 94-150-2. Norme Française. Sols: Reconnaissance et Essais – Essai statique de pieu sous effort axial – Partie 1: en compression et Partie 2: en traction.

Une série de quatre essais d’amplitudes et taux moyens croissants (CC2 à CC5) est ensuite enchaînée, provoquant des déplacements significatifs (environ 25mm pour seulement 700 cycles cumulés). L’essai statique rapide R2 indique que la rigidité du pieu a fortement augmenté. L’essai statique R3 confirme ce résultat et montre que la capacité post cyclique du pieu est maintenant de 1480 kN soit un accroissement de 27% par rapport à l’essai S1 sur le pieu vierge.

CS1 CR1 CR2

CR3

CC1

CC2 à 5

Benzaria O., Puech A., and Le Kouby A. 2012. Cyclic axial load-tests on driven piles in overconsolidated clay, Offshore Site Investigation and Geotechnics, SUT, London

Benzaria A., Puech A. et Le Kouby A. 2013. Essais cycliques axiaux sur des pieux forés dans l’argile des Flandres. Proceedings 18th ICSMGE, Paris

Jamiolkowski, M.B., Lo Presti, D.C.F. & Manassero, M. 2003. Evaluation of Relative Density and Shear Strength of Sands from CPT and DMT. Soil Behavior and Soft Ground Construction, ASCE, GSP No. 119, 201-238.

Jardine, R.J., Standing, J.R.: OTO 2000. 008 - Pile Load Testing Performed for HSE Cyclic Loading Study at Dunkirk, France; Volume 1".

Kuwano R. 1999, "The stiffness and yielding anisotropy of sand." PhD thesis, University of London (Imperial College).

Puech A., Canou J., Bernardini C., Pecker A., Jardine R., and Holeyman A. 2012. SOLCYP: a four year JIP on the behavior of piles under cyclic loading. Offshore Site Investigation and Geotechnics, SUT, London

Puech A. et Benzaria O. 2013. Effet du mode de mise en place sur la réponse statique et cyclique de pieux dans l’argile surconsolidée des Flandres. Proceedings 18th ICSMGE, Paris

Puech A., Benzaria O., Thorel L., Garnier J., Foray P. et Jardine R. 2013. Diagrammes de stabilité cyclique de pieux dans les sables. Proceedings 18th ICSMGE, Paris

Tsuha C.H.C., Foray P.Y., Jardine R.J., Yang Z.X., Silva M., Rimoy S. 2012. Behaviour of displacement piles in sand under cyclic axial loading. Soils and Foundations 52(3), June 2012, 393–410,

Figure 9. Courbes charge-déplacement en tête obtenues lors des essaisde chargements réalisés sur le pieu F4.

6 CONCLUSIONS

On a présenté les résultats des essais statiques et cycliques en compression effectués sur les pieux forés à la tarière creuse installés sur le site de Loon-Plage constitué de sable dense des Flandres.

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Essais cycliques axiaux sur des pieux forés dans l’argile surconsolidée des Flandres

Cyclic axial load tests on bored piles in overconsolidated Flanders clay

Benzaria O. Fugro GeoConsulting, Nanterre, France - IFSTTAR, Paris, France

Puech A. Fugro GeoConsulting, Nanterre, France

Le Kouby A. IFSTTAR, Paris, France

RÉSUMÉ : Dans le cadre du projet national SOLCYP, quatre pieux forés instrumentés ont été installés dans l’argile fortementsurconsolidée des Flandres. Ces pieux de 13 mètres de fiche et 420mm de diamètre ont été soumis à des séries d’essais de chargements statiques et cycliques axiaux. Cette communication présente les principaux résultats des essais statiques conventionnelset des essais cycliques en compression répétés et alternés.

ABSTRACT: As part of the national project SOLCYP, four bored piles were installed in the high OCR Flanders clay. The 13m long,420mm diameter instrumented piles were submitted to extensive series of static and cyclic load tests. This paper presents key resultsfrom conventional static tests and cyclic one-way and two-way tests in compression.

MOTS-CLES: SOLCYP, pieux forés, argile surconsolidée, chargements cycliques.

KEYWORDS : SOLCYP, bored piles, overconsolidated clay, cyclic loading

1 INTRODUCTION

Le projet national SOLCYP (Puech et al., 2012) a pour objectif principal le développement d’une méthodologie pour le dimensionnement des pieux soumis à des chargements cycliques. Dans ce cadre, des essais sur pieux réels ont été conduits sur deux sites expérimentaux du Nord de la France: le site de Merville constitué d’argile des Flandres et le site de Loon-Plage constitué de sable dense. Sur le site de Merville ont été installés quatre pieux métalliques battus, quatre pieux forés et deux pieux vissés moulés.

Un précédent article (Benzaria et al., 2012) était consacré à la description du contexte expérimental et à la présentation des résultats obtenus sur les pieux battus. Cette communication présente les résultats des essais statiques et cycliques exécutés sur les pieux forés.

2 ARGILE DES FLANDRES

Le site expérimental se situe sur la commune de Merville (59) dans le Nord de la France. Il se caractérise par une couverture de limons sableux à argileux de 3,5m d'épaisseur dans lequel fluctue la nappe phréatique (-2m environ lors des essais) sous laquelle on rencontre la formation d'argile des Flandres, particulièrement homogène sur toute la zone, et d'une puissance de 40m environ.

L'argile des Flandres, géologiquement comparable à l'argile de Londres et à l'argile de Boom, s’est déposée il y a 50 millions d’années (Eocène) dans un golfe marin qui couvrait toute la zone Nord de la France, de la Belgique et du Sud Est de l’Angleterre. Elle a été recouverte par des formations tertiaires dont la sédimentation s’est poursuivie jusqu’au Pleistocène supérieur. Le niveau du sol se situait alors probablement à 200m au-dessus du niveau actuel. Les formations sus-jacentes se sont érodées. Le processus d’érosion a été suivi au Quaternaire par le dépôt d’alluvions du Flandrien. Le matériau a été soumis a des cycles de chargement/déchargement et à des processus périglaciaires qui associés à des phénomènes de cimentation

chimique et de vieillissement ont fortement conditionné son degré de surconsolidation (OCR) apparent (Josseaume, 1998)

L’argile des Flandres présente des caractéristiques voisines de celles des argiles de Londres et de Boom (Borel et Reiffsteck, 2005) :

- faible teneur en eau (de l’ordre de 30%) - forte plasticité (IP voisin de 50) - forte fissuration notamment au-delà de 5m de profondeur

Figure 1. Profils de résistance au cône qt et OCR sur le site de Merville

Une campagne spécifique d'investigations a été réalisée au

droit du plot d'essais comportant des essais au piezocône (CPTU), des essais au pressiomètre Ménard (PMT), des carottages continus et une série d'essais de laboratoire sur carottes.

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L’OCR a été estimé à partir du CPT et de la relation de Mayne (1991) : OCR = k. (qt-σv0)/σ’v0 avec k=0,5. Le facteur k est fonction du type de matériau. La valeur de 0,5 a été retenue car elle donne des valeurs d’OCR compatibles avec l’épaisseur de surcharge supposée et le gradient de qn = qt-σv0 dans l’argile profonde (au-delà de 8m). A noter que pour l’argile de Londres, Powell et al., 1989 suggèrent des valeurs de k supérieures à 1.

Figure 2. Conditions de sol au niveau du plot d’essais de Merville Les essais triaxiaux de type UU (non consolidé, non drainé) ou CIU (consolidé isotropiquement, non drainé) montrent des ruptures prématurées de type fragile, caractéristiques de ce type d’argile plastique fissurée et fortement surconsolidée. La rupture se caractérise par la formation de plans de cisaillement contenant des particules réorientées, comme noté par Bond et Jardine (1991). Les valeurs de la résistance au cisaillement non draînée Cu sont corrélées à la résistance au cône par un facteur Nkt élévé, [Nkt= (qt – σvo) /Cu =20], compatible avec la nature du matériau. Les valeurs de pression limite pressiométrique nette sont assez bien reliées à Cu par la relation d’Amar et Jezequel (1998) : Cu = pl*/12+30, avec Cu en kPa et pl* en MPa

3 INSTALLATION ET CHARGEMENT DES PIEUX

Les quatre pieux sont géométriquement identiques (D=420mm, fiche 13m). Ils ont été exécutés à l’aide d’une tarière à axe creux vissée dans le sol sans extraction notable de matériau (Figure 3a) puis extraite sans dévissage tandis que le béton est injecté simultanément par l'axe creux. La partie basse est munie d'un manchon télescopique. Les pieux sont équipés d’un train d’extensomètres amovibles de type LCPC introduits dans un tube de réservation positionné entre les armatures (Figure 3b).

Les pieux ont été testés deux mois après leur mise en place. Les programmes de chargement comportaient des essais statiques de référence à paliers d’une heure selon la norme NF P 94-150 (1999), des essais de chargement rapides (réduction des paliers à 3mn) et des essais de chargement cycliques axiaux de type répété ou alterné. Une description plus précise des dispositifs (Figure 4) et modes de chargement est indiquée dans (Benzaria et al., 2012).

La caractérisation complète d’un chargement cyclique suppose la définition des paramètres suivants :

Qm: valeur moyenne de la charge sous chargement cyclique, Qc : demi-amplitude du chargement cyclique, Qmax: charge maximale (Qmax = Qm + Qc) N : nombre de cycles (les essais ont été conduits à la rupture

ou à grand nombre de cycles N>1000)

Figure 3. Pieux forés a) tarière à axe creux. b) schéma d’instrumentation à l’aide d’extensomètres amovibles de type LCPC.

b)

a)

Figure 4 : Dispositif de chargement des pieux en compression.

f : fréquence des cycles (en général 0,5Hz) Qu : capacité statique ultime selon le mode considéré. L’essai est dit répété si Qc< Qm et alterné si Qc>Qm.

4 ESSAIS STATIQUES

La Figure 5 montre la courbe charge-déplacement en tête obtenue pour l’essai statique de référence sur le pieu F1, vierge de tout chargement, ainsi que la courbe de fluage représentant la vitesse de déplacement du pieu lors de chaque palier de chargement.

Qu =

900

kN

à

Z0 =

42

mm

Figure 5. Courbe charge-déplacement en tête et courbe de fluage obtenues pour l’essai statique de référence sur le pieu F1.

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On note : - une rupture de type ductile (contrairement aux observations

de Benzaria et al., 2012 sur le pieu battu) ; - une entrée du pieu en grands déplacements pour un

déplacement en tête de l’ordre de 3% du diamètre. La charge de rupture Quc conventionnelle pour 10% de déplacement de la tête peut être extrapolée avec confiance à 900kN ;

- une charge de fluage QF nette vers 670kN soit QF/Quc # 0,75.

Figure 6. Distributions des charges obtenues pour l’essai statique de référence sur le pieu F1.

Figure 7. Courbes locales de mobilisation du frottement latéral obtenues pour l’essai statique de référence sur le pieu F1.

Les mesures extensométriques ont permis de déterminer la distribution des charges le long du pieu (Figure 6) ainsi que les courbes locales de mobilisation du frottement latéral selon la procédure décrite dans Benzaria et al. (2012). Les courbes locales de transfert de charges (dites aussi courbes t-z) sont données sur la Figure 7. La mobilisation du frottement est très rapide (entre 1 et 2%) du diamètre du pieu. Le caractère ductile de la rupture est confirmé. Les frottements sont modestes (f <50kPa) et très inférieurs à ceux observés sur les pieux métalliques battus installés sur ce même site. 5 ESSAIS CYCLIQUES

5.1 Essais cycliques répétés

La Figure 8 représente la séquence d’essais de chargements réalisés sur le pieu F2. Leurs caractéristiques sont données dans la Tableau 1.

Figure 8. Courbes charge-déplacement en tête obtenues lors d’essais de chargements statiques et cycliques sur le pieu F2.

On observe que : - le taux de chargement maximal contrôle la réponse du pieu ; - lorsque ce taux demeure inférieur à un seuil critique, les

déplacements restent non significatifs y compris pour de grands nombres de cycles (N>1000). Les boucles d’hystérésis sont fermées (Figure 9) ;

- lorsque le seuil critique est franchi, le pieu cumule rapidement des déplacements le conduisant vers la rupture Les essais CC3, CC8, CC9 et CC10 ont dépassé 12mm (3%D) de déplacement cyclique permanent considéré comme critère de rupture (Figure 8) ;

- le seuil critique est très voisin de Qmax/Quc =0,9 - les chargements cycliques, y compris lorsqu’ils conduisent à

la rupture, n’ont pas significativement affecté la capacité statique du pieu.

Tableau 1 : Caractéristiques des chargements appliqués au pieu F2

C : cyclique R : statique rapide Quc = 900kN (pieu F1)

TestF2-

Type Date f(Hz)

Qm /Quc

Qc /Quc

N


CC1 C 16/05/2011 0,5 0,50 0,25 3408

CC2 C 17/05/2011 0,5 0,58 0,25 4834

CC3 C 17/05/2011 0,5 0,58 0,33 2021

CR1 R 17/05/2011

CC4 C 17/05/2011 0,5 0,25 0,20 1013

CC5 C 17/05/2011 0,5 0,40 0,20 1000

CC6 C 17/05/2011 0,5 0,40 0,30 1088

CC7 C 17/05/2011 0,5 0,50 0,30 602

CC8 C 17/05/2011 0,5 0,50 0,40 81

CC9 C 17/05/2011 0,1 0,50 0,40 24

CC10 C 17/05/2011 0,5 0,50 0,40 85

CR2 R 17/05/2011

CC1 à 3 CR1 CC4 à 10 CR2

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La comparaison des essais CC9 et CC10 effectués à mêmes taux de chargement met en outre en évidence un effet de fréquence dans le domaine proche de la rupture : la rupture est accélérée lorsque la fréquence diminue.

Figure 9. Courbes charge-déplacement en tête obtenues d’essais de chargements réalisés sur le pieu F2.

5.2 Essais cycliques alternés Neuf essais cycliques alternés ont été effectués sur le pieu F3. Les caractéristiques de ces essais sont telles que Qm/Quc< 0,2 et 0,2<Qc/Quc<0,5. Pour des raisons de montage hydraulique (deux vérins en opposition de phase), il n’a pas été possible d’appliquer une charge maximale Qmax supérieure à 0,7 Quc.

Tous les essais effectués dans ces conditions sont du type stable avec des boucles d’hystérésis fermées comme illustré sur la Figure 10.

Figure 10, Courbe charge-déplacement en tête obtenue lors de l’essai de chargement CC9 réalisé sur le pieu F3. 5.3 Notion de stabilité cyclique L’ensemble des résultats obtenus en compression pure et en compression alternée (mais également en traction pure et en traction alternée) suggèrent que pour le type de matériau et de pieu considéré il existe deux zones de fonctionnement bien différenciées : - une zone dans laquelle les chargements cycliques même en

grand nombre (N>1000) ont peu d’effet sur le comportement du pieu : accumulation non significative des déplacements permanents, rigidité cyclique constante ;

- une zone dans laquelle le pieu évolue très rapidement vers la rupture.

La zone de stabilité cyclique est étendue mais pourrait être affectée par le caractère répété ou alterné des chargements.

L’analyse complète des résultats et la définition de critères de rupture précis devrait permettre de proposer ultérieurement des diagrammes de stabilité cyclique au sens de Poulos (1988) ou Karlsrud et al. (1986)

6 CONCLUSIONS

CC9 CC10 On a présenté des résultats d’essais statiques et cycliques en compression pure et alternée effectués sur des pieux forés à la tarière creuse installés sur le site de Merville constitué d’argile surconsolidée des Flandres.

CC6

Sous chargement statique la rupture n’est pas fragile contrairement aux observations faites sur les pieux battus. Les frottements mobilisés sont nettement inférieurs (Benzaria et al., 2012).

Il semble exister sous chargement cyclique une zone étendue à l’intérieur de laquelle la stabilité est assurée pour un grand nombre de cycles. Le seuil critique, au moins pour les essais répétés, semble voisin de Qmax/Quc =0.9. Les chargements cycliques, y compris lorsqu’ils conduisent à la rupture, n’affectent pas significativement la capacité statique du pieu.

Un article à cette même conférence (Puech et Benzaria, 2013) analyse le comportement statique des deux types de pieux battus et forés en liaison avec la nature et le comportement mécanique de l’argile des Flandres.

7 REMERCIEMENTS

Les résultats présentés dans cette communication ont été acquis dans le cadre du Projet National Français SOLCYP. SOLCYP est un projet de recherche sur le comportement des pieux soumis à des sollicitations cycliques, regroupant 12 entreprises et bureaux d’études du génie civil et 6 organismes universitaires et de recherche. Il est piloté par l’IREX et financé par les partenaires, l’Agence Nationale de la Recherche, le Ministère de l’Ecologie, du Développement Durable et de l’Energie et la Fédération Nationale des Travaux Publics. Les auteurs remercient les partenaires du projet d’avoir autorisé la publication de ces données.

8 REFERENCES

AFNOR.1999. NF P 94-150-2. Norme Française. Sols: Reconnaissance et Essais – Essai statique de pieu sous effort axial – Partie 1: en compression et Partie 2: en traction.

Benzaria O., Puech A and Le Kouby A. 2012. Cyclic axial load-tests on driven and bored piles in overconsolidated clay, Offshore Site Investigation and Geotechnics, SUT, London.

Bond A.J. and Jardine R.J. 1991. Effects of installing displacement piles in a high OCR clay. Géotechnique, 41(3) 341-363.

Borel S. and Reiffsteck P. 2006. Caractérisation de la déformabilité des sols au moyen d’essais en place. Géotechnique et Risques Naturels, LCPC, GT81.

Josseaume H. 1998. Propriétés mécaniques de l’argile des Flandres à Dunkerque et Calais. Revue Française de Géotechnique, N°84.

Karlsrud, K., Nadim F. and T. Haugen (1986). Piles in clay under cyclic loading: Field tests and computational modeling. Proc., 3rd int. Conf. on Nun. Meth. In offshore Piling, 165-190, Nantes, France, May 1986.

Mayne P.W. 1986. CPT indexing of in situ OCR in clays. Proceedings ASCE Spec. Conf. “In situ 86”, Blacksburg, Virginia.

Powell J.J.M., Quaterman R.S.T. and Lunne T. 1989. Interpretation and use of piezocone test in UK. Proc. Geotechnology Conference: penetration testing in UK, Birmingham, Thomas Telford, London

Puech A., Canou J., Bernardini C., Pecker A., Jardine R., and Holeyman A. 2012. SOLCYP: a four year JIP on the behavior of piles under cyclic loading. Offshore Site Investigation and Geotechnics, SUT, London.

Puech A. et Benzaria O. 2013. Effet du mode de mise en place sur la réponse statique de pieux dans l’argile surconsolidée des Flandres. Proceedings 18th ICSMGE, Paris.

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Fondations superficielles glissantes pour l’offshore profond – Méthodologie de dimensionnement

Deep Offshore Sliding Footings – Design Methodology

Bretelle S. GHD, Perth, Australie

Wallerand R. Subsea7, Paris, France

RÉSUMÉ : Les caractéristiques géotechniques des sols rencontrés en offshore profond conduisent de manière récurrente à desfondations de très grandes dimensions. Ceci induit des difficultés d’installation qui stimulent la recherche de solutions audacieusespour les fondations dont le dimensionnement intègre des sollicitations horizontales transitoires. Les efforts horizontaux résultent deconditions transitoires pour les pipelines (arrêt et redémarrage de l’exploitation induisant des changements de température dans lespipelines). Ceux-ci induisent un glissement de la fondation. Ces efforts horizontaux étant variables dans le temps, le sol est soumis àdes sollicitations cycliques. Cet article détaille les différentes étapes proposées pour le dimensionnement des fondations glissantes,destinées au support des structures liées aux pipelines, en prenant en compte les effets de la dégradation cyclique sur la capacitéportante, et l’évaluation des tassements.

ABSTRACT: Due to the geotechnical characteristics of offshore deep-water soils, footing size can become very large. This may induce installation challenges, which in turn stimulates smart design solutions allowing for horizontal displacement of the footing.Horizontal loads results from flow lines expansion during start-up and shut-down of the production, as large temperature changes will induce dilatation. If horizontal loads are above the horizontal capacity of the footing, the foundation will slide. Horizontal loads varying with time, soil cyclic loadings are induced. This paper presents the different steps of the design for the proposed concept, allowing the footing to slide. This design includes cyclic soil degradation effects on bearing capacity and settlements. MOTS-CLÉS: Offshore, fondation, glissante, pipeline, cyclique, dégradation, tassement. KEYWORDS: Offshore, footing, sliding, pipeline, cyclic, degradation, settlement.

1 INTRODUCTION

Les caractéristiques géotechniques des argiles molles rencontrées en offshore profond conduisent de manière récurrente à des fondations de très grandes dimensions. Ceci induit des difficultés d’installation qui stimulent la recherche de solutions audacieuses pour les fondations dont le dimensionnement intègre des sollicitations horizontales transitoires.

Les efforts horizontaux résultent des sollicitations horizontales transitoires. C'est le cas notamment des structures intermédiaires ou des structures fin de ligne, qui permettent la connexion entre les pipelines et les différents éléments de l'architecture du champ (puits, risers, manifolds). Les efforts appliqués à ces structures sont transmis au sol par l'intermédiaire de fondations superficielles faites de tôles d'acier. Les efforts verticaux proviennent du poids propre de la structure, de la section de conduite qu'elle supporte, des différents connecteurs et vannes associées. Les efforts horizontaux sont eux provoqués par la dilatation ou la contraction de la conduite (arrêt et redémarrage de l'exploitation induisant des changements de température et de pression dans les pipelines). Ceux-ci induisent un glissement de la fondation (le coefficient de sécurité au glissement est inférieur à 1). Ces efforts horizontaux étant variables dans le temps, le sol est soumis à des sollicitations cycliques.

L’évaluation des effets résultants de la dégradation cyclique des propriétés des sols (ici la résistance au cisaillement) permettra d’assurer que la fondation reste stable sous efforts verticaux et moments.

L’évaluation des tassements est nécessaire. Le cisaillement cyclique (environ 2000 cycles sur la durée de vie de l’ouvrage)

est source de tassements spécifiques, qui sont dus aux déplacements répétés de la fondation.

Cet article détaille les différentes étapes proposées pour le dimensionnement des fondations glissantes, destinées au support des structures liées aux pipelines, et posées sur des argiles molles.

2 COMPARAISON PIPELINES-FONDATIONS

Le tableau 1 rappelle les concepts d’interaction sol-pipeline et sol-fondation. Il permet d’illustrer les différences entre les comportements admis pour les pipelines et ceux généralement requis pour les fondations superficielles (White - Cathie 2010).

Les grandes familles de différences qui sont décrites ci-dessous affectent directement la perception du dimensionnement, qui intègre naturellement les déplacements pour les pipelines, et qui par contre recommande des coefficients de sécurité importants sur la capacité portante (incluant les efforts horizontaux) des fondations pour se prémunir des déplacements.

Pour ce qui concerne les pipelines, il peut s’avérer économiquement impossible d’empêcher le pipeline de se déplacer, en l’enfouissant par exemple, sur des dizaines de kilomètres, et à des profondeurs d’eau de plus de 1000m. Par contre, des modèles d’interaction sol-pipeline intégrant le déplacement permettent d’optimiser les dimensions des pipelines (les épaisseurs des tubes en particulier).

Pour ce qui concerne les fondations, même liées aux pipelines, il reste d’usage de les dimensionner sans autoriser leur déplacement.

Ce choix conduit à des fondations de très grandes dimensions, en particulier sur les argiles marines en grande

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profondeur. Il pose aussi des problèmes structuraux au niveau de la connexion du pipeline et de la fondation.

Tableau 1. Comparaison des comportements admis pour les

ipelines et pour les fondations. p Paramètre Pipeline Fondation

Géométrie de l’interface sol-

fondation

Incertaine, la pénétration dépend des

méthodes d’installation, de la forme du fond marin

Connue, contrôlée

Critère de dimensionnement

en service

Déplacement autorisé, lois de comportement

spécifiques

Fixe, déplacements minimes de la

fondation

Sol Perturbé pendant l’installation et les cycles

Similaire à l’état initial

Charges La réponse du pipeline

est affectée par l’interaction avec le sol

Les charges ne sont pas affectées

par le sol de fondation

Méthodes de dimensionnement

Non disponible (ou encore contestées), les

hypothèses de sol enveloppes (hautes ou

basses) doivent être considérées

Disponibles, les hypothèses de

sol basses sont conservatives pour

la stabilité

3 DIMENSIONNEMENT SOUS CHARGES PERMANENTES

La première étape traite du dimensionnement sous charges permanentes qui requiert un coefficient de sécurité supérieur à 1 avec la prise en compte concomitante des efforts verticaux et moments (Cathie 2008).

La stabilité au sens usuel du terme doit être assurée. Les

charges permanentes sont essentiellement liées au poids de la structure.

La problématique du dimensionnement des fondations superficielles connait de nombreux développements récents et il existe une normalisation (ISO 19901-4 2011) afférente. Celle-ci permet la prise en compte de manière concomitante des efforts verticaux N, horizontaux T ainsi que des moments M.

On rappelle en préambule les principes du dimensionnement intégrant ces efforts M, N, T, sous la forme d’enveloppes 3D (Randolph 2005).

Les mécanismes de rupture sont rappelés sur la figure 1.

L’approche traditionnelle de dimensionnement de fondations superficielles consiste à simplifier le chargement (M, N, T) en un chargement (N, T) appliqué sur une surface réduite. Cette approche conduit à des surdimensionnements quand ils sont comparés à des calculs par éléments finis tridimensionnels.

L’approche proposée dans l’API-ISO permet l’utilisation d’enveloppes de rupture 3D, utilisant les valeurs de : Nult charge verticale de rupture sous chargement

vertical seul (M=0 et T=0) ; Tult charge horizontale de rupture sous chargement

horizontal seul (M=0 et N=0) ; Mult moment de renversement ultime en condition de

renversement seul (T=0 et V=0). La surface limite dans l’espace (N/Nult, M/Mult, T/Tult) est

alors définie par des calculs spécifiques 3D ou disponibles dans certaines références pour des formes de fondation types. Un exemple est proposé sur la figure 2.

Figure 2 : Enveloppe de rupture (Gourvenec 2007)

4 DIMENSIONNEMENT SOUS CHARGES HORIZONTALES effort vertical NLes efforts horizontaux résultent de conditions transitoires pour les pipelines (arrêt et redémarrage de l’exploitation induisant des changements de température dans les pipelines).

Un coefficient de sécurité suffisant doit être obtenu pour que la fondation soit stable (pas de rupture sous charges verticales et moments), tout en autorisant son déplacement horizontal.

effort horizontal

L’évaluation présentée dans cet article se limite au cas des argiles molles trouvées dans les grands fonds.

4.1 Vérifications de stabilité

L’évaluation des effets résultants de la dégradation cyclique des propriétés des sols (ici la résistance au cisaillement) se décompose en trois étapes :

Moment M

Une évaluation des contraintes de cisaillement est effectuée sous la fondation le long de surfaces de rupture ; Figure 1 : Mécanismes de rupture

l’effet des cycles est intégré en utilisant des résultats usuels d’essais de laboratoire caractérisant la dégradation cyclique ;

une vérification de la stabilité de la fondation est effectuée le long des surfaces de rupture précédentes, avec les caractéristiques réduites déterminées précédemment.

4.1.1 Évaluation des contraintes de cisaillement Il est suggéré d’utiliser un programme de calcul aux éléments finis pour obtenir les contraintes de cisaillement sous la fondation.

Il n’est pas requis d’utiliser un modèle autorisant des grands déplacements (2 à 3 mètres sont observés pour les pipelines), ni

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de représenter fidèlement les cycles (les structures concernées subissent 1000 à 5000 cycles), ce qui nécessiterait des temps de calculs importants et des outils complexes.

La modélisation proposée ici applique un effort vertical représentatif du poids de la structure, et un effort horizontal « à la rupture », c’est-à-dire saturant les contraintes de cisaillement à l’interface sol fondation.

La figure 3 présente un modèle type de calcul.

Figure 3 : Modèle de calcul Ce calcul permet de déterminer l’étendue des zones

sollicitées sous la fondation et les contraintes de cisaillement appliquées au sol de fondation.

A partir de ce calcul, on détermine le taux de cisaillement moyen (à différentes profondeurs sous la fondation (en général de 0 à B, où B est la largeur de la fondation).

4.1.2 Dégradation cyclique La dégradation cyclique est déterminée à partir d’essais de laboratoire cycliques, qui permettent d’obtenir des enveloppes de dégradation en fonction du nombre de cycles.

Figure 4 : Dégradation cyclique (Argile de Drammen, OCR=1,

d’après Andersen 2004) On lit sur l’axe horizontal le rapport /su et sur l’axe

vertical le ratio cy/su. Le nombre de cycles correspond aux différentes courbes sur l’abaque.

Les notations suivantes sont utilisées : τ0 = cisaillement moyen sans cycles τcy = cisaillement dégradé, après N cycles su = résistance au cisaillement moyen

4.1.3 Vérification de la stabilité après dégradation cyclique Le calcul présenté en 4.1.1 peut être repris en modifiant les valeurs de Su pour tenir compte de la dégradation cyclique.

Le coefficient de sécurité obtenu est alors représentatif de la stabilité de la fondation après un grand nombre de cycles.

4.2 Calculs des tassements

L’évaluation des tassements (élastique, consolidation, fluage) est nécessaire. Le cisaillement cyclique (environ 2000 cycles sur la durée de vie de l’ouvrage) est source de tassements additionnels, qui sont dus aux déplacements répétés de la fondation. Les tassements additionnels suivants sont considérés : Tassements induits par le chargement cyclique de

l’argile profonde (données de triaxiaux DSS publiées), Dégradation de la couche en contact avec la fondation

du fait du glissement répété, Erosion par les bords de la fondation (cause potentielle

de réduction de la surface effective sous la fondation).

4.2.1 Tassements élastiques, consolidation, fluage Les tassements élastiques de consolidation et de fluage peuvent être déterminés à partir des méthodes usuelles analytiques, ou à partir du modèle éléments finis décrit plus haut (avec des modèles de type Soft Soil pour Plaxis).

4.2.2 Réduction cyclique du module d’élasticitéPour les tassements cycliques, il faudra tenir compte : Des déformations dues aux cycles ; Du développement de surpressions interstitielles dû aux

cycles. Dans les deux cas, on utilisera les résultats des calculs

décrits plus haut, et les enveloppes obtenues par essais de laboratoire.

Les notations suivantes sont utilisées (différentes de celles du chapitre précédent) :

τav = cisaillement moyen sans efforts horizontaux τcy = cisaillement moyen avec efforts horizontaux su = résistance au cisaillement moyen La figure 5 présente des résultats types pour la

détermination des déformations () dues au cisaillement cyclique.

Figure 5 : Déformations moyennes de cisaillement cyclique (Argile de Drammen, OCR=1, d’après Andersen 2004)

La déformation ci-dessus est multipliée par l’épaisseur de la

couche concernée pour obtenir le tassement supplémentaire. La figure 6 présente l’augmentation de pression interstitielle

en fonction du nombre de cycles. La déformation supplémentaire est alors obtenue par : = (u / ’v)*’v / M Avec u / ’v : lu sur la figure 6 ’ : contrainte effective verticale vM : module de déformation

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Figure 6 : augmentation de pression interstitielle due au cisaillement de 1000 cycles (Argile de Drammen, OCR=1, d’après Andersen 2004)

Figure 7 : Exemple de calcul avec dégradation sous les bords (cerclées).

5 CONCLUSION 4.2.3 Dégradation de la couche de contact

La couche de contact (les 20 à 30 cm d’argile directement en contact avec la face inférieure de la fondation) subit de façon répétée un état de rupture, sous l’effet du glissement.

Cet état de rupture reste en général confiné au niveau de cette couche d’interface, car une couche de meilleures caractéristiques (croute) est souvent présente entre 0 et 2 m de profondeur. La croute agit comme une barrière de protection qui limite le transfert des ruptures par cisaillement vers les couches inférieures.

Cet article a détaillé les différentes étapes proposées pour le dimensionnement des fondations glissantes, destinées au support des structures liées aux pipelines, dans des conditions géotechniques d’argiles grands fonds.

Il est très difficile, voire impossible de modéliser le processus cyclique dans cette couche supérieure, du fait des redistributions de contraintes entre les bords et le centre de la fondation et bien entendu du fait des sollicitations cycliques.

Le tassement à long terme et sous un nombre de cycle représentatif de la vie de la structure sera obtenu par cumul des différentes valeurs de tassements calculées suivant la démarche proposée. La stabilité de la structure sera ainsi assurée, et son fonctionnement ne sera pas altéré par les tassements ou rotations dus au chargement cyclique.

6 REMERCIEMENTS

Il est donc proposé de retenir une approche simplifiée qui considère les différents aspects de la dégradation.

Le modèle éléments finis décrit plus haut permet de considérer le moment maximum qui sollicite la fondation, associé au poids de la fondation et à l’effort horizontal qui sature le frottement au niveau de l’interface. Ce calcul peut être mené avec les caractéristiques « intactes » dans la couche de contact, puis avec les caractéristiques remaniées dans cette même couche.

Je tiens à remercier mon mari et mon fils pour leur patience pendant la période de rédaction de cet article, ainsi que pour leur relecture.

La comparaison directe des résultats permet d’estimer un tassement et une rotation supplémentaire à considérer dans le dimensionnement.

Bien entendu, cet article n’aurait pas vu le jour sans mon expérience en France, à Terrasol pour la maitrise des fondamentaux géotechniques, chez Saipem, pour l’ouverture Internationale et offshore, chez Cathie Associates pour la pratique continue des projets lies aux grands fonds et enfin en Australie, où GHD me donne l’opportunité de travailler sur des grands projets d’infrastructure.

7 REFERENCES 4.2.4 Erosion par les bords La concentration de contraintes est un phénomène bien connu qui dans ce cas particulier peut créer des tassements supplémentaires par érosion près des angles.

Les skis ont depuis longtemps apporté une réponse appropriée en relevant les bords pour limiter l’effet de l’angle.

Andersen 2004 Cyclic clay data for foundation design of structures subjected to wave loading International Conf. on “Cyclic behaviour of Soils and Liquefaction Phenomena” Keynote lecture CBS04, Bochum, Germany.

Cathie et Al. 2005 Pipeline Geotechnics State of the Art.

Les fondations glissantes devront comporter un dispositif similaire pour éviter de racler progressivement toute une bande de sol.

Cathie et Al 2008 Design of sliding foundations for subsea structures. Gourvenec, S. (2007), Shape Effects on Capacity of Rectangular

Footings Under Combined Load, Géotechnique, 57(8), pp. 637–646.

L’estimation par le calcul de ce phénomène reste à développer, par contre, les conséquences peuvent être évaluées en retirant arbitrairement les éléments de sol de l’interface sous une partie de la fondation. On pourra enlever les éléments sur une bande correspondant au mouvement calculé des pipelines qui se connectent sur la fondation, puisque ce mode de comportement est disponible dans les modèles de pipelines.

ISO 19901-4 2003 modified 2011 API RP2GEO Geotechnical and foundation design considerations.

Randolph et Al - 2005 Challenges of offshore geotechnical engineering. White, D., Cathie, D.N. 2011, 'Geotechnics for subsea pipelines',

Frontiers in Offshore Geotechnics II, The Netherlands, CD, pp. 87-123.

Une autre méthode, moins pénalisante consiste à reprendre les résultats du calcul avec les moments présentés au paragraphe précédent 4.2.3, pour déterminer les contraintes dans la couche d’interface.

En comparant ces contraintes à la contrainte ultime (Nc*Su) ou en utilisant les zones plastifiées du modèle éléments finis, l’étendue de la zone remaniée (sa largeur sous la fondation) peut être estimée. On peut alors reprendre le calcul précédent en enlevant une zone plus limitée. La figure 7 illustre un exemple de calcul.

2335

Proposition d’une loi t-z cyclique au moyen d’expérimentations en centrifugeuse

Proposal of cyclic t-z law by means of centrifuge experiments

Burlon S., Thorel L. Université Paris-Est, IFSTTAR, Département GERS, Paris, France Mroueh H. LGCgE, Lille, France

RÉSUMÉ : Cet article propose une extension de la loi t-z de Frank et Zhao (1982) pour le calcul des déplacements d’un pieu soumis àdes charges axiales cycliques. Les potentialités de cette nouvelle loi sont présentées et une comparaison est effectuée entre des résultats expérimentaux issus d’essais en centrifugeuse pour quatre pieux soumis à des charges cycliques et des résultats numériquesobtenus au moyen de cette nouvelle loi. Pour chaque cas, le déplacement de la tête de pieu en fonction du nombre de cycles de chargement est analysé. Une discussion est menée pour mieux cerner l’influence de chaque paramètre de la nouvelle loi t-z cyclique développée et envisager les développements futurs à mettre en œuvre.

ABSTRACT: This paper includes an extension of the t-z law proposed by Frank and Zhao (1982) for the displacements calculation ofa pile subjected to cyclic axial loads. The potential of this new law are presented and a comparison is made between experimental results obtained from centrifuge tests for four piles subjected to cyclic loads and numerical results obtained by this new law. For each case, the displacement of the pile head according to the number of load cycles is analyzed. A discussion is conducted to better understand the influence of each parameter of the new cyclic t-z law and to consider future developments to implement.

MOTS-CLÉS : Fondation profonde, charge axiale, sollicitations cycliques, centrifugeuse, loi t-z.

KEYWORDS: Deep foundations, axial load, cyclic load, centrifuge tests t-z curve.

1 INTRODUCTION

Le comportement des pieux soumis à des charges axiales monotones peut être appréhendé par des calculs mettant en œuvre des lois d’interaction locale de type t-z. Ces lois permettent, pour chaque section d’un pieu, d’associer le déplacement relatif de l’interface sol-pieu à la contrainte de cisaillement mobilisée. Cette approche de calcul du comportement des pieux est généralement mise en œuvre,

e 7

sont très ent sur les

ur les sols fins e charges

ge du pieu et

cliques (fondations d’éoliennes, de structures pétrolières, etc.). Toutefois, des lois t-z cycliques, développées notamment pour l’ingénierie pétrolière, existent pour rendre compte des effets de tels chargements (Chin et Poulos 1992 et Randolph 1986).

Sur la base de ces lois, cet article propose une extension de la loi t-z formulée par Frank et Zhao. Les aptitudes de cette nouvelle loi, pour rendre compte des phénomènes de durcissement ou de radoucissement cyclique ou de rochet et de relaxation (Lemaître et Chaboche 2009), sont présentées à l’échelle locale. Une comparaison entre des résultats expérimentaux issus d’essais sur modèles réduits centrifugés pour quatre pieux soumis à des charges cycliques et des résultats numériques obtenus au moyen de cette nouvelle loi est ensuite proposée. Cette comparaison concerne l’évolution du déplacement en tête de pieu en fonction du nombre de cycles. Une discussion des résultats est ensuite présentée de manière à mieux cerner l’influence de chaque paramètre de la nouvelle t-z cyclique développée et à esquisser les développements futurs à mener.

2 MISE AU POINT D’UNE LOI T-Z CYCLIQUE

2.1 Principes

Dans sa version la plus générale, la loi t-z cyclique entre le frottement qs et le déplacement tangentiel ut est traduite par l’équation 1. Les neuf paramètres de cette loi sont présentés dans le tableau 1.

(1)

Les paramètres auxiliaires A et R sont définis de la manière suivante : le paramètre A contrôle l’augmentation de la raideur de

l’interface sol-pieu lors des déchargements. Cette augmentation est d’autant plus importante que le déchargement est réalisé à un niveau de chargement élevé ;

le paramètre R gère l’augmentation de la raideur de l’interface sol-pieu à chaque inversion de cycle. Il comprend les paramètres et définis dans le tableau 1.

Lors du premier cycle de chargement, la relation entre le frottement qs et le déplacement relatif ut peut être simplifiée :

(2)

Le second terme comprenant les paramètres , et traduit le durcissement monotone. Il est maximal lorsque le

comme le recommande l’Eurocode7 Partie 1 (Eurocod2005), lorsque les exigences de la structure portée en termes de déplacement sont essentielles. En France, ces loislargement utilisées et s’appuient essentiellempropositions de Frank et Zhao (1982) à la fois poet les sols grenus. Elles sont toutefois limitées au cas daxiales monotones ne dépassant la charge de fluane permettent pas de rendre compte des phénomènes observés dans le cas où le pieu est soumis à des charges cy

ttu

t

u

ss eueqq 10

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déplacement relatif ut atteint la valeur . En l’absence de durcissement, la relation se simplifie sous la forme suivante :

(3)

On peut remarquer que le paramètre correspond au

déplacement relatif de l’interface sol-pieu permettant de mobiliser environ 63 % du frottement maximal en l’absence de durcissement. Pour des chargements monotones, les paramètres qs0 et peuvent être déterminés respectivement selon la norme française d’application de l’Eurocode 7 (AFNOR 2012) relative aux fondations profondes et selon les lois de Frank et Zhao (1982) (Figure 1). Les autres paramètres sont à caler sur des essais cycliques de chargements de pieu. Actuellement, ils ne peuvent pas être déduits à partir de propriétés usuelles du sol comme le module pressiométrique, la pression limite ou la résistance de cône. Tableau 1. Définition des paramètres de la loi t-z cyclique

Paramètres généraux i nombre d’inversion de cycle de chargement (i ≥ 1)

qs ;i frottement mobilisé à l’inversion de charge (qs ;i = 0 pour i = 1)

ut ;i déplacement relatif de l’interface sol-pieu au cycle i (ut ;i = 0

pour i = 1)

somme des valeurs absolues des déplacements relatifs

Paramètres auxiliaires

Mo ilisatiob n du frottement sous chargement monotone – 2 paramètres qs0

[kPa] frottement mobilisable sous chargement monotone pour de

grands déplacements

[cm]

paramètre de mobilisation du frottement (plus ce paramètre est faible, plus le déplacement pour mobiliser un frottement

ide) important est faible et plus l’interface sol-pieu est rigRadoucissement/durcissement monotone – 3 paramètres

[c ] m

paramètre fixant le déplacement pour laquelle le durcissement est maximal

[kPa] paramètre contrôlant l’amplitude du durcissement

paramètre de calage égal à 2 Phénomènes cycliques – 4 paramètres

qs [k ] Pa amplitude de radoucissement ou de durcissement cyclique

[cm]

paramètre contrôlant la vitesse de radoucissement ou de durcissement cyclique

paramètre contrôlant l’adaptation, l’accommodation et le rochet ou la relaxation

paramètre contrôlant la vitesse d’adaptation, d’accommodation et de rochet ou de relaxation

0

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0 0.2 0.4 0.6 0.8 1 1.2 1.4

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q s [k

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Frank et Zhao - Sols finst-z cyclique - Sols finsFrank et Zhao - Sols pulvérulentst-z cyclique - Sols pulvérulents

tu

ss eqq 10

Figure 1. Comparaison entre la loi de Frank et Zhao et la loi t-z cyclique pour une sollicitation monotone.

2.2 Potentialités de la loi t-z cyclique

La loi t-z cyclique a été élaborée de manière à rendre compte les principaux phénomènes cycliques observés lors d’essais de cisaillement cyclique en laboratoire :

1;; ititts uuu le durcissement ou le radoucissement cyclique pour des essais de cisaillement en déplacement symétrique (Figure 2a) ;

la relaxation ou non pour des essais de cisaillement en déplacement non symétrique (Figures 2b et 2c) ;

le rochet plus ou moins important pour des essais de cisaillement en contrainte non symétrique.

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Figures 2. Modélisation d’essai de cisaillement suivant différentes conditions (a : déplacement symétrique, b : déplacement non symétrique – relaxation, c : déplacement non symétrique – relaxation nulle)

2.3 Comportement de la base du pieu

La loi utilisée pour modéliser l’enfoncement de la base d’un pieu up a été développée sur le même modèle que la loi précédente. Elle est formulée directement en termes d’effort axial mobilisable sous la base du pieu et comprend neuf paramètres dans sa version la plus évoluée. Pour des chargements monotones, les paramètres Qp0 et sont toujours calés respectivement selon la norme française d’application de l’Eurocode 7 relative aux fondations profondes (AFNOR 2012) et selon les lois de Frank et Zhao (1982).

Un paramètre complémentaire pr (variant entre 0 et 1) a été introduit de manière à prendre en compte le fait que la pointe d’un pieu sollicitée axialement peut se trouver au-dessus de sa position d’origine. Plus il est faible, moins l’effort mobilisé sous la pointe du pieu, tant que ce dernier est au-dessus de sa position initiale, est important (Figure 3).

0

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400

600

800

1000

1200

1400

1600

1800

-3 -2 -1 0 1 2 3

up [cm]

Qp [

kN]

Figure 3. Modélisation de la mobilisation de l’effort de pointe

3 PRÉSENTATION DES ESSAIS EN CENTRIFUGEUSE

Le modèle réduit considéré, à l’échelle du 1/23ème, est soumis à une accélération centrifuge de 23×g. Les propriétés du sable sont les suivantes (Jardine et al. 2009, Andria-Ntoanina et al. 2010) : dmin = 1395-1408 kg/m3, dmax = 1755 kg/m3, s = 2650 kg/m3, d10=0,15 mm, d50=0,207-0,210 mm, d60=0,23 mm et CU=1,49-1,43. Le pieu modèle a les caractéristiques suivantes : forme cylindrique « pointe » plate, diamètre de 18 mm, longueur de 590 mm, fiche de 560 mm, barreau d’aluminium de masse totale 0,405 kg, rugosité à l’état « neuf » définie par Rt=112 μm, Ra=25,9 μm soit Rt/d50=0,54 et Ra/d50=0,13, rugosité à l’état « usé » (après 10 essais) définie par Rt=90 μm, Ra=23,9 μm soit Rt/d50=0,44, Ra/d50=0,12. Les massifs sont

reconstitués par pluviation de sable sec dans un conteneur double (Figure 4).

Les expérimentations réalisées comprennent deux séries d’essais monotones permettant de déterminer la résistance ultime du pieu en compression (Qp) et en traction (Qs). Quatre essais cycliques ont ensuite été réalisés : la charge atteint d’abord la valeur Vm puis oscille entre Vmax = Vm + Vc et Vmin = Vm - Vc (Tableau 2).

Tableau 2. Programme de chargement des essais cycliques

Vm/Qp Vc/Qp

Essai 1 (traction cyclique) -0,33 0,05

Essai 2 (traction cyclique) -0,133 0,133

Essai 3 (cyclique alterné – compression dominante) 0,3 0,4

Essai 4 (cyclique alterné) 0 0,133

4 ANALYSE DES RÉSULTATS OBTENUS

4.1 Présentation des résultats

Les paramètres de calcul (Tableau 3) sont calés sur les deux essais monotones en traction (Figure 4) et en compression (Figure 5) en supposant le module de Young du pieu égal à 10 GPa et sur l’essai cyclique 3 qui présente le comportement apriori le plus complexe à modéliser.

Tableau 3. Propriétés des paramètres de la loi t-z pour le frottement axial et pour la résistance de pointe

Frottement axial qs0 102 kPa 1 cm 30 kPa 1,8 cm 2 qs - 83 kPa 25 cm 20 0,05

Résistance de pointe Qp0 1615 kN 0,1 cm ___ ___ ___

Qp 300 kN 5 cm 6 0,1 pr 0,05

4.2 Analyse des résultats

La figure 6 propose, pour les quatre essais, la comparaison entre les résultats expérimentaux et les résultats numériques. Expérimentalement, la rupture du pieu est obtenue pour les essais 2, 3 et 4 avec un défaut de résistance à la traction du sol. Pour l’essai 1, alors que l’effort de traction appliqué est plus important de tous les essais réalisés, la rupture n’a pas été atteinte pour le nombre de cycles effectués. Il est néanmoins très probable que le nombre de cycles effectués est insuffisant. Pour l’essai 3, la rupture en compression peut être observée puisque l’enfoncement du pieu est supérieur à 10 % de son diamètre. Les efforts de traction atteints traduisant la rupture du pieu pour les essais 2, 3 et 4 sont très différents : 1204,6 kN, 843,22 kN et 317 kN. Ce résultat montre que le frottement mobilisable le long du fût du pieu, après un grand nombre de cycles de chargement, varie en fonction du chargement appliqué. Dans l’état actuel, la loi t-z proposée n’est pas en mesure de rendre compte ce phénomène car les paramètres qs et uts ne traduisent pas suffisamment les effets des cycles sur la résistance de l’interface.

Les résultats obtenus indiquent toutefois que le modèle développé rend compte de manière plutôt satisfaisante des déplacements du pieu lors des premiers cycles de chargement. Pour des nombres de cycles élevés, les tendances restent plutôt bien appréhendées même si les amplitudes de déplacement ne sont pas correctes.

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Figure 4. Modélisation de l’essai de traction

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5 CONCLUSION

Une loi d’interaction t-z cyclique a été développée pour le calcul des pieux sous charge axiale. Des tentatives de validation ont été effectuées en confrontant des résultats numériques obtenus avec cette loi et des expérimentations en centrifugeuse pour des chargements cycliques en traction et en compression alternés ou non. Des améliorations significatives restent à apporter. Les premiers résultats sont encourageants car ils traduisent au moins dans leurs grandes lignes les principaux comportements du pieu.

Calcul - Essai-3

+/- B/10

6 REMERCIEMENTS

Les résultats présentés dans cette communication ont été acquis dans le cadre du Projet ANR et du Projet National Français SOLCYP (SOLlicitations Cycliques sur les Pieux).

7 REFERENCES

Eurocode 7 – Part 1. (2004) Calcul Géotechnique – Partie 1: Règles Générales. Comité Européen de Normalisation, Bruxelles.

Frank, R. et Zhao, S. (1982). Estimation par les paramètres pressiométriques de l'enfoncement sous charge axiale de pieux forés dans des sols fins. Bulletin Liaison Laboratoire Ponts Chaussées 119 17-24.

Chin J. T. and Poulos H. G. (1992) Cyclic axial loading analyses: a comparative study. Computers and Geotechnics, 13, 137-158.

Randolph M.F. (1986) RATZ: Load transfer analysis of axially loaded piles, Report Geo: 86033, Department of Civil Engineering, The University of Western Australia.

-50

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0

10

0 200 400 600 800 1000

CyclesEssai-4

w [m

m]

-B/10

Calcul - Essai-4

Mesures - Essai-4

Figures 6. Modélisation des quatre essais cycliques (w : déplacement axial du pieu – > 0 : tassement – < 0 : soulèvement)

Lemaitre, J. et Chaboche, J.L. (2009). Mécanique des Matériaux Solides (éd. 2ème). Paris: Dunod.

AFNOR. (2012) Dimensionnement des fondations profondes. Norme NF P 94-292, Paris.

Jardine R.J., Zhu B.T., Foray P. & Dalton, C.P. (2009). Experimental arrangements for the investigation of soil stresses developed around a displacement pile. Soil and Foundations 49(5): 661-673.

Andria-Ntoanina I., Canou J. et Dupla J.-C. (2010) Caractérisation mécanique du sable de Fontainebleau NE34 à l’appareil triaxial sous cisaillement monotone. Rapport SOLCYP, 23p.

2339

Deformation behavior of single pile in silt under long-term cyclic axial loading

Comportement d’un pieu isolé sous chargement axial cyclique de longue durée dans un limon

Chen R.P., Ren Y., Zhu B. , Chen Y.M.MOE Key Laboratory of Soft Soils and Geoenvironmental Engineering, Zhejiang University, Hangzhou, China

ABSTRACT: Evaluating the response of piles to cyclic loading is a crucial part in the design of piled-embankment over soft ground.In this paper, a series of large-scale model tests were performed to investigate the response of pile in silt under cyclic axial loading.Heavily instrumented piles were used in the tests. The study is focused on the accumulation of permanent dispalecemtn of the pilesunder long-term cyclic loading. Piles were tested at differernt cyclic loading levels and subjected up to 50,000 cycles of loading ineach test. The accumulated settlement was found to be strongly dependent on the characteristics of the applied cyclic loads. The pileswere found not to produce any increase in accumulated settlement if the cyclic loading amplitude is less than a certain thresholdvalue. A simple method is proposed to predict the accumulated settlement of single pile due to very large number of loading cycles.The idea of a cyclic deformation diagram for analyzing the influence of charcteristics of cyclic loads on the deformation behavior wasalso developed.

RÉSUMÉ : L'évaluation de la réponse des pieux vis-à-vis d’un chargement cyclique est un élément essentiel dans la conception desremblais sur sols mous. Dans cet article, une série d'essais sur maquette à grande échelle a été réalisée pour étudier la réponse de pieuxsous chargement axial cyclique dans le limon. L'étude est centrée sur l'accumulation des déplacements permanents des pieux souschargement cyclique de longue durée. Les pieux ont été testés à différents niveaux de charge cyclique et soumis à 50000 cycles pourchaque test. Les tassements accumulés sont fortement dépendants des charges cycliques appliquées. Aucune augmentation detassements n’est constatée si l'amplitude de chargement cyclique est inférieure à un certain seuil. Une méthode simple est proposéepour prédire le tassement cumulé d’un pieu isolé en fonction d’un très grand nombre de cycles de chargement. L'idée d'un diagrammede déformation cyclique pour analyser l'influence des caractéristiques des charges cycliques sur le comportement en déformation aégalement été développée.

EYWORDS: Pile; model test; accumulated settlement; cyclic loading

1 INTRODUCTION

Piles are commonly used to support high-speed railwayembankment in soft ground, which are exposed not only to theheavy loads from superstructure self-weight, but also to thelong-term “one-way” cyclic loads induced by high-speed trainsthroughout their service life. However, available designexperiences on the long-term response of the pile in silt tocyclic axial loading are very limited, due to the fact that theexisting data obtained from laboratory tests and fieldmeasurement are insufficient. This results in uncertainty in thedesign and always leads to an over-conservative design of thepile foundation.

The response of pile subjected to cyclic loading is verycomplex and model test is the most effective and reliable way tostudy it and its influencing factors. Laboratory and field tests(Chan and Hanna 1980; Lee and Poulos 1991; Karlsrud et al.1993) have shown that there are two main effects of cyclic axialloading on piles: (1). a reduction in load capactiy and pile-soilsystem stiffness; (2). an increase in settlement of piles. Poulos(1989) reviewed several cyclic loading tests performed in sandand stated that the accumulation of permanent displacementwith increasing load cycles was expected to dominate under“one-way” cyclic loading, particularly if “strain-softening” behavior can occur at the pile-soil interface. The accumulationof the permanent displacement principally depends on the cyclicload level. Briaud and Felio (1986) analyzed the published dataand concluded that a load threshold exists above which failureoccurs by plunging due to cyclic loading and the value of thisthreshold is 80% of the ultimate pile capacity on average.

In previous studies, most of the piles were loaded less than500 cycles of loading. There have been few researches on theresponse of pile to long-term cyclic loading at present.

This paper describes the results of a series of large-scalemodel tests on single stiff piles in saturated silt to study theaccumulation of permanent displacement of the piles. The testshave been performed using heavily instrumented model piles. Asimple method for predicting the accumulation of permanentdisplacement of pile to long-term cyclic axial loading isproposed. The idea of a cyclic deformation diagram foranalyzing the influence of charcteristics of cyclic loads on thedeformation behavior was also developed.

2 DESCRIPTIONS OF EXPERIMENTS

2.1 Test site and soil characteristics

The present large-scale model tests were carried out in a big soiltank at Zhejiang University (Fig 1). This soil tank has andimension of 15×5m in plan view and a depth of 6m.

The soil used in the tests is low cohesive silt. Grading testsshowed that the soil contains 10% sand, 85% silt and 5% clay.The characteristics of the soil are summarized in Table 1.Laboratory tests show that the prepared soil had an averagewater content of 28.5% and an effective internal friction angleof 30o.

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Figure 1. The big soil tank at Zhejiang University

Table 1. Main soil properties

Property Value

D50 particle size (mm) 0.032

Specific gravity, GS 2.69

Plastic limit, WL 22.6

Liquid limit, IL

Plasticity index, PI

31.7

9.1

2.2 Model pile

The used instrumented model pile was closed-ended steel tubepile with an outer diameter (d) of 168 mm, a wall thickness of 7mm. The pile had a cone-shaped tip end with a cone angle of60o. The model pile was designed to be assembled from foursegments to give a full length of 4.2 m. The instrumentationconsisted of axial load cells (ALC), total pressure transducers(TPT) and pore pressure transducers (PPT).

Figure 2. Schematic diagram of the model pile (Unit: mm)

2.3 Characteristics of cyclic loading

The characteristics of the applied cyclic load are uniquelydefined using two independent parameters:

/s uSLR P P (1)/c uCLR P P (2)

in which uP refers to the static ultimate capacity of the pile incompression, and sP is the minimum in a load cycle and cP isthe cyclic load amplitude. In present study, sP and cP simulatedthe self-weight from the superstructure and the cyclic loadinduced by the high-speed trains, respectively. A visualinterpretation of the load ratios is given in Fig. 3.

PU

P

CLR=

SLR=0.2

0.20.4

0.60.8

PC

PS

CLR=PC/PU

SLR=PS/PU

0

Figure 3. The characteristics of cyclic loading

In this study, the cyclic load frequency of 3Hz was used forthe tests and the sampling frequency was of 50 Hz.

3 TEST RESULTS

3.1 Accumulated settlement

The overall pattern of accumulated settlement of the pile withSLR of 0.3 is presented in Fig. 4 by normalizing theaccumulated permanent displacement s by the pile diameter d.The values of the CLR in this series ranged from 0.1 to 0.6. Itcan be found that the ways in which displacement developed ishighly dependent on the amplitudes of cyclic load which can berepresented by the cyclic load ratio (CLR).

In the test with the smallest CLR of 0.1, extreme smallpermanent displacement, of approximately 0.04%d wasproduced in the first three cycles and remained nearly constantfrom cycle No. 3 to 50,000. For the CLRs ranging from 0.2 to0.5, the permanent accumulated displacement increasedgradually with the increasing number of cycles and also withthe increasing magnitude of cyclic load. For the tests with theCLR of 0.2, 0.3, 0.4 and 0.5, the permanent displacements at theend of the tests were 0.15%d, 0.35%d, 0.56%d and 1.26%d,respectively. The permanent displacement increased rapidly atinitial stage and had the highest rate of displacement increase inthe first few cycles, and then it kept increasing continuouslywith a decreased rate of displacement increase and seemed toincrease without a final and constant value. For the test withvery large cyclic load, such as CLR=0.6, the pile head moveddownward in a very “unstable” way marked by a quick plunging during the test and the pile failed with a total accumulateddisplacement of 10%d in 2,147 cycles.

100 101 102 103 104 1050.00

0.02

0.04

0.06

s/d

(%)

N

CLR=0.1SLR=0.3

(a). CLR=0.1

100 101 102 103 104 1050.0

0.5

1.0

1.5

2.0

2.5

3.0

s/d

(%)

N

CLR=0.1CLR=0.2CLR=0.3CLR=0.4CLR=0.5CLR=0.6

SLR=0.3

s=10%d after2174 cycles

(b). CLR=0.1~0.6Figure 4. Normalized permanent displacement (s/d) with number of

cycles (SLR=0.3)

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The aforementioned patterns of behavior shown in Fig. 4 canalso be found in other tests with SLR of different values. Theaccumulated displacement for each test fell qualitatively intoany of the three distinct patterns shown in Fig. 4:

(1) no accumulated displacement, as exhibited during the testwith CLR of 0.1;

(2) continuing displacement, as exhibited during the testswith CLR ranging from 0.2 to 0.5;

(3) failure, as exhibited during the test with CLR of 0.6.For the first case, the pile-soil system seems not to be

influenced by the cyclic loading and is in elastic range; onlysmall accumulated displacement was produced during the firstfew cycles. For the second case, the pile-soil system wasinfluenced to some degree and partially entered plastic range;the pile head showed continuing downward movement withoutany apparent limit and the accumulation of displacementdepended on both of cyclic load level and number of load cycles,and high cyclic load level and large number of cycles producedlarger permanent displacement. For the last case, the cyclicloading had brought severe damage to the pile-soil system andthe pile fully entered the plastic range; the pile head movedcontinuously downward at a rapid rate up to the end of the testand a plunging failure might occur in some cases.

Thus, to divide the accumulated displacement responses forthe tests with a given SLR two critical values of CLR aredefined here, named minimum cyclic load ratio (MCLR) andfailure cyclic load ratio (FCLR), respectively. For CLR smallerthan the MCLR, the pile was in elastic range; for CLR greaterthan the FCLR, the damage to the pile-soil system was severeand the always caused “failure”.

The MCLR was found to be of 0.1 in all the tests and shownto be unaffected by the SLR, and it can be inferred that if theapplied cyclic loads remained less than 10% of the ultimate pilestatic capacity, the response of the pile can be considered to betotal elastic and the permanent displacement was negligibleafter first several cycles.

The FCLR was found to be of 0.5 for the tests with SLRranging from 0.2 to 0.4. However, in the case of the test withSLR of 0.1, the pile produced large permanent displacementwith CLR of 0.4, and it showed the tendency that lesser cyclicloads were required to cause large permanent displacement forthe pile with very small SLR. Briaud and Felio (1986) reviewedthe previous cyclic load tests and concluded that a threshold ofpeak load ratio (CLR+SLR) existed above which largepermanent displacement occurred and the value of thatthreshold was about 0.8 on average. However, the tests resultssuggest that the large permanent displacement depended moreon the magnitude of the cyclic load rather than the peak cyclicload. It can be inferred from the results that large permanentdisplacement occurred if the magnitude of the applied cyclicload exceeds the 50% of the ultimate pile static capacity.

3.2 Prediction method

To investigate the evolution of the permanent displacement inthe tests in which the permanent displacement are identified as“continuing displacement”, the results are replotted on double logarithmic scales and the evolution of the permanentdisplacement is evaluated in terms of the dimensionless ratio

0( ) N

s s

s N s ss s

(3)

which expresses the magnitude of the permanent displacement( )s N caused by cyclic loading in terms of the displacement ss

that would occur in a static load test when the load is equivalentto the maximum cyclic load (as defined by ( ) uSLR CLR P ).The 0s and Ns refer to the permanent displacement in first andN’th cycle, respectively.

The results, plotted in Fig. 5, show that the trend in the datafollows the exponential behavior which appears as straight linesin double logarithmic axes. This suggests that the permanentdisplacement due to cyclic loading can be predicted by thefollowing power model:

( ) b

s

s N ANs

(4)

where A and b are two parameters. It is observed in Fig. 5 thatall slopes are almost equal. This suggests that b is independentof the load characteristics within the observed range. It isintroduced into (4) to represent the influence of loadcharacteristics on parameter in the following form:

( 1) ( )m nA a SLR CLR (5)

where a , m and n are three calibration parameters. Clearly,when 0CLR , then 0A and no accumulated displacementwill occur under static load. Also, when 0SLR then

( )nA a CLR indicates that the accumulated displacementdepends only on CLR. Thus, substituting (5) into (4) gives thefollowing model for accumulated permanent displacement:

( ) ( 1) ( )m n b

s

s N a SLR CLR Ns

(6)

The expression in Equation (6) was fitted to the data in Fig.5 to empirically determine values of these parameters and back-calculated parameters a , m , n and b for the tests are 0.054,0.68, 1.24 and 0.23. The predicted results are shown by thedotted lines in Fig. 5 and it appears that the influences of theload characteristics on permanent displacement are reflectedwell in the prediction. The closeness of the fit up to 45 10cycles indicates that, in the absence of further experimental data,it might be reasonable to extrapolate beyond 45 10N . Furtherdata are, of course, required to confirm this hypothesis.

100 101 102 103 1040.01

0.1

1SLR=0.3

s/s

s

N

Experimental:CLR=0.2CLR=0.3CLR=0.4

Figure 5. Measured and predicted accumulated displacementThe dotted lines are obtained using Equation (6)

4 CYCLIC DEFORMATION DIAGRAM

Poulos (1988) proposed the idea of cyclic stability diagram toinvestigate the capacity degradation caused by cyclic loading. Inthis study, similar concept is used and the idea of a diagramnamed cyclic deformation diagram is developed. The cyclicdefromation diagram for the model piles is shown in Fig. 6. Inthe diagram, the aforementioned three types of displacementresponse are represented by different symbols. Therefore, threemain regions can be identified on the diagram shown in Fig. 6:

(1) A stable (elastic) region I in which the cyclic loadinghas no influence on the pile responses and the displacementresponse is the type of “no accumulated displacement”.

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(2) A serviceability region II in which the cyclic loadinghas some influence on the pile response and the displacementresponse can be identified as “continuing displacement”.

(3) A unstable region III in which cyclic loading causessevere damage for the pile to produce very large permanentdisplacement and in some cases a plunging failure occurs.

0.0 0.2 0.4 0.6 0.8 1.00.0

0.2

0.4

0.6

0.8

1.0 S

D

C

B

III

II

I

Proposed lower boundary of unstable zoneProposed upper boundary of stable zoneTest results - no accumulated displacementTest results - continuing displacementTest results - failureMcAnoy et al. (1982) - continuing settlementMcAnoy et al. (1982) - failure at N=564Stevens (1978) - no settlementStevens (1978) - continuing settlementStevens (1978) - plunging failureKarlsrud et al. (1986) - failure after 100 cycles

Cyc

liclo

adra

tio(C

LR)

Static load ratio (SLR)

zone I: Stablezone II: Serviceabilityzone III: Unstable

A

Figure 6. Cyclic deformation diagram for pile in silt (N=50,000)

The upper boundary to the cyclic permanent displacement isthe straight line (CS: SLR+CLR=1) that represent thecombinations of SLR and CLR necessary to cause a failure ofpile without cyclic effects being considered. The other two linesplotted in this diagram represent the approximate boundariesbetween the stable region, the serviceability region and theunstable region. These lines are defined by the followingrelations:

Upper boundary of stable zone (line AB and BC):

0.5 0 0.41.2 1 0.4 1.0CLR SLR

CLR SLR SLR

(7)

Lower boundary of unstable zone (line DC):

5 1, 0 1CLR SLR SLR (8)

Fig. 6 also plots the other test results of field or model testson axial cyclically loaded pile. It can be seen that theseproposed lines are consistent with the experimental data andthus it is indicated that the proposed three regions are capable ofreasonably identifying the deformation behavior of pile undervarious load combinations. A diagram such as shown in Fig. 6represents the permanent displacement of a pile for a specifiednumber of cycles, N. As N increases, the stable region willremain unchanged and the unstable region may increase as thepermanent displacement increases.

In the pile design, it is very convenient to determine thedeformation behavior of the pile to cyclic loading using thisdiagram. The most conservative design is to have the cyclicloads in the stable region which means that pile will not beaffected by cyclic loading and issues of the permanentdisplacement can be totally ignored. If the designed cyclic loadis in the serviceability region, the permanent displacementaccumulates in “stable” way and depends on both of the number of cycles and the load characteristics; and it can be predictedusing the proposed simple method mentioned above. For a safedesign, it should avoid the cyclic load to be in the unstableregion in which cyclic loading will result in very largepermanent displacement and even a plunging failure.

5 CONCLUSION

A series of tests were conducted on large-scale model pilessubjected to long-term cyclic axial loading. The deformationbehavior of the piles in silt to cyclic loading was investigated.

The evolution of permanent displacement highly depends onthe magnitude of cyclic load. In general, the accumulation ofpermanent displacement increases with increasing cyclic loadamplitude and increasing number of cycles. However, the pilebehaves in an elastic manner and does not accumulate anydeformation after the first few cycles of loading if themagnitude of cyclic load is less than 10% of the ultimate pilecapacity. Very large permanent displacement, even plungingfailure, occurs when magnitude of cyclic load exceeds 50% ofthe ultimate pile capacity. This suggests that the magnitude ofthe cyclic load be kept below 50% of the ultimate capacity toavoid large permanent displacement in the design.

These results provide a better understanding of thedeformation behavior of pile in silt to long-term cyclic axialloading, and can be used to optimize the designs of pilefoundations that resist cyclic loads in service.

6 ACKNOWLEDGEMENTS

The work was supported by the National Natural ScienceFoundation of China (Grant Nos. 51225804 and U1234204).

7 REFERENCES

ASTM 2010. D2487-10. Standard Practice for Classification of Soils forEngineering Purposes (Unified Soil Classification System). ASTMInternational.

Briaud J.L and Felio G.Y. 1986. Cyclic axial loads on piles: Analysis ofexisting data. Canadian Geotechnical Journal, 23, 362-371.

Chan S.F. and Hanna T.H. 1980. Repeated loading on single piles insand. Journal of Geotechnical Engineering Division, 106, 171-188.

Karlsrud K., Nadim F. and Haugen T. 1986. Piles in clay under cyclicaxial loading-field tests and computational modeling. Proc., 3rd Int.Conf., Numerical Methods in Offshore Piling, Nantes, France, 165-190.

Karlsrud K., Nowacki F. and Kalsnes B. 1993. Response in soft clayand silt deposits to static and cyclic loading based on recentinstrumented pile load test. Proc. SUT Int. Conf, Kluwer, Dordrecht,549-584.

Lee C.Y. and Poulos H.G. 1991. Tests on model instrumented groutedpiles in offshore calcareous soil. Journal of GeotechnicalEngineering, 117, 1738-1753.

McAnoy R.P.L., Cashman A.C. and Purvis D. 1982. Cyclic tensiletesting of a pile in glacial till. Proc., 2nd Conf., Numerical Methodsin Offshore Piling, Austin, Tex., 257-292.

O’Riordan N., Ross A. and Allwright R. 2003. Long-term settlement ofpiles under repetitive loading from trains. Transportationgeotechnics, Thomas Telford, London, 67-74.

Poulos H. G. (1988). “Cyclic stability diagram for axially loaded piles.” Journal of Geotechnical Engineering, 114, 877-895.

Poulos H.G. 1989. Cyclic axial loading analysis of piles in sand. Journalof Geotechnical Engineering, 115, 836-852.

Stevens J.B. 1978. Prediction of pile response to vibratory loads. Proc.,10th OTC Conf., Houston, Tex., Vol. 3, 2213-2223.

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Time-Varying Dynamic Properties of Offshore Wind Turbines Evaluated by ModalTesting

Étude expérimentale de l’évolution temporelle des propriétés dynamiques d’éoliennes maritimes

Damgaard M., Andersen J.K.F.Vestas Turbines R&D, Denmark

Ibsen L.B., Andersen L.V.Department of Civil Engineering, Aalborg University, Denmark

ABSTRACT: Modal frequencies and damping ratios of civil engineering structures are often used as damage-sensitive features, sincechanges in the dynamic characteristics of the structures may indicate structural damage. For offshore wind turbine structures, themodal parameters are influenced by environmental impacts that change boundary conditions, irreversible soil deformations andinherent structural properties. The excitation frequencies related to the environmental loads and the passage of blades past the towerare so low that a proper estimate of the modal parameters are needed in order to avoid strong resonance of the wind turbine structure.In this paper, free vibration tests and a numerical Winkler type approach are used to evaluate the dynamic properties of a total of 30offshore wind turbines located in the North Sea. Analyses indicate time-varying eigenfrequencies and damping ratios of the loweststructural eigenmode. Isolating the oscillation oil damper performance, moveable seabed conditions may lead to the observed timedependency.

RÉSUMÉ: Les fréquences modales et les taux d'amortissement des structures de génie civil sont souvent utilisés comme indicateur dedommages car l’évolution de la réponse dynamique des structures peut indiquer des dégâts structuraux. Pour des structures comme leséoliennes maritimes, les paramètres modaux sont influencés par la déformation irréversible des sols, les propriétés structurellesinhérentes et les conditions environnementales qui peuvent changer les conditions aux limites. Les fréquences d’excitations liées aux charges environnementales et aux passages des palles sont si basses qu’une estimation correcte des paramètres modaux est nécessairepour éviter une forte résonance de la structure de l'éolienne. Dans cet article, des tests vibratoires et une approche numérique du typeWinkler sont utilisés afin d’évaluer les propriétés dynamiques de 30 éoliennes maritimes situées en mer du Nord. Les analysesrévèlent le changement des fréquences propres et des taux d'amortissement de la plus basse fréquence propre structurelle en fonctiondu temps. En isolant la performance de l'amortisseur oscillant à huile, les changements de conditions du fond marin peuventdémontrer une dépendance temporelle.

KEYWORDS: Free vibration; modal; offshore wind turbine; p-y curve; scour; winkler approach.

1 INTRODUCTION

Recently, offshore wind turbine towers and blades haveincreased significantly in height and length, respectively, withonly a small increase in weight. Therefore, the dynamicresponse of the wind turbine structure occurs in a frequencyrange close to the excitation frequencies related toenvironmental and structural harmonic loads. In this context,sufficient geometrical and material damping in the structure andsoil are required to counteract large amplitudes of vibration.Especially for wind parks characterised by a large degree ofwind-wave misalignment, a proper estimate of the inherentdamping is needed due to low aerodynamic forces out of therotor plane.

The aim of this paper is to investigate the time-varyingeigenfrequency f1 and inherent modal damping δ1 of the lowesteigenmode Φ(1) for offshore wind turbines installed on amonopile foundation. Experimental modal analysis of offshorewind turbines have been studied by several researchers. Basedon free vibration tests, Tarp-Johansen et al. 2009 and Damgaardet al. 2011 have used “rotor-stop” tests to determine each damping contributor to the measured inherent modal dampingδ1 of an offshore wind turbine. Versteijlen et al. 2011 andDevriendt et al. 2012 used the same modal approach to obtainreliable damping estimates. In addition, Versteijlen et al. 2011considered operational modal analysis in order to include theaerodynamic effects on the structure. The theory has beenwidely used for civil engineering structures like bridges andbuildings. However, in the last years the application of

operational modal analysis on wind turbines has been publishedin many excellent papers, see for instance Hansen et al. 2006and Tcherniak et al. 2010. A thorough data processing of morethan 650 free vibration tests on 30 offshore wind turbinestructures are presented in the paper. The variation in thedynamic properties is supported by a numerical Winklerapproach that estimates the modal parameters for differentenvironmental conditions.

2 STRUCTURE AND SITE CONDITIONS

A total of 30 Vestas V90-3MW turbines located in the NorthSea are considered. Each tower is installed on a monopileconnected by a grouted transition piece to the tower base. Thetower height is approximately 60 m, the monopile diameter 4.3m and the water depth 8 m w.r.t. LAT. For each turbine anoscillation damper is placed in the top of the tower. It consistsof a pendulum partly immersed in highly viscous oil, capable ofoscillating in the horizontal directions. The soil consists mainlyof cohesionless soil in the top layers with friction angles φkhigher than 30° followed by cohesive soils with undrained shearstrength cu higher than 90 kPa.

3 MODAL PARAMETER ESTIMATION

By use of two accelerometers placed in the nacelle, the modalparameters of each wind turbine are experimentally estimated

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from the acceleration decay when the turbine generator shutsdown and the blades pitch out of the wind, see Figure 1. Hence,assuming that the wind turbine structure behaves as a single-degree-of-freedom (SDOF) system, the eigenfrequency f1 andmodal damping δ1 are determined by least-squares fitting of alinear function to the zero crossings and to the natural logarithmof the rate of decay of the vibration, respectively. It should benoticed that a wind turbine structure has two closely spacedmodes occurring at nearly identical frequencies (Damgaard etal. 2012), where vibrational energy is transferred from thehighest to the lowest damped mode. Hence, for the dampingestimation of each free vibration test it is ensured that theacceleration of the structure only takes place in the fore-aftdirection y.

Figure 1. Raw output acceleration signal during a “rotor-stop”.

3.1 Winkler Approach

Offshore wind turbines supported by pile foundations aresubjected to lateral cyclic loads. The load-deflection behaviouris often evaluated by a Beam on Nonlinear Winkler Foundation(BNWF) model due to its computationally efficiency andpractical versatility. The tower and pile are modelled asBernoulli-Euler beams and the soil-structure interaction isincorporated via so-called p-y curves suggested by DNV 2011,see Fig. 2. The soil consists of a series of independent soillayers with smooth horizontal boundaries, i.e. no shearing canbe transmitted across the boundaries. Rather than modelling thesoil as a number of discrete springs connected to the elementnodes, this paper uses a consistent approach, where the soil ismodelled as a continuous spring over each element. The nodalforces are then obtained via numerical integration. The reader isreferred to Damgaard et al. 2011 for more information about thecomputational model.

Figure 2. Beam on nonlinear Winkler foundation (BNWF) model.

3.1.1 Soil Damping EstimationIn general, attenuation of wave propagation in the soil isdetermined from geometric damping, i.e. the radiation of wavesinto the subsoil, and material damping caused by the slippage ofsoil grains with respect to each other. However, extensivestudies of wind turbines on a homogeneous or layered groundmade by Andersen 2008 show that geometric dissipation is

insignificant at frequencies below 1 Hz. From the continuummechanics it is known that material damping is related to therelative motion of material points, and the energy dissipation isfrequency-dependent. For a given frequency and deformationlevel, the soil material damping can be approximated to anequivalent viscosity. Based on a static deformation analysis,using the Winkler approach, the following procedure is used todetermine the soil damping ratio ζsoil of the lowest eigenmodeΦ(1): A 10-minutes time-domain simulation of the wind turbine

structure is conducted for a power production situation witha normal turbulence model (IEC 2005) using the aeroelasticcode FLEX (Øye 1996). A correct estimate of the structuraleigenfrequency f1 in the FLEX model is ensured byextending the tower until the eigenfrequency f1 of theWinkler model is reached.

Based on the maximum overturning moment at thetower/foundation interface from the FLEX simulation andincluding wave loads, the horizontal pile deformation ineach nodal point below the seabed is evaluated.

Assuming a load-displacement cycle after the generatorshuts down, as indicated in Figure 3a, the irreversible soildeformations are a measure of energy dissipation. Hence, theenergy dissipation in Figure 3a can be transformed to anequivalent viscous damping model as shown in Figure 3b. Using the theory of linear structural dynamics, the soildamping ζsoil of the lowest eigenmode Φ(1): is determined fromthe global damping matrix C, the angular eigenfrequency ω1,the eigenmode Φ(1) and the modal mass M1 given by

11

(1)(1)T

soil 2 M

CΦΦ(1)

The virgin curve in Figure 3a is determined by the p-y curveformulation given by DNV 2011. The unloading phase isdetermined by the initial stiffness E*

py. Assuming separationbetween the pile and the soil, a shear drag pdrag is introduced.For cohesionless soils, the shear drag depends on the verticaleffective stress σ´

v (Ovesen et al. 2006) given by pdrag=0.6D σ´v,

whereas for cohesive soils the undrained shear strength cu mustbe considered, i.e. pdrag=0.7Dcu.

Figure 3. Hysteresis Loop Method (Nielsen 2004): (a) Load-displacement curve after the wind turbine generator shuts down, (b)Hysteresis loop implied by viscous damping in a harmonic motion withthe amplitude A and the angular eigenfrequency ω1.

4 INTERPRETATION OF RESULTS

Experimental modal testing of 30 offshore wind turbines in theperiod 2006-2011 is presented in Figure 4a and Figure 4b interms of the modal damping δ1 and the eigenperiod T1 of thelowest eigenmode Φ(1), respectively. Using a lognormalprobability distribution, the 5% quantile of the modal dampingδ1 and the eigenperiod T1 are estimated to 0.11 and 2.94 s,respectively. This corresponds to an eigenfrequency f1 of 0.34Hz. As indicated in Figure 4a and Figure 4b, the scatter of theestimated parameters is high. Increasing the R-square valuefrom 0.95 to 0.99, meaning that the fit of the accelerationamplitude peaks and zero crossings explains 99% of the totalvariation in the data about the average, seems to reduce thescatter to a certain extent, see Figure 4c and Figure 4d. Overall,

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a tendency of decreasing modal damping and eigenfrequency isobserved for increasing acceleration level. High structuralaccelerations induce irreversible soil deformations and therebysoil damping activation. However, the oil damper performanceis characterised by optimal damping for low levels ofaccelerations, which may explain the observed behaviour. Inaddition, distinct non-linear soil behaviour occurs for highaccelerations, which reduces the secant stiffness Es and therebythe eigenfrequency f1.

An almost identical mean value and standard deviation ofthe modal parameters have been observed for each wind turbine.Hence, the variation of the modal parameters in Figure 4c andFigure 4d might be caused by the following conditions:

Figure 4. Free vibration tests for a total of 30 offshore wind turbines: (a)Damping histogram, (b) Eigenperiod histogram, (c) Damping vs.acceleration level, (d) Eigenfrequency vs. acceleration level.

Tower damper performance Tidal variation Wind variation Temperature dependent modal parameters Moveable seabed and scour around the foundation

Figure 5. Selected turbine investigation: (a) Tidal and wind variation asa function of time, (b) Modal parameters as a function of time. Data arecollected with same acceleration level and slope of generator speed.

It has been observed that the mass pendulum of the towerdamper in some tests moves exactly with a phase identical tothe phase of the wind turbine, resulting in almost no additionaldamping. To eliminate the variation of the tower damperperformance, data for each turbine is investigated for the sameslope of generator speed when the blades pitch out of the windand for the same acceleration level. As an example, Figure 5shows the comparison of the measured 10-minutes wind speed

and tidal variation together with the modal parameters for aselected turbine. One one-year measurements of the tidal levelsat the wind park show only a maximum difference betweenhighest and lowest astronomical tide of 2 m. It might then beassumed that the tidal variation at the wind park has negligibleimpact on the magnitude of the modal parameters. The sameconclusion can be drawn regarding the variation in the windspeed and temperature during the tests. The aerodynamicdamping is very low, when the blades pitch out of the wind, anda temperature change from -73° to 93° only changes theYoung’s modulus of elasticity Esteel with 5% (Nielsen 2004).Based on a Winkler model this corresponds to a change in theeigenfrequency f1 of only 0.5%. In conclusion, assuming thatthe tower damper contributes with the same damping value inFigure 5b, the time-dependent modal parameters might becaused by erosion of soil particles near the monopilefoundation.

Figure 6. Scour and backfilling analysis based on a Beam on NonlinearWinkler Foundation model: (a) Eigenfrequency f1 as a function of scourdepth and backfilling height, (b) Soil damping δsoil as a function of scourdepth and backfilling height.

1.1 Scour and Backfilling

When a pile is installed in a loose sedimentary bed, a scour holewill form around the pile. The phenomenon is of highimportance, since the structural eigenfrequency and soildamping contribution will change and in worst case lead tofatigue damage and, eventually, failure. Based on experimentaltests, Sumer et al. 1992 stated that the mean value of theequilibrium scour depth for a vertical cylinder in steady currentis given by 1.3D, where D is the diameter of the cylinder.However, for combined current and wave conditions the scourdepth is difficult to determine, since wave action tends to reducethe scour depth (Høgedal and Hald 2005).

As no scour protection is present for the investigated windturbine structures in this paper, the variation of theeigenfrequency f1 and soil damping δsoil, caused by sedimenttransportation at seabed, is estimated using a Winkler approach.Different scour depths and backfill heights are considered for awind speed of 13 m/s with maximum depth and height equal to1.3D, respectively. The vertical effective stress p0 is reducedlinearly with depth to a depth equal to 3D below the base of thecurrent scour hole. As expected, Figure 6a shows a decreasingeigenfrequency f1 for increasing scour depth. Assumingcohesionless backfill material with a friction angle φk of 28°,

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the eigenfrequency f1 tends to increase for increasing backfillheight. The material soil damping δsoil highly depends on thepile deflection and the initial stiffness E*

py for each soil layer,see Figure 3. However, the pile deflection at the base of thescour hole only increases to a certain scour depth, and the initialstiffness E*

py depends on the strength of each soil layer and thescour depth. Hence, for increasing scour depth the piledeflection and initial soil stiffness might increase or decreaserelative to each other. This may in turn explain the observedbehaviour of the soil damping δsoil in Figure 6b.

Over a period of time the relative density Id of the backfillmaterial might be increased due to the presence of wavesinducing depth compaction (Sørensen et al. 2010). Hansson etal. 2005 have reported friction angles above 40° forFrederikshavn sand. Figure 7 shows the eigenfrequency f1 andsoil damping δsoil as a function of the strength of the backfillmaterial after the scour hole is replaced by the backfill material.

In conclusion, the Winkler approach shows a variation ofthe eigenfrequency f1 caused by sediment transportation at theseabed level of 8%. The model indicates a soil damping δsoil inthe range of 0.05-0.08 logarithmic decrement. Hence,comparing these results with the experimental findings in Figure5, the time-varying modal parameters of the investigatedoffshore wind turbines might be caused by sedimenttransportation at seabed.

2 CONCLUSION

Wind energy is a rapidly growing interdisciplinary field thatinvolves many different disciplines within civil engineering.The dynamic behaviour of the wind turbine structure isdetermined by a complex interaction of components and sub-systems. A full understanding of the structural modalparameters is crucial in order to assess the fatigue damageaccumulation during the lifetime of the wind turbine structure.

Experimental and numerical investigations of the dynamicproperties of offshore wind turbine structures installed on amonopile foundation have been presented in this paper. Basedon a total of 665 free vibration tests, time-varying modalparameters are observed, which is supported by a Winklerapproach. Several interesting observations can be made:

Figure 7. Strength of backfilled material based on a Beam on NonlinearWinkler Foundation model: (a) Eigenfrequency f1 as a function of thefriction angle φk

’ of the backfill material, (b) Soil damping δsoil as afunction of the friction angle φk

’ of the backfill material.

Experimental testing indicates a high variation in theeigenfrequency f1 and the modal damping δ1. A 5% quantile

of 0.11 logarithmic decrement is observed, which correspondsvery well with the findings for each considered turbine. Eliminating the tower damper performance tends to reduce

the large variation of the modal parameters. However,distinctly time-varying eigenfrequencies f1 and modaldamping values δ1 are still obtained.

A Beam on a Winkler foundation model indicates that theobserved time-dependencies might be caused by sedimenttransportation at seabed. Scour development and backfillingchange the eigenfrequency f1 with 8%, and the soil dampingδsoil varies in the range 0.05-0.08.

3 ACKNOWLEDGEMENTS

The authors are grateful for the financial support from theresearch project Cost Effective Monopile Design.

4 REFERENCES

Andersen L.V. 2008. Assessment of Lumped-Parameter Models for RigidFootings. Computers and Structures. 88, 1333-1347.

Damgaard M, Ibsen L.B, Andersen L.V. and Andersen J.K.F. 2012.Natural Frequency and Damping Estimation of an Offshore WindTurbine Structure. Proc. of the 20th Int. Offshore and PolarEngineering Conf., Rhodos, Greece. 300-307.

Damgaard M, Ibsen L.B, Andersen L.V., Andersen J.K.F. and AndersenP. 2012. Damping Estimation of Prototype Bucket Foundation forOffshore Wind Turbines Identified by Full Scale Testing. Proc. ofthe 5th Int. Operational Modal Analysis Conf. (IOMAC), Guimarães,Portugal, 300-307.

Devriendt C., Jordaens P.J., De Sitter G. and Guillaume P. 2012.Damping Estimation of an Offshore Wind Turbine on a MonopileFoundation. Proc. of the EWEA 2012 Conf., Copenhagen.

DNV 2011. Design of Offshore Wind Turbine Structures. Det NorskeVeritas AS.

Hansen M.H., Thomsen K., Fuglsang P., and Knudsen T. 2006. TwoMethods for Estimating Aeroelastic Damping of Operational WindTurbine Modes from Experiments. Wind Energy. 9, 179–191.

Hansson M., Hjort T.H. and Thaarup M. 2005. Data Report 0408Fredrikshavn Sand. Technical Report, Aalborg University.

Høgedal M. and Hald T. 2005. Scour Assessment and Design for Scourfor Monopile Foundations for Offshore Wind Turbines. CopenhagenOffshore Wind, Copenhagen.

IEC 2005. International Standard. Wind Turbines – Part 1: DesignRequirements. European Committee for Electrotechnical Standardi-zation, Brussels.

Nielsen S.R.K. 2004. Linear Vibration Theory. Aalborg TekniskeUniversitetsforlag, Denmark.

Ovesen N.K., Fuglsang L. and Bagge G. 2006. Lærebog i Geoteknik.Polyteknisk Forlag, Denmark.

Sumer B.M., Fredsøe J., and Christiansen N. 1992. Scour Around aVertical Pile in Waves. Journal of Waterway, Port, Coastal andOcean Engineering, ASCE. 117, 15-31.

Sørensen S.P.H., Ibsen, L.B. and Frigaard P. 2010. ExperimentalEvaluation of Backfill in Scour Holes around Offshore Monopiles.Proc. of the 2nd Int. Symposium on Frontiers in OffshoreGeotechnics, Perth, Australia.

Tarp-Johansen N.J., Andersen L., Christensen E.D., Mørch C., KallesøeB. and Frandsen S. 2009. Comparing Sources of Damping of Cross-Wind Motion. The European Offshore Wind Conference &Exhibition, Stockholm.

Tcherniak D., Chauhan S., Rosseth M., Font I., Basurko J. and SalgadoO. 2010. Output-Only Modal Analysis on Operating WindTurbines: Application to Simulated Data. European Wind EnergyConf., Warsaw, Poland.

Versteijlen W.G., Metrikine A.V., Hoving J.S., Smid E. and De VriesW.E. 2011. Estimation of the Vibration Decrement of an OffshoreWind Turbine Support Structure Caused by its Interaction with Soil.Proc. of the EWEA Offshore 2011 Conf., Amsterdam.

Øye S. 1996. FLEX 5 User Manual. Lyngby.

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Numerical investigation of dynamic embedment of offshore pipelines

Étude numérique de l’ancrage dynamique de conduites enterrées maritimes

Dutta S., Hawlader B. Memorial University, St. John’s, Canada.Phillips R. C-CORE, St. John’s, Canada.

ABSTRACT: Pipelines are one of the key components of offshore oil and gas development programs. Deep water pipelines are oftenlaid on the seabed and penetrate into soil a fraction of their diameter. High operating temperature and pressure generate axial stressthat could buckle the pipeline laterally. The embedment and formation of soil berm have a significant effect on lateral resistance. Theembedment of a pipeline depends on stress concentration at the touchdown point (TDP) and dynamic laying effects. In this study, large deformation finite element modelling of dynamic penetration of offshore pipeline is presented. The Coupled EulerainLagrangian (CEL) technique is used to develop finite element model. The pipe is first penetrated into the seabed followed by a small amplitude cyclic lateral motion. Results from the finite element models are compared with centrifuge test results. High plastic shearstrain is obtained around the pipeline during cyclic loading which causes significant pipe embedment. The shape of soil berm is different from that of monotonic pipe penetration.

RÉSUMÉ : Les conduites enterrées sont un des éléments clés des programmes de développement de pétrole et de gaz. Des conduites enterrées en eau profonde sont souvent mises sur le plancher océanique et pénètrent dans le sol sur une fraction de leur diamètre. La température et la pression de fonctionnement élevées génèrent une contrainte axiale qui peuvent déformer la conduite latéralement.L'ancrage et le sol encaissant ont un effet significatif sur la résistance latérale. L'enfouissement d'une conduite dépend des concentrations de contraintes et des effets dynamiques de la pose. Dans cette étude, une modélisation par éléments finis en grande déformation de la pénétration dynamique de la conduite est présentée. Une technique de type Eulérien Lagrangien (CEL) est utilisée pour développer le modèle éléments finis. Le tuyau est d'abord mis en place dans le fond marin puis subit un mouvement cyclique de faible amplitude latérale. Les résultats des modèles éléments finis sont comparés avec les résultats d’essais en centrifugeuses.D’importantes valeurs de la déformation plastique sont obtenues autour de la canalisation lors du chargement cyclique ce quinécessite un ancrage suffisant de la conduite. La forme du sol encaissant est différente de celle du tuyau mis en place statiquement.

KEYWORDS: pipelines, dynamic embedment, clay, large deformation analysis.

1 INTRODUCTION

As-laid pipelines are commonly used in deepwater. During installation the as-laid pipeline could be penetrated a fraction of its diameter into the seabed (Bruton et al. 2006), and a soil berm could be formed. The soil around the pipelines provides not only the thermal insulation and hydrodynamic stability to the pipe but also resistance to pipeline walking and lateral buckling during high operating temperature and pressure. Accurate assessment of as-laid pipe embedment is extremely difficult. Depending upon sea state, vessel conditions, pipe stiffness and soil conditions, the pipeline might experience both in-plane and out-of-plane cyclic motion during installation (Westgate et al. 2010, 2012), which causes dynamic embedment of the pipeline.

The penetration of a pipeline under static load can be obtained using bearing capacity theory, analytical solution or finite element techniques. In the current engineering practice, two additional factors are used to estimate the embedment of pipelines: (a) additional vertical force near the TDP (the point where the pipe first touches the soil) due to catenary effects and (b) dynamic lay effects. A number of methods have been proposed in the past to estimate these factors (Carneiro et al. 2010, Oliphant and Yun 2011). For example, Randolph and White (2008) proposed an empirical equation to calculate the touchdown lay factor (flay) using pipe submerged weight, bending rigidity, horizontal component of effective tension, lay angle, water depth and seabed stiffness. The embedment factor for dynamic lay effects (fdyn) varies between 2 and 10 (Lund 2000, Bruton et al. 2006). This wide range of variation in this factor makes the assessment of pipe embedment very difficult.

During installation, both vertical and lateral pipe motions can soften the seabed soil near the pipe. Soil softening/remolding together with water entrainment can reduce the undrained shear strength of soil. Field observation (Westgate et al. 2010) and physical modeling using geotechnical centrifuge (Cheuk and White 2011) show that the horizontal cyclic motion, although small amplitude, has a significant effect on pipe embedment.

The main purpose of this study is to conduct large deformation finite element (FE) analysis for dynamic events during the installation of pipeline. Coupled Eulerian Lagrangian (CEL) technique is adopted in the analysis using ABAQUS FE software. Four FE models are developed for two different soils: kaolin and high plasticity clays (plasticity index for kaolin is 34 and for high plastic clay is 100-130, Cheuk and White 2011). The results are compared with the centrifuge test results vailable in the literature. a

2 PROBLEM DEFINITION.

The problem considered in the present finite element (FE) modelling is shown in Fig.1. During laying, offshore pipelines usually penetrate vertically into the seabed due to its self-weight and catenary effect near the touchdown zone (TDZ). The vessel movement from wave loading could cause small amplitude cyclic motions in the x-direction. As the pipeline is under a vertical load (p0), the lateral movement in the x-direction could cause additional vertical penetration as shown by Stage-II and III in Fig. 1.

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Figure 1. Problem statement.

3 FINITE ELEMENT MODELLING.

ABAQUS 6.10 EF-1 is used in the present finite element analysis. As the embedment of pipe in the seabed is large deformation problem, the conventional finite element techniques in Lagrangian approach cannot simulate the complete process realistically as numerical difficulties are generally encountered for such large displacements. Therefore, in this study the Coupled Eulerian Lagrangian (CEL) technique currently available in ABAQUS FE software is used. In CEL, the soil flows through the fixed mesh without having any numerical issues. The FE modeling using CEL for pipe embedment into the seabed is presented by the authors previously (Dutta et al. 2012 a&b). A soil domain of 8m×3m ×0.04m (length × height × thickness) is used in this study. The soil is modelled as Eulerain elements and the pipe is modelled as Lagrangian elements. The 1.5 m void space above the soil is required to accommodate the displaced soil mass (Eulerian materials) during pipe displacement. Zero velocity boundary conditions are applied at all faces of the Eulerian domain to make sure that Eulerain materials are within the domain and cannot move outside. However, at the seabed-void interface, no boundary condition is provided so that the soil can flow to the void. That means, the bottom of the model is restrained from any vertical movement, while all the vertical faces are restrained from any lateral movement. The pipe is modeled as a rigid body. During penetration, especially in cyclic loading, the remolding of soil near the pipe could cause significant reduction in undrained shear strength. Smooth pipe/soil interface condition is used for the present analysis. Mesh sensitivity analysis is also performed and an optimum mesh size of 0.04m×0.04m is used (Dutta et al. 2012 a).

The loading is performed in three different stages. First, the geostatic conditions are applied to bring the seabed to in-situ condition. Second, the pipe is penetrated applying a vertical load (p) which is the combined effect of submerged unit weight of the pipe and laying effects. In the third step, 40 cycles of small amplitude (0.05D) lateral displacement are applied using displacement boundary conditions under the constant vertical load p0. Plastic shear strain develops near the pipe during penetration. In the present FE analyses the degradation of undrained shear strength as a function of plastic shear strain is adopted using the following model (Einav and Randolph 2005 Wang et al. 2009 and Zhou and Randolph 2009).

su=[rem+(1-rem)exp(-3/95)]su0 (1)

where trem S1 , St is the soil sensitivity, is the

accumulated equivalent plastic shear strain, su0 is the intact undrained shear strength of soil and ξ95 is the accumulated plastic shear strain at 95% undrained shear strength degradation. The variation of su0 with depth is shown in Fig. 1 and the von-Mises yield criteria is adopted.

In this study four cases are simulated and the results are compared with centrifuge test results of Cheuk and White (2008). Two tests (KC-04 & KC-05) are in kaolin clay and two

(HP-06 & HP-07) are in high plasticity clay. Table 1 shows the parameters used in the FE analyses. The vertical load p for initial static penetration and during cyclic motion are also shown in Table 2. T able 1. Parameters for finite element modelling.

PipePipe diameter, D (mm) Lateral displacement during cyclic motion

800

± 0.05D

Soil Properties KaolinClay

HighPlasticity

Clay Undrained modulus of elasticity, Eu 500su 500su Poisson’s ratio, u 0.495 0.495Undrained shear strength at mudline, sum (kPa) 0.75 0.40 Gradient of shear strength increase, k (kPa/m) 1.6 2.5 Submerged unit weight of soil, (kN/m3) Remoulded soil sensitivity, St

Accumulate absolute plastic shear strain for 95% degradation of soil strength, 95

6.0 4.0 10

3.0 1.7 10

Table 2. Centrifuge test conditions (Cheuk and White 2011). KC-04 KC-05 HP-06 HP-07Pipe vertical load, p (kN/m) 1.17 2.23 1.47 2.61 Initial static embedment, win/D 0.08 0.12 0.10 0.22 Pipe vertical load at cyclic motion, p0 (kN/m)

1.13 2.17 1.43 2.52

4 RESULTS.

The pipe was initially penetrated under a static vertical load p. The initial static embedment (win) for this load is shown in Table 2. After initial penetration a small amplitude cyclic lateral load is applied (e.g. Fig. 2 for KC-05, u = pipe lateral displacement) to simulate the first 40 cycles (Stage-I) of centrifuge tests. The normalized lateral resistance for KC-05, where su0(i) in the horizontal axis is the intact undrained shear strength at pipe invert, is shown in Fig. 3(a) and comapred with centrifuge test results Fig.3(b). The lateral resistance is slightly higher than that obtained in centrifuge test. This might be due to the limitation of the soil shear strength degradation model (Eq. 1). It is very difficult to measure and model the behaviour of soil near the pipeline under cyclic loading. However, using this simplified model (Eq. 1) in ABAQUS CEL the lateral resistance during cyclic movement is reasonably simulated.

Figures 4(a),4(b) and 4(c) show the lateral resistance for other three simulations. As shown, the shape of the lateral resistance plot is different, which mainly depends on soil shear strength profile, shear strength degradation, sensitivity of soil, and applied vertical load. The depth of the invert of the pipe normalized by pipe diameter (D) with number of load cycle is shown in Fig. 5 for comparison the centrifuge test. the present FE model reasonably simulates the embedment

of the pipe with the soil parameters listed in Table 1. Figure 5(a) shows that the depth of embedment does not increase

x

Z

x

z

D

p0

P

B

ipe

su0 = sum+kz

w

erm

SS

Stage-I

tage-II tage:-III

s

k

um

w/D

Figure 2. Pipe embedment during lateral motions

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significantly after 20-30 load cycles for kaolin clay. However, the pattern is somehow different for high plastic clay as shown in Fig. 5(b) where the pipes continue to penetrate even after 20-30 load cycles.

0

0.1

0.2

0.3

0.4

0.5

0.6

-2 -1 0 1 2H/su0(i) D

KC-05 (CEL)

0

0.1

0.2

0.3

0.4

-2 -1 0 1 2

w/D

H/su0(i) D

KC-04 (CEL)

0

0.1

0.2

0.3

0.4

0.5

-2 -1 0 1 2

w/D

H/su0(i) D

HP-06 (CEL)

Figure 3.Test KC-05 (a) present study (b) centrifuge test.

From field experience in deepwater pipeline projects in West Africa, Casola et al. (2011) suggested that the embedment of pipelines is more than D/3 and the lateral cyclic load has little effect on pipe embedment. Their observed behaviour might be applicable if su0 and the gradient (k) of su0 is high as they found in these projects. However, for the cases analyzed in the present study, the cyclic loading has a significant effect even after embedment of D/3 as shown in Fig. 5(b).

Figure 4. FE results (a) KC-04 (b) HP-06 and (d) HP-07.

5 DYNAMIC EMBEDMENT

In the current engineering practice the effects of laying and dynamic embedment are assessed separately. The lay effect on vertical load (p) is obtained by multiplying the submerged unit weight of the pipe by an empirical lay factor (flay). The monotonic embedment (wmon) for this load p is calculated using the bearing capacity theory. The effect of small amplitude cyclic lateral motion is incorporated using another empirical factor known as dynamic embedment factor (fdyn). Finally, the total embedment (wf) is calculated as wf = fdyn wmon. In some projects

(e.g. Oliphant and Yun 2011) a combined empirical factor (=flayfdyn) is also used that accounts for both laying and dynamic effects. Table 3 shows the calculated values of fdyn for the four tests simulated in this study. Analyzing field data of a 200 km offshore pipeline in shallow to deep water, Oliphant and Yun (2011) showed that an average value of fdyn of approximately 7 could be used for estimation of pipeline embedment. Lund (2000) suggested that the value of fdyn in the field varies between 2 and 10.

q for two

with horizontal cyclic motions (a) Kaolin (b) High plasticity clay.

(a) (b) KC-05(Centrifuge)

Stag

e-I 6 EQUIVALENT PLASTIC STRAIN AND BERM SHAPE.

During penetration the soil around the pipeline is softened as a function of plastic shear strain as shown in Eq. 1. The

w/D

e uivalent plastic shear strain at the end of penetration

0

0.1

0.2

0.3

0.4

0.5

0 5 10 15 20 25 30 35 40

w/D

No of Cycles

Figure 5.Pipeline embedment clay cases (KC-04 and KC-05) are shown in Figs. 6(a) & 6(b). A significant plastic strain (>500%) is developed near the pipe. The white broken lines in Figs. 6(a) & 6(b) show the boundary above which the equivalent plastic shear strain is greater than 95 (=10). That means 95% degradation of undrained shear strength occurred in the soil above this line (see Eq. 1). In other words, the undrained shear strength of the soil in the zone above this line is almost near the remoulded undrained shear strength. In order to show the effects of lateral cyclic loading on penetration, an analysis is performed for monotonic penetration as shown in Fig. 6(c). The geometry and soil property used in this analysis is same as KC-05 in Fig. 6(b), except the pipe moved monotonically downward to the depth of final embedment in KC-05 using a displacement boundary condition. The shear strain and berm formation for monotonic penetration is shown in Fig. 6(c). the equivalent plastic shear strain near the pipe in monotonic loading is significantly lower (Fig. 6c) than that obtain in lateral cyclic loading (Figs. 6a&b). The maximum plastic shear strain developed near the pipe in cyclic loading is almost 5 times higher than that of monotonic loading. Also the maximum equivalent plastic shear strain is less than 95, which means that the shear strenth reduction due to softening is less

0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

-2 -1 0 1 2H/su0(i) D

HP-07 (CEL)(c) (b)

0.6

KC-04(Cheuk and White 2008)(a) KC-05 (Cheuk and White 2008)KC-04 (Present study)KC-05 (Present study)

(a)

0

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0 5 10 15 20 25 30 35 40

w/D

No of Cycles

0.8

HP-06 (Cheuk and White 2008)(b) HP-07 (Cheuk and White 2008)

HP-06 (Present study)

HP-07 (Present study)

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than 95%. The softening has a significant effect on the shape of the berm and soil movement around the pipe. the highly softened displaced soil mass formed flat berms on the top of the seabed extended over a large distance in cyclic loading. In addition, the soil is in contact with the pipe almost up to the top of the berm. However, for monotonic loading the displaced soil mass formed a berm mainly near the pipe and the berm height is more than that of in cyclic loading. That means, the soil deformation in monotonic and cyclic loading is significantly dif

ning behaviour of soil will provide more accurate results.

able 3. Dynamic embedment factoK K H

ferent. In a parametric study Dutta et al. (2012 b) showed that the

effect of softening on the vertical penetration resistance in monotonic loading is not very significant, which is because of less plastic strain developed near the pipe. However, for cyclic loading huge plastic strain is developed in a zone near the pipe which causes significant reduction in undrained shear strength. That means, the zone of considerable softening is higher in cyclic loading. As in offshore the small amplitude lateral cyclic loading near the touchdown zone is commonly encountered from the motion of the vessel, the analyses for cyclic motion with strain softe

T r, fdyn. C-04 C-05 HP-06 P-07Initial static embedment (win/D) 0.08 0.12 0.10 0.22 Final embedment (w/D) 0.34 0.50 0.48 0.77 fdyn 4.25 4.16 4.8 3.5 Figure 6. Equivalent plastic strain.

ated. The following conclusions can be drawn from this

BAQUS FE software can simulate the pipeline

oading is

ipe in

c loading. Forty cycles of small amplitude lateral loading increased the

initial static embedment. The cycles.

The work has been funded by C-CORE, MITACS and NSERC h is greatly acknowledged.

9

Bru

Che

Geotechncial Journal (accepted). Zhou, H. and Randolph, M. F. (2009). Numerical investigations into

cycling of full-flow penetrometers in soft clay. Géotechnique, 59 (10), 801-812.

The shape of the berm depends on type of loading; spread over a large area in cyclic and mounted near the pmonotonic loading. The softening of soil in a zone near the pipe is significantly higher in cyclic loading compared to monotoni

embedment by a factor of 4-5 ofembedment is higher in initial loading

8 ACKNOWLEDGEMENTS

Discovery grant whic

REFERENCES

ton, D., White, D., Cheuk, C. And Bolton, M. (2006). Pipe/soil interaction behaviour during lateral buckling, including large amplitude cyclic displacement tests by the Safebuck JIP. Proc. Offshore Technology Conference, Houston, Texas, USA. OTC 17944.

neiro, D. Gouveia, J. and ParCar rilha, R. 2010. Feedback analysis of pipeline embedment over as-laid survey results. Proc. Int. Conf. on Offshore Mechanics and Arctic Engineering, Shanghai, China. OMAE2010-20410.

ola, F., El-chayeb, A., Greco, S. and CarlucCas ci, A. 2011. Characterization of pipe soil interaction and influence on HP/HT pipeline design. Proc. Int. Conf. on Offshore and Polar Engineering Conference, Hawaii,USA,pg:111-121.

Cheuk, Y.C. and White, J.D. 2008. Centrifuge modelling of pipe penetration due to dynamic lay effects. Proc. Int. Conf. on Offshore Mechanics and Arctic Engineering, Estoril, Portugal. OMAE2008-57923. uk, Y.C. and White, J.D. 2011. Modelling the dynamic embedment of seabed pipelines. Géotechnique 61 (1), 39-57.

ta, S., Hawlader, B. and Phillips, R. 2012a. Finite elemDut ent modeling of vertical penetration of offshore pipelines using Coupled Eulerian Lagrangian approach. Proc. Int. Conf. on Offshore and Polar Engineering Conference, Rhodes,Greece,pg:343-348.

ta, S., Hawlader, B

(a)KC-04

Dut . and Phillips, R. 2012b. Strain softening and rate effects on soil shear strength in modeling of vertical penetration of offshore pipelines. Proc. Int. Pipeline Conference,Alberta, Canada. IPC2012-90233.

Einav, I. and Randolph, F.M. 2005. Combining upper bound and strain path methods for evaluating penetration resistance. Int. J. Numer. Meth. Engng., 63:1991-2016.

(b)KC-05 Lund, K.H. 2000. Effect of increase in pipeline soil penetration from installation. Proc. Int. Conf. on Offshore Mechanics and Arctic Engineering, New orleans, USA. OMAE2000-PIPE5047.

Oliphant, J. and Yun, J.G. 2011. Pipeline embedment prediction using as-laid data. Proc. Int. Conf. on Offshore Mechanics and Arctic Engineering, Rotterdam, The Netherlands. OMAE2011-50095.

Randolph F.M. and White J.D. 2008. Pipeline embedment in deep water: Processes and quantitative assesment. Proc. Offshore Technology Conference, Houston, Texas, USA. OTC 19128.

ng, D., White, J.D. and Rand

7 CONCLUSION

Large deformation finite element analyses are conducted to assess the embedment of as-laid offshore pipelines in clay. The effects of small amplitude cyclic lateral loading are investig

Wa olph, F.M. 2009. Numerical simulations of dynamic embedment during pipe laying on soft clay. Proc. Int. Conf. on Offshore Mechanics and Arctic Engineering, Hawaii, USA. OMAE2009-79199.

stgate, J.Z., White, J.D. and Randolp

(c) Monotonic loading

We h, F.M. 2010. Pipeline laying and embedment in soft fine-grained soils: Field observations and numerical simulations. Proc. Offshore Technology Conference, Houston, Texas, USA. OTC 20407.

Westgate, J.Z., White, J.D. and Randolph, F.M. 2012. Modelling the embedment process during offshore pipe laying on fine-grained soils. Canadian

study. The Coupled Eulerian Lagrangian (CEL) method currently available in Aembedment. The plastic shear strain near the pipeline in cyclic lsignificantly higher than that of in monotonic loading.

2351

Post Cyclic Behaviour of Singapore Marine Clay

Le comportement post-cyclique de l’argile marine de Singapour

Ho J., Goh S.H., Lee F.H. National University of Singapore

ABSTRACT: In this paper, the post-cyclic behaviour of remoulded Singapore Marine Clay is examined. Cyclic triaxial tests, followed by monotonic loading to failure, were performed on normally consolidated specimens (38mm diameter by 76mm height)within a cyclic strain range of approximately 0.7% to 1.4%. Results herein reveal that the effective stress paths under post-cyclic monotonic loading may take on different forms depending on the mean effective stress state of the specimen at the end of the cyclicloading phase. By normalizing the mean effective stress (p’) against the effective consolidation pressure (pc’) of the specimen, the effective stress paths during the post-cyclic monotonic loading may be approximately grouped into three different regimes, accordingto the normalized mean stress (p’/pc’). Within each normalized mean stress regime, the monotonic soil response is independent of the effective consolidation pressure, the cyclic strain amplitude and number of cycles applied during cyclic loading. The results suggestthat the normalized mean stress after cyclic loading may be an important parameter in determining the subsequent stress path underundrained monotonic loading to failure.

RÉSUMÉ : Dans ce papier, le comportement post-cyclique de l’argile marine remaniée de Singapour est étudié. Des essais triaxiauxcycliques suivis de chargement monotone jusqu’à la rupture ont été effectués pour des déformations cycliques comprises entre 0,7% et 1,4% sur des échantillons normalement consolidés. Les résultats démontrent que les chemins de contrainte durant le chargementmonotone post-cyclique peuvent être de formes différentes et dépendent de l’état de contrainte moyenne effective de l’échantillon à lafin de la phase du chargement cyclique. En normalisant la contrainte moyenne effective (p’) par la contrainte de consolidation (pc’),les chemins de contrainte pendant le chargement monotone post-cyclique peuvent être regroupés en trois différents groupes selon la contrainte moyenne normalisée (p’/pc’). Dans tous les cas, la réponse monotone du sol est indépendante de la contrainte effective de consolidation, de l’amplitude de la déformation cyclique et du nombre de cycles appliqué pendant le chargement cyclique. Les résultats suggèrent que la contrainte moyenne normalisée après le chargement cyclique pourrait être un paramètre important pourdéterminer, sous chargement monotone non-drainé, les chemins de contrainte jusqu’à la rupture .

KEYWORDS: Post-cyclic clay behaviour, Mean effective stress, Cyclic stress reversal

1 INTRODUCTION

Previous studies have shown that, during undrained compression loading, the effective stress path of a normally consolidated clay after cyclic loading is similar to the effective stress path of an overconsolidated clay (Hyde and Ward 1985, Matsui et al. 1992, Yasuhara et al. 1992). Researchers have adopted different frameworks in the analysis of the post-cyclic undrained shear strength of clays. One common approach is to estimate the undrained shear strength based on the maximum shear strain during the applied cyclic loading (Thiers and Seed 1969, Sangrey and France 1980). Another method of analysis is to relate the post-cyclic undrained shear strength with the apparent overconsolidation ratio induced by the cyclic loading (Matsui et al. 1992, Yasuhara et al. 1992). Apart from the post-cyclic undrained shear strength, this induced apparent overconsolidation ratio from cyclic loading was also used to determine how the subsequent monotonic effective stress path approaches the critical state line (Yasuhara et al. 1992).

However, one possible limitation in past studies is the relatively fast rates of cyclic loading used, which typically ranges from 0.5Hz to 1 Hz. At such loading rates, it is uncertain if excess pore pressure in the clay specimens will be able to dissipate. For this reason, the reliability of pore pressure measurements during cyclic loading phase may be doubtful. Fast loading rates on normally consolidated clays during undrained triaxial tests prevents equilibration of excess pore pressure within test specimens, which results in a higher pore pressure within the middle one-third portion of the specimen

(Wood 1982). This can lead to errors in effective stress calculations. Few attempts had been made to overcome the issue of unequalized pore pressures during cyclic loading. For example, Matsui et al (1992) the specimen to stand in an undrained state for 1 hour after cyclic loading and prior to post-cyclic monotonic loading to allow for equalization of pore pressures. On the other hand, Diaz-Rodriguez et al (2000) used a longer equalization period of 12 hours. Another approach is to allow drainage before post-cyclic loading to allow for pore pressures accumulated during cyclic loading to be dissipated (Yasuhara et al. 1992). The drawback of this approach is that it alters the pore pressures within specimens, leading to discontinuities in effective stress paths between the cyclic loading and post-cyclic loading phases. Intuitively, the effective stress response of a clay undergoing cyclic loading should be indicative of its post-cyclic behavior if post-cyclic monotonic loading is conducted immediately after cyclic loading. Due to possible errors effective stress measurements and discontinuities between cyclic and post-cyclic effective stress paths, a direct comparison between the cyclic and post-cyclic behavior of clays has hitherto not been possible. This objective of the study reported herein is to re-visit the issue of post-cyclic behaviour of soft clay, while ensuring adequate equilibration of excess pore pressure. 2 CYCLIC AND POST CYCLIC TESTING PROGRAM

A series of two way strain-controlled undrained cyclic triaxial tests were performed on remoulded specimens (38mm diameter by 76mm height) of normally consolidated Singapore Upper

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Marine Clay, the standard properties of which have been reported by Tan (1983). After cyclic loading, the specimens were immediately subjected to standard consolidated undrained triaxial monotonic loading to failure. According to Ho et al. (2012), when undrained cyclic triaxial tests on clays are conducted at a sufficiently slow rate for pore pressure equilibration, intrinsic strain rate effects on pore pressure measurements, effective stress paths and stress-strain relationships are negligible. Since the focus of this study is not on strain rate effects, all tests were conducted at relatively slow rates. Both mid-plane and base pore pressure transducers were used and pore pressure equilibration is considered to be achieved when both transducers produce similar excess pore pressure measurements. All cyclic and post-cyclic triaxial tests were performed using the GDS Enterprise Level Dynamic Triaxial Testing System. Table 1 shows the experimental matrix. Experimental data presented in this study was recorded at 2-second intervals.

Table 1. Experimental Matrix.

Cyclic Amplitude No

Effective Consolidation Pressure, pc’

(kPa) Amplitude

(mm) Strain(%)

Period (min)

No. of Cycles

1 - - 0 2 53 104 155 206

501 ≈1.4 10

1007 - - 0 8 29 3

10 411 512 613 2014 3015

1001 ≈1.4 14

10016 - - 017 218 319 420 521 622 1023 3024

2001 ≈1.4 60

100

3 EXPERIMENTAL RESULTS AND DISCUSSION

3.1 Cyclic Loading

Normalized stress plots and stress-strain relationships during the cyclic tests are summarized in Figure 1. The stress path parameters, i.e. deviator stress (q) and mean effective stress (p’), are normalized against the effective consolidation pressure (pc’) for easy comparison between specimens subjected to different consolidation pressures. Post-cyclic effective stress paths are included in Figure 1. The critical state line is also plotted based on data from monotonic triaxial compression tests where the effective angle of friction for Singapore Upper Marine Clay is found to be 25.4 degrees. The initial yield locus and the state boundary surface are assumed to be elliptical.

For Singapore Upper Marine Clay specimens subjected to the same effective confining pressure, the stress paths and stress-strain relationships shown on Figure 1 for different number of applied cycles are similar, reflecting consistency among the specimens. Due to positive excess pore water

pressure generated during cyclic loading, the mean effective stress generally decreases during the reloading phase of each cycle. However, after a certain number of load cycles, the mean effective stress is observed to increase at the later part of the loading, just before the maximum deviator stress is reach, as illustrated in Figure 2. The turning point marking this change in mean effective stress is hereby termed as “stress reversal” point. These stress reversal points correspond to a decrease in shear-induced excess pore water pressure, which would seem to imply dilative behavior. As the stress reversal points appear after the normalized mean effective stress decreases beyond a certain value, post-cyclic monotonic tests are conducted after different number of load cycles to investigate the factors governing the onset of this stress reversal behaviour.

3.2 Post-cyclic Loading

Normalized stress plots and stress-strain relationships during the post-cyclic monotonic tests are summarized in Figure 3, for the tests listed in Table 1. From Figure 3, the form of the effective stress paths under post-cyclic monotonic loading depends on the normalized mean effective stress state of the specimen at the start of the post-cyclic loading phase. These post-cyclic effective stress paths may be approximately grouped into three different regimes, according to the normalized mean effective stress.

When the normalized mean effective stress state of the clay specimen at the start of post-cyclic monotonic loading is greater than 0.6, stress reversal is generally absent and post-cyclic shearing shows either a decrease or no change in mean effective stress. This is akin to that of lightly over-consolidated and normally consolidated clays which tend to increase in density when sheared.

On the other hand, when the normalized mean effective stress state of the clay specimen at the start of post-cyclic monotonic loading falls below 0.5, stress reversal becomes evident and the effective stress path becomes similar to that of heavily over-consolidated clays.

Within the range of 0.5 to 0.6, the effective stress path is approximately vertical indicating that this is a boundary zone between occurrence or otherwise, of stress reversal.

Unlike the effective stress paths, the normalized stress-strain relationships are relatively similar. The post-cyclic undrained strength is almost equal to that without cyclic loading. This means that the undrained shear strength of Singapore Upper Marine Clay is not significantly influenced by cyclic loading. This observation agrees with the findings of Yasuhara et al. (1992).

3.3 Effect of Cyclic Strain Amplitude

In this section, the results of four additional tests conducted at a lower strain amplitude are presented; the aim being to investigate whether strain amplitude has any effect on the stress reversal. Table 2 summarizes the four additional tests conducted.

Table 2. Additional Cyclic and Post-cyclic Tests.

Cyclic Amplitude No

Effective Consolidation Pressure, pc’

(kPa) Amplitude

(mm) Strain(%)

Period (min)

No. of Cycles

1 102 153 204

100 0.5 ≈0.7 14

110

Figure 4 presents the normalized stress plots and stress-strain relationships obtained from the additional tests. As the applied cyclic strain amplitude for these four tests has reduced by half to 0.7%, the number of load cycles required to reach the same mean effective stress state as previous tests with strain amplitude of 1.4% has increased proportionally. However, as Figure 4 shows, the post-cyclic effective stress paths of these four tests can still be categorized under the three normalized mean effective stress regimes previously discussed. The boundaries of these three regimes remain the same despite the

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increase in number of load cycles and reduction in applied cyclic strain amplitude. This implies that the form of the effective stress paths of Singapore Upper Marine Clay under post-cyclic monotonic loading is dependent primarily on the mean effective stress state at the end of the cyclic loading phase.

4 CONCLUSIONS

The effective stress paths of Singapore Upper Marine Clay under post-cyclic monotonic loading takes on three different forms depending on the mean effective stress state of the specimen at the end of the cyclic loading phase. Although it is commonly accepted that the undrained post-cyclic effective stress path of a normally consolidated clay is similar to the effective stress path of an overconsolidated clay (Matsui and Abe 1981, Hyde and Ward 1985, Yasuhara et al. 1992), this study shows that this condition holds only when the normalized mean effective stress state of the clay specimen at the start of

post-cyclic monotonic loading falls below 0.5. When the normalized mean effective stress state of the clay specimen at the start of post-cyclic monotonic loading is greater than 0.6, the effective stress path of the specimen still follows that of lightly over-consolidated and normally consolidated clays. Within each of these normalized mean effective stress regimes, the post-cyclic clay behavior is governed by the normalized mean effective stress after cyclic loading and independent of the effective consolidation pressure, the cyclic strain amplitude and number of cycles applied during cyclic loading.

Furthermore, stress reversal points observed during cyclic loading phase becomes evident when mean effective stress state of the clay specimen at the start of post-cyclic monotonic loading falls below 0.5. Thus, the occurrence of stress reversal during cyclic loading is indicative of dilative behavior and post-cyclic monotonic loading in this regime exhibits effective stress path of an overconsolidated clay.

Figure 1. Effective Stress Paths and Stress-Strain Relationships for (a) pc’ = 50kPa, (b) pc’ = 100kPa, and (c) pc’ = 200kPa.

p’ / pc’ increases

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Figure 2. Stress Reversal Points.

Figure 3. Post-cyclic Effective Stress Paths and Stress-Strain Relationships.

Figure 4. Post-cyclic Effective Stress Paths and Stress-Strain Relationships for Post-Cyclic Tests at 0.7% Cyclic Amplitude.

5 ACKNOWLEDGEMENTS

The authors are grateful to National University of Singapore for the provision of laboratory facilities, without which the research will not be possible. In particular, the main author will like to thank National University of Singapore for the research opportunity given through the award of research scholarship.

6 REFERENCES

Diaz-Rodriguez J.A., Moreno P. and Salinas G. 2000. Undrained shear behavior of mexico city sediments during and after cyclic loading. Proc. 12th World Conference on Earthquake Engineering, 1652-1660.

Ho J.H., Kho Y., Goh S.H. and Lee F.H. 2012. Cyclic triaxial testing of soft clays. Proc. 25th KKCNN Symposium on Civil Engineering, 347-350.

Hyde A.F.L. and Ward S.J. 1985. A pore pressure and stability model for a silty clay under repeated loading. Géotechnique 35 (2), 113-125.

Matsui T., Bahr M.A. and Abe N. 1992. Estimation of shear characteristics degradation and stress-strain relationship of saturated clays after cyclic loading. Soils and foundations 32 (1), 161-172.

Sangrey D.A. and France J.W. 1980. Peak strength of clay soils after a repeated loading history. Proc. International Symposium on Soils Under Cyclic and Transient Loading, 421-430.

Tan S.L. 1983. Geotechnical properties and laboratory testing of soft soils in Singapore. Proc. 1st Internation Seminar on Constrtuction Problems in Soft Soils, 1-47.

Thiers G.R. and Seed H.B. 1969. Strength and stress-strain characteristics of clays subjected to seismic loading conditions. Proc. ASTM STP 450, 3-56.

Wood, D.M. 1982. Laboratory investigations of the behaviour of soils under cyclic loading: a review. In: Soil mechanics – Transient and cyclic Loads, eds. Pande, G.N. and Zienkiewicz, O.C. John Wiley & Sons Ltd, Chichester.

Yasuhara K., Hirao K. and Hyde A.F.L. 1992. Effects of cyclic loading on undrained strength and compressibility of clay. Soils and Foundations 32 (1), 100-116.

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Centrifuge test and numerical modeling for a suction bucket monopod foundation

Essai en centrifugeuse et la modélisation numérique d'une fondation de type : caisson à succion

Kim D.J., Youn J.U., Jee S.H., Choi J. Hyundai Engineering and Construction, Seoul, Korea

Choo Y.W., Kim S., Kim J.H., Kim D.S. Department of Civil and Environmental Engineering, KAIST, Daejeon, Korea

Lee J.S. Department of Civil and Environmental Engineering, Wonkwang University, Iksan, Korea

ABSTRACT: A centrifuge load test for a preliminary design of a monopod suction bucket foundation was performed. The target site was the Yellow Sea of Korea and the prototype foundation was a steel monopod caisson with a diameter of 15.5m for a 3MW turbine. The seabed conditions comprised of a dense silty sand above layers of sandy silt were reproduced to a model soil profileusing soil samples collected at nearby seashores. Horizontal load and overturning moment were applied and monitored in the test, with vertical load being simulated by self-weight of the bucket model. Series of numerical analysis were performed in order to validate test conditions and compare the effects of soil parameters.

RÉSUMÉ : Un test de chargement en centrifugeuse pour étudier le design préliminaire d'une fondation de type « caisson à succion » a été réalisé. Le site cible était la mer Jaune de Corée et la fondation prototype était un caisson unique en acier caisson de15.5 m de diamètre pour une éolienne de 3 MW. La stratigraphie du fond marin, sable limoneux dense et limon sableux, a été reproduite pour faire un profil de sol modèle en utilisant des échantillons de sol prélevés sur le rivage à proximité. Un chargementhorizontal et un moment de renversement ont été appliqués et contrôlés pendant l'essai, le chargement vertical était simulé en utilisant le poids propre du modèle. La modélisation numérique a été réalisée afin de valider les conditions d'essai et de comparerles effets du choix des paramètres de sol.

KEYWORDS: Suction bucket foundation, Monopod bucket foundation, Offshore wind, Centrifuge Test, Numerical Modeling MOTS-CLÉS : Caisson à succion, Caisson de fondation, Éolienne Offshore, centrifugeuse, modélisation numérique

1 INTRODUCTION

Suciton bucket (also termed as suction caisson or suction pile) has been considered as a viable alternative to conventional foundations for offshore wind turbines, becuase it has features appropriate for installing large foundations in offshore environment with minimal environmental problems (Byrne and Houlsby 2003, Houlsby et al. 2005, Villalobos 2006, LeBlanc et al. 2009, Hung and Kim 2012, Oh et al. 2012). In Korea, major offshore wind farm projects are planned in the Yellow Sea near the south western coast of Korea. The soil profiles are mainly composed of layers of silty sand and sandy silt.

A preliminary design was performed for field testing of suction bucket foundations, and centrifuge load tests were performed to verify and compare alternative designs. In this paper, a centrifuge test for a steel monopod with a diameter of 15.5 m and a length of 10.5 m is described. Expected horizontal load combined with moment load was applied in the test.

Numerical analyses were performed to validate the centrifuge test model conditions such as model weight and soil boundary distance. In addition, the effects of soil parameters such as elastic modulus, internal friction angle, dilation angle, cohesion and wall interface friction angle, on foundation behaviour were evaluated.

2 CENTRIFUGE MODELING

A centrifuge test was performed with a geotechnical centrifuge at KAIST (Korea Advanced Institute of Science and Technology) in South Korea. It has a maximum capacity of 240 g-ton and 5 m radius (Kim et al. 2012). Detailed description of the centrifuge test for this study can be found in Choo et al. (2012) and Kim et al (2013). The procedures and results are briefly described here.

The soil conditions at the target site were replicated in the model soil container for the centrifuge test. Natural soil samples collected at the Western coastal areas near the target site were used after verifying that the properties of model materials were comparable the soil samples from the target site (Figure 2). The model profile was formed in two layers of dense silty sand and medium dense sandy silt up to the depth of 32 m, which was about two times the diameter of the foundation.

(a) Silty Sand (SM) Layer

(b) Sandy Silt (ML) Layer

Figure 1. Comparison of grain size distributions between target site samples and model soil

A 1/70 scaled model was used for the test. Horizontal load

by a displacement controlled actuator was applied and

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monitored at the model tower which was located at 33.0 m height from the foundation top in prototype scale. The horizontal displacement of the tower was measured at multiple points so as to calculate the horizontal displacement and rotation of the foundation.

The load – displacement curve of the test are shown in figure 2. The load is presented in moment, which is the horizontal load multiplied by the vertical eccentricity of the load from the foundation top. The displacement is shown in terms of the rotation of the foundation. Gradual decrease in the slope was observed and the method by Villalobos (2006) was used to define the yield load, which was 198 MN-m.

Figure 2. Moment – rotation angle curve of the centrifuge test

The model and nearby soil after the test are shown in Figure 3. Tilting of the foundation by the horizontal and moment load induced several mm of heave in the passive side and 20 to 30 mm of subsidence behind the bucket. Positions of the model before and after the load test are shown in Figure 4.

Figure 3. Model and nearby soil after the test

Figure 4. Comparison between positions before and after the test

3 NUMERICAL ANALYSIS

3.1 Model setup and analysis procedures

Numerical modeling in this study was performed using FLAC 3D V 5.0 based on the finite-difference method and explicit scheme (Itasca, 2012). The numerical model was modified from a model used in Kim et al. (2013) and detailed descriptions are given for modeling and analysis procedures.

Soil elements were modeled by Mohr-Coulomb failure criterion with linear elasticity up to plastic yield and the bucket body and tower parts were modeled by linear elastic solid elements. In order to represent the load conditions, a solid circular tower was additionally modeled on top of the bucket top lid and horizontal displacement was applied on the top face of the tower. Half section model mesh and boundary conditions were used for the analysis because of the symmetry of the foundations and load conditions. Approximately 4800 elements were used in the model. Actual steel deformation properties were used in the analysis (E = 200 GPa, ν = 0.30). The mesh for the analysis is shown in Figure 5.

Figure 5. Mesh for numerical model (Bucket body shown in magnified scale)

The base properties for the model are shown in Table 1. The submerged unit weight of the steel used for the bucket body was modified from actual value, because the weight of the centrifuge model bucket was increased by the connection between the bucket body and the vertical rod.

Table 1. Base properties for numerical analysis

ParametersItems

Bucket SM Layer ML Layer

Submerged unit weight (γsub, kN/m3)

75.9 9.50 8.60

Elastic modulus (E, MPa)

200,000 10 10

Poisson ratio (ν)

0.3 0.3 0.3

Internal friction angle (φ)

- 33.7 34.5

Dilation angle (ψ)

- 11.7 0

Cohesion (c, kPa)

- 16.1 5.2

Friction angle between bucket wall and soil (δ, )

- 22.5 -

Coefficient of earth pressure at rest (K0)

- 0.5 0.5

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Bottom face nodes were fixed in vertical displacement, and side face nodes were fixed in horizontal displacement. Coulomb criterion interface elements were applied in contacting faces between the bucket body and soil in order to model sliding and separation behaviour. Shear and normal stiffness values were set to 200 MPa/m which was larger than ten times the elastic modulus of surrounding soil (Itasca, 2005).

The analysis was run in three stages. The first stage simulated the initial K0 soil condition. The second stage simulated the installation of the bucket in the soil. The third stage was the loading stage where the top of the loading tower was horizontally moved in every step and unbalanced forces were calculated as the resistance of the foundation.

The ramping algorithm was used for the loading velocity control, in which the loading velocity was linearly increased with step to ud,max per step (1�10-6 m/step in this study) till prescribed steps were run and kept constant afterwards (Itasca, 2012).

3.2 Analyses and results

Cases considered in this study are summarized in Table 2 and a plot of displacement contour for C2 case is shown in Figure 6. Table 2. Analysis cases Items Values A1. Reference case Parameters in Table 1 A2. Bucket weight and vertical load

Bucket weight 2220 kN Vertical load 5750 kN

B1. Horizontal boundary distance from model center (5D for reference case)

2D

C1, C2. Elastic modulus of SM layer (E, MPa)

20, 5.0

D1, D2. Internal friction angle (φ)

38.7, 28.7

E1. Dilation angle (ψ) 3.7 F1. Cohesion (c, kPa) 0.1 G1. C2 + F1

Figure 6. Contour of displacement for case C2 The prototype of the centrifuge test was modeled with larger

thickness in the wall and the top plate than the preliminary design due to limitations in fabrication. The vertical rod and the connecting part between the bucket body and the rod were designed to have sufficient stiffness and strength for the centrifuge. These resulted in a heavier prototype and vertical load than the target structure in the preliminary design. Therefore, the effect of heavier structure weight was analyzed in the numerical analysis. The load – displacement curve is shown in Figure 7. Slight decrease in the resistance was observed for the reduced weight and vertical load, after around 0.003 ~ 0.005D.

Figure 7. Load – displacement curves for different foundation weight and vertical load

The centrifuge model soil container had a radius of 447.5 mm, which was about two times the diameter of the model foundation. The results between 2D and 5D horizontal boundary distances are compared in Figure 8. The difference was negligible between the horizontal boundary distances considered.

Figure 8. Load – displacement curves for different horizontal boundary distances

Different elastic moduli resulted in a noticeable variation in

the slopes of the curves (Figure 9). Therefore, proper estimation of the elastic modulus and application in the numerical model are thought be important for the load – displacement behaviour in the conditions of this study.

Figure 9. Load – displacement curves for different elastic moduli of the silty sand layer

The effect of variations in the internal friction angle and

dilation angle of the silty sand layer was considered (Figure 10). Slight changes in slopes were observed after around 0.005 ~ 0.01D, but they were found to be relatively small in the displacement range considered in this study.

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Figure 10. Load – displacement curves for different internal friction angles and dilation angles of the silty sand layer

The resistance was affected by the cohesion of the surrounding soil (Figure 11).

Figure 11. Load – displacement curves for different cohesions of the silty sand layer

Figure 12 shows the result when the elastic modulus and

cohesion were decreased from the reference values. The curve is closer to the centrifuge test result than others. However, this does not mean that this set of parameters are the actual properties, but provides a guide on which parameters are more influential than others and how the numerical model can be improved. Further researches are needed to model the nonlinearity and the dependency on confining stress of elasticity of the silty sand layer.

Figure 12. Comparison of load – displacement curves for the centrifuge test result and numerical model

4 CONCLUSIONS

The load – displacement behaviour of a monopod suction bucket foundation was studied by a centrifuge test and numerical modeling. The centrifuge model test was performed with a model soil which represented key soil characteristics of the target site. Horizontal load combined with overturning moment was applied according to the preliminary design of an offshore wind tower. In the centrifuge test, the foundation and

soil behaviour was observed for a wide load range from the initial to the post-yield load, so that the foundation design be verified and improved based on the test result. A series of numerical modeling were performed to validate the centrifuge test condition and study the effects of soil parameters on the load-displacement curves. It was found that the increased weight and vertical load provided slight increase in the resistance. The effect of the limited horizontal boundary distance in the tested centrifuge model was analysed to be minimal. Soil parameters such as elastic modulus and cohesion were found to have significant impacts than other factors in this study on the load – displacement behaviour of the monopod foundation in the silty sand layer. Refinement of the numerical model related to these parameters and elaborate estimation of them are important for realistic modeling of the foundation behaviour.

5 ACKNOWLEDGEMENTS

This study was supported by a grant from the Offshore Wind-energy Foundation System (OWFS) R&D program (10 CTIP E04) of Korea Institute of Construction & Transportation Technology Evaluation and Planning funded by Ministry of Land, Transport and Maritime Affairs and Hyundai Engineering and Construction, Co., Ltd.

6 REFERENCES

Byrne B.W. and Houlsby G.T. 2003. Foundations for offshore windturbines, Philosophical Transactions of the Royal Society of London, Series A: Mathematical and Physical Sciences, 361 (1813), 2909~2930.

Choo Y.W., Kim D.J., Kim S., Kim J.H., Kim D.S., Jee S.H. and Choi J.H. 2013. Centrifuge Tests of Monopod and Tripod Bucket Foundations for Offshore Wind Turbine Tower. Proc. of Asiafuge 2012, Indian Institute of Technology Bombay, Mumbai, India.

Houlsby G.T., Ibsen L.B. and Byrne B.W. 2005. Suction caissons for wind turbines, Frontiers in Offshore Geotechnics : ISFOG, 75~93.

Hung L.C. and Kim S.R. 2012. Evaluation of vertical and horizontal bearing capacities of bucket foundations in clay, Ocean Engineering, 52, 75~82.

Itasca. 2012. FLAC(Fast Lagrangian Analysis of Continua) 3D User's Manual, Itasca Consulting Group, Minneapolis, MN.

Kim D.J., Choo Y.W., Kim S., Kim J.H., Choi H.Y., Kim D.S., Lee M.S. and Park Y.H. 2013. Bearing capacity of monopod bucket foundations for offshore wind tower via centrifuge and numerical modeling, Journal of the Korean Geotechnical Society, under review. (in Korean)

Kim D.J., Choo Y.W., Lee J.S., Kim D.S., Jee S.H., Choi J., Lee M.S. and Park Y.H. 2013. Numerical Analysis of Cluster and Monopod Suction Bucket Foundation, OMAE2013-10480, under review.

Kim, D.S, Kim, N.R., Choo, Y.W., and Cho, G.C. (2012), "A newly developed state-of-the-art geotechnical centrifuge in South Korea," KSCE Journal of Civil Engineering, 17 (1), 77~84 (doi:10.1007/ s12205-013-1350-5).

LeBlanc C., Ahle K., Nielsen S. A. and Ibsen L. B. 2009. The monopod bucket foundation, Recent experience and challenges ahead. Eoropean Offshore Wind 2009 Conference & Exhibition, Stockholm, Sweden.

Oh M.H., Kwon O., Kim K.S. and Jang I. 2012. Economic feasibility of bucket foundation for offshore wind farm. Journal of the Korea Academia-Industrial cooperation Society, 13 (4), 1908~1914.

Villalobos F.A. 2006. Model Testing of Foundations for Offshore Wind Turbines. Ph.D. Dissertation, University of Oxford, UK.

2359

A large deformation finite element analysis solution for modelling dense sand

Solution d'analyse par éléments finis d’une large déformation pour la modélisation de sable dense

Li X.1,2, Hu Y.1, White D.11 University of Western Australian, Perth, Australia2 Beijing Jiaotong University, China

ABSTRACT: To capture the softening behaviour of dense sand, an extended Mohr-Coulomb model was developed using a critical state framework. The model extends Bolton’s correlations to capture dilatancy and peak strength, and is compatible with the remeshing andremapping strategies used in large deformation finite element analysis. This model is initially being used to simulate the behaviour of sand layers during foundation and spudcan penetration into uniform and stratified soils, but is applicable to a variety of problems that cannot be accurately simulated using conventional M-C plasticity alone.

RÉSUMÉ : Pour attraper le comportement s’adoucissant de sable, un modèle de Mohr-Coulomb étendu a été développé en utilisant un cadre critique d’état. Le modèle étend les corrélations de Bolton pour capturer la dilatance et la résistance de pic, et est compatible avec lesstratégies de remaillage et remappage. Ce modèle est initialement utilisé pour simuler le comportement des couches de sable lors de lapénétration du caisson vers les sols feuilletés. Donc, il sera applicable à une variété de problèmes qui ne sont pas bien capturées en utilisant laplasticité M-C conventionnel.

KEYWORDS: Critical state; Large deformation analysis; Remeshing and mapping algorithm; Dilation; Shear band; Biaxial test.

1 INTRODUCTION

Sand can display dilation and strain-softening during shearing under certain stress and relative density conditions. There are numerous constitutive models developed to capture these characteristics (Manzari and Dafalias 1997; Li et al. 1999). However, to be able to implement such a constitutive model into finite element software for large deformation analysis, a relatively simple model is essential with the minimum of control variables involved. This is to ensure that the large deformation analysis can be kept stable.

Large deformation of sand has not been analysed widely since large deformation doesn’t occur in general when a conventional foundation is placed on sand. However, when foundations – such as the spudcan foundations beneath offshore drilling rigs – are placed on sand overlying clay in offshore design, it is more likely for the sand layer to experience large deformation (Yu et al. 2010). Although large deformation of layered soils has been studied extensively for stiff clay over soft clay soils using large deformation FE analysis (LDFE) and centrifuge tests, fewer LDFE studies for sand over clay conditions have been executed since to date no suitable modelling approach exists for efficient simulation of the large strain behaviour of sand.

This paper describes an investigation into the dependency of bearing capacity on the large strain shearing characteristics of sand. An extended Mohr-Coulomb (MC) model was developed, which features strain-dependent hardening and softening using a critical state framework. The model uses state dependent dilatancy and friction angles. The controlling relations have been calibrated for a number of well-characterised sands, demonstrating that the model is a practical approach that can capture the specific responses of particular soils. The model was implemented in LDFE analysis (Hu and Randolph 1998a, 1998b) using the remeshing and interpolation technique with small strain model (RITSS).

The results of LDFE/RITSS with the extended MC model show that the volumetric and softening behaviour of sand has a significant influence on the penetration resistance of foundations during large penetration. When a shear band forms in sand, its dilatancy angle reaches zero and the sand finds the critical state. For foundations on uniform sand, this model shows how the variation in the bearing capacity factors Nq and N is linked to density and initial stress state, as well as the fundamental strength property, the critical state friction angle.

The extended CSMC model coupled with LDFE shows great potential to capture sand behaviour through large deformations in a simple and efficient computational framework.

2 CRITICAL STATE MOHR-COULOMB (CSMC) MODEL

2.1 State dependent dilatancy angle and friction angle

Using the critical state concept, Been and Jefferies (1985) proposed a state parameter, to identify the current soil density state and to predict the subsequent shearing behaviour. The state parameter, is defined as:

ce e (1) where e is the current void ratio; ec is the critical state void ratio at current stress. The state parameter can be used to indicate the current volume change tendency of the sand and be linked to the dilation angle (Jefferies 1993; Manzari and Dafalias 1997; Li et al. 1999; Li 2002).

Been and Jefferies (1985) reported that both the peak friction angle p and dilatancy angle decrease with increasing . This idea also can be extended to loose sand where negative dilatancy (or contraction) occurs. A simple single parameter relation can be written as: tan A (2) where A is a constant and is suggested as 1.2 (Li et al. 2013). The parameter A serves as a scale factor to the dilatancy angle, and it influences dilatancy angle in both the negative and positive regions of the state parameter , i.e. both dense and loose sands.

For a better fit to experimental data, a three-parameter relation can be written as:

s ( )tan (1 exp )nign mA (3)

where m, n are constants; n is a parameter controlling the curve shape; m is a parameter majorly influenced the curve shape with positive state parameter, i.e. loose sand.

Bolton (1986) linked peak friction and dilation angles by:

p c a (4) where c is critical friction angle; a is a constant. However, the value of a varies with soil stress condition and soil type (Li et al. 2003). Thus, the energy equation proposed by (Taylor 1948) is preferred here: tan tan tanc (5)

Combining Eqs. 3 and 5, the relation between the mobilized friction angle and soil state parameter is illustrated in Fig. 1 with the variation of parameter A. The current state-dependent dilatancy angle and friction angle can be substituted into any modified Mohr-Coulomb (MC) model such as the hyperbolic MC model (Abbo and

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Sloan 1995). This extension allows the MC model to capture soil hardening and softening behavior based on a critical state concept.

3 MODEL CALIBRATION

To implement state-dependent dilatancy and friction angles in the extended Mohr-Coulomb model developed here, the following parameters must be selected through the model calibration process (see Li et al. 2013 for further details): (1) Soil critical state line (CSL). A power relation (Li and Wang 1998) can be more accurate than the conventional log-linear CSL for sand under a confining pressure no more than 2MPa:

'

( )ca

pe ep

(6)

where ec is the critical void ratio at mean effective stress p; e is the critical void ratio as mean effective stress diminishes to zero; pa is a reference pressure taken as, pa = 101 kPa (atmospheric pressure) for convenience; p is the mean effective stress; is the slope of CSL in e versus (p/ pa) plane, which is similar to the conventional compression index; is a dimensionless constant. In this paper, is also termed as compression index and is termed as compression power for convenience. For sand, is typically 0.75 and the compression index can be estimated as 0.01Cu where Cu is the coefficient of uniformity of sand; e is estimated as 0.85emax+0.15emin where emax and emin are the maximum and minimum void ratios of the sand. (2) Dilatancy parameter A for Eq. 2 or dilatancy parameters A, m, n for Eq. 3. For Eq. 2, A = 1.2 can be selected. For Eq. 3, m, n can be estimated as 3.5, 0.75 respectively. A is to be calibrated by experimental data and is typically in the range of 0.3 to 1.0.

(3) Young’s modulus E and Poisson’s ratio . The stiffness of sand varies with void ratio and stress state. Good predictions can be made using the following equation (Hardin and Richart 1963; Wang et al. 1990; Li et al. 1999; De and Basudhar 2008):

2

0(2.97 ) '

1 a

e pE Ee p

(7)

where E0 is suggested as 6~10 MPa (Carraro et al. 2009). The bulk and shear moduli, K and G can be calculated by the usual elastic relations from and E.

4 IMPLENTATION OF CSMC IN LDFE

4.1 LDFE with RITSS technique

Large deformation FE (LDFE) analysis is conducted by remeshing and interpolation technique with small strain (RITSS) (Hu & Randolph 1998a, b). This approach is coupled with a finite element package named AFENA (Carter & Balaam, 1995). To avoid large mesh distortion and achieve large deformation simulation, a series of small strain analysis increments (using AFENA) are combined with fully automatic remeshing of the entire domain, followed by interpolation of all field variables (such as stresses and material properties) from the old mesh to the new mesh.

During the mapping of field variables, some mapping error is inevitable. The fewer number of variables that must be carried to describe the current material state, the less error will be introduced after each mapping, thus the more accurate and convergent the large deformation analysis. When CSMC constitutive model is implemented to the LDFE/RITSS, void ratio e is the only extra variable required to be interpolated in addition to the stress field. Thus, numerical stability can be kept.

In the mesh generation/remeshing algorithm, the angle in one triangle element is limited in the range of 26~111. Two criteria are used to trigger mesh refinement: (1) the distortion ratio (which is the shortest distance from the mid node to a straight line joining the corner nodes, divided by the length of that straight line) exceeding

0.02; (2) the ratio between the maximum and minimum element edge lengths exceeding 100.

0.10 0.05 0.00 -0.05 -0.10 -0.15 -0.2010

20

30

40

50

tan=tan+tanc

tan = A(1-exp(3.5*sign()*||0.75)where = e - e

c,

c=31o

Fric

tion

angl

e

(o )

State parameter

A=2A=1A=0.5A=0.2

Fig. 1 Effect of parameter A on friction angle

4.2 Biaxial test

The calibration of the model parameters is illustrated using a single element simulation of a triaxial test and by a fully meshed simulation of a biaxial test, both in Ottowa sand (Alshibli et al. 2003). The close match of the prediction and the experimental data for a single element triaxial test provides the model parameters A = 0.36, m = 8, n = 0.75 (Fig. 2).

When the calibrated parameters were applied to the bi-axial element test conditions, a much lower peak is observed (Fig. 3). However, if the dilatancy angle is increased, as the parameter A in equation 2 is raised from 0.36 to 0.6, the CSMC model shows a similar peak as the experimental data (Fig. 3). Bolton (1986) has also suggested that the dilatancy angle in plane strain test is about 1.6 times of that in triaxial test. This shows that different parameters might be needed for triaxial and biaxial test conditions. In the biaxial test, the softening behaviour is captured very well.

0 5 10 15 20 250

1

2

3

4

5

6

7

8

Prin

cipa

l Stre

ss ra

tio

=

1/3

Axial strain 1 (%)

Single element triaxial test (A=0.36) Laboratory drained triaxial test

Ottawa sand, 3=100 kPa

Soil properties: Cu

=1.4, D50

=0.22mm, eini

=0.53, Id=0.9

Model parameters: CSL: e

c=0.64-0.014(p/101)0.75,

c=36o

Dilation: tan=A(1-expsign()8||0.75

)

Fig. 2 Model calibration by single element triaxial test

Fig. 4 depicts the shear band formed in a biaxial test using the

CSMC model. A single shear band is formed first at 2% axial strain. Subsequently, a double shear band begins to form at 3% axial strain and evolves gradually. This phenomenon is consistent with the observation in Alshibli et al. (2003).

The soil in the shear band yields and dilates gradually to the critical void ratio for this stress level, which is 0.61. The dilatancy angle decreases continuously until the soil reaches the critical state, mobilising c. However, the soil outside the shear band remains at the initial void ratio, i.e. 0.54. The local strain in the shear band exceeds the external strain. The single element simulation (Fig. 2) shows a much slower decrease in the principal stress ratio after the peak than the biaxial test (Fig. 3). This confirms that the measured axial strain in laboratory tests that undergo localisation is only an apparent value (Fig. 4).

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0 1 2 3 4 5 6 70

1

2

3

4

5

6

7

8

Prin

cipa

l Stre

ss ra

tio

=

1/3

Axial strain 1 (%)

Laboratory drained biaxial test Simulation A=0.60 (rough) Simulation A=0.36 (rough) Simulation A=0.36 (smooth)

Ottawa sand, 3=69.4 kPa

Soil properties: Cu=1.4, D50=0.22mm, eini=0.54, Id=0.87Model parameters: CSL: ec=0.64-0.014(p/101)0.75, c=36o

Dilation: tan=A(1-expsign()8||0.75

)

Fig. 3 Biaxial test result of Ottawa sand simulated by FEM

0.00 0.02 0.04 0.06 0.08 0.10 0.12 0.140.04

0.00

-0.04

-0.08

-0.12

-0.16

Vol

umet

ric s

train

v

Axial strain 1

Ottawa sand, 3=69.4 kPa Numerical bi-axial test Laboratory bi-axial test Single element test

Fig. 4 v-1 relation in biaxial test

Strain localization is critical in explaining some laboratory test

results where after the peak, the deviatoric stress, q, often decreases to a stable value much earlier for axial strain than volume strain (Samieh and Wong 1997; Salgado et al. 2000; Alshibli et al. 2003). In Alshibli et al. (2003), the stress starts to oscillate around a stable value after 10% axial strain, whilst the volume strain continuously increases even over 25% axial strain. Once a “central” shear band of soil at the critical state is formed, the apparent shear strength of whole sample reaches the critical value. However the volume of whole sample still increases with yielding of soil at the margins of the shear band (Fig. 5).

0.52

0.54

0.54

0.54

0.54

0.56

0.56

0.56

0.56

0.58

0.58

0.58

0.58

0.6

0.6

0.6

0 50 100

0.52

0.54

0.54

0.54

0.54

0.54

0.54

0.56

0.56

0.56

0.56

0.56

0.560.

56

0.56

0.58

0.58

0.58

0.6

0 50 100 0.52

0.52

0.54

0.54

0.54

0.54

0.56

0.56

0.56

0.56

0.56

0.58

0.58

0.58

0.58

0.58

0.58

0.58

0.6

0.6

0.6

0 50 100

e

0.70.680.660.640.620.60.580.560.540.520.5

Fig. 5 Void ratio field: (a) 1=2%; (b) 1=3%; (c) 1=9%

The geometry of the specimen affects the shearing behavior. Biaxial simulation results with different sample aspect ratios are shown in Fig. 6. The 1-1 relation is nearly identical in all three cases. However, the v-1 relation is dependent on the aspect ratio of the soil specimen and the shape of shear band formed.

5 BEARING CAPACITY OF A FOUNDATION ON SAND

The bearing capacity of circular plate on sand is analyzed by both limit analysis (using the ABC program, Martin 2004) and LDFE. In soil with self-weight the bearing capacity factor N is often coupled with Nq although the two parts are not simply superposable. An Nq-N bounding index is defined as:

surf

Dq (8)

where D is a representative self-weight stress beneath the footing and qsurf is the surface surcharge. The coupled Nq and N bearing capacity can be characterized by an integrated bearing capacity Nq that varies with and is defined as:

uq

surf

qN

q (9)

0.00 0.01 0.02 0.03 0.04 0.05 0.060

1

2

3

4

5

6

7

8

Prin

cipa

l Stre

ss ra

tio

=

1/3

Axial strain 1

100 X 300 Sample 100 X 200 Sample 100 X 100 Sample

Bi-axial test with fully rough boundarye

ini=0.54, I

d=0.87,

c=36o

tan=0.75(1-expsign()3.5||0.75

)

c = 3.85

(a) 1-1 relation

0.00 0.01 0.02 0.03 0.04 0.05 0.060.008

0.004

0.000

-0.004

-0.008

-0.012

-0.016

-0.020

-0.024

Vol

umet

ric s

train

v

Axial strain 1

100 X 300 Sample 100 X 200 Sample 100 X 100 Sample

(b) v-1 relation

Fig. 6 Effect of sample geometry on biaxial shearing behaviour Limit analysis using ABC shows that the integrated Nq factor for a rough circular foundation can be approximated as (Fig. 7):

2 tan (1 0.48 tan )1 0.0025qN e

(10)

0.01 0.1 1 10 1008

16

32

64

128

256

512

1024

Inte

grat

ed b

earin

g ca

paci

ty fa

ctor

Nq

=qu/q

surf

Boundary condition =D/qsurf

==36o

==32o

==28o

==24o

==20o

Limit analysis of circular plate penetration into sand Best fit lines: N

q=(1+0.48tan/(1+0.0025))e2tan

Fig. 7 Coupled bearing capacity factor Nq for a circular foundation Referring to Fig. 7, the integrated Nq factor approaches a constant value with the decrease of the Nq-N bounding index . That ultimate value, e2tan, can be regarded as Nq. Similarly, the integrated Nq of rough strip foundation can be calculated as,

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5 tanexp (1 0.91tan )1 0.0025qN

(11)

6 CONCLUSION

However, plasticity limit analysis involves certain assumptions: (1) an associated flow rule, i.e. = ; (2) rigid plastic strength. The FEM method can consider the effect of soil stiffness and soil dilatancy angle on bearing capacity factor and the CSMC model allows the progressive changes in strength and stiffness during bearing failure to be captured.

Calculations of the Nq bearing capacity factor for a circular plate on weightless sand have been performed using LDFE and the MC model. The results show that both stiffness and dilatancy angle have a significant influence on the soil bearing capacity. The bearing capacity factor Nq varies by up to 50% for a realistic range of stiffness. The variation of Nq induced by the variation of dilatancy angle is no more than 15%. An empirical relation can be drawn for the estimation of Nq, as:

In this paper, the classic Mohr-Coulomb (MC) model is extended to simulate soil hardening and softening behaviour based on critical state (CS) soil mechanics. Friction and dilation angles are linked with soil state parameter in an MC model. This new critical state Mohr-Coulomb (CSMC) model is verified by single element tests and large deformation finite element (LDFE) analysis using the RITSS method. The newly developed CSMC model can be easily applied to large deformation analysis and shows good stability.

ACKNOWLEDGEMENTS

This research is supported by The National Basic Research Program of China (973 Program, No. 2012CB026104) and the ARC Discovery Project DP1096764. The third author is supported by an ARC Future Fellowship and Shell.

2 tan(0.6 0.06 ln )surfqE

qN e (12) For the plate on weighted sand, the integrated Nq is found to

vary with soil stiffness, soil weight, soil dilatancy angle and soil dimension (as shown in Fig. 8). The FEM results (Fig. 9) show that the integrated Nq approaches its ultimate value Nq if is smaller than 2, as follows:

REFERENCES

Abbo, A.J. and Sloan, S.W. 1995. A Smooth Hyperbolic Approximation to the Mohr-Coulomb Yield Criterion. Computers and Structures 54(3): 427-441.

2 tan0.015(atan 0.3)(0.65 sin )expqEN

D

(13)

Alshibli, K.A. Batiste, S.N. and Sture S. 2003. Strain localization in sand: plane strain versus triaxial compression. Journal of Geotechnical and Geoenvironmental Engineering, 129(6): 483-494.

For all the cases, the integrated bearing capacity factor can be written as (seeing Fig. 8),

Been, K. and Jefferies, M.G. 1985. A state parameter for sands. Géotechnique, 35(2): 99-112.

tan

0.5

0.45

(0.95 0.009( ) tan )1 0.02

qdsurf

qd q

E dN eDq

EN ND

(14)

Been, K., Jefferies, M.G., and Hachey, J. 1991. The critical state of sands. Geotechnique, 41(3), 365–381.

Bolton, M.D. 1986. The strength and dilatancy of sands. Geotechnique, 36(1): 65-78.

Carter, J.P. and Balaam, N.P. 1995. AFENA users manual: Geotechnical Research Center, University of Sydney.

Hu, Y.X. and Randolph, M.F. 1998a. H-adaptive FE analysis of elasto-plastic non-homogeneous soil with large deformation. Computers and Geotechnics, 23(1-2): 61-83.

0 50 100 150 200 250 300 3500.5

0.4

0.3

0.2

0.1

0.0

Plate penetration into uniform sandE=30MPa, =30.6o

q surf/D

Coupled bearing capacity factor Nq

=10g/cm3, =2o, D=4m=10g/cm3, =32o, D=4m=2g/cm3, =32o, D=40m=2g/cm3, =2o, D=40m Eq. 14

Hu, Y. & Randolph, M. F. 1998b. A practical numerical approach for large deformation problems in soil. Int. J. Numerical and Analytical Meth. Geomech. 22(5): 327-350.

Li X. Hu, Y.X. and White, D. 2013. Development of a critical state hyperbolic Mohr-Coulomb model for sand in large deformation FE analysis. Submitted to Geotechnique.

Li, X.S., Dafalias, Y.F., and Wang, Z.L. 1999. State-dependent dilatancy in critical-state constitutive modelling of sand. Candian Geotechnical Journal, 36(4): 599–611.

Ling, H.I. and Yang, S. 2006. A unified sand model based on critical state and generalized plasticity. J. of Eng. Mech., 132: 1380-1391.

Manzari, M.T., and Dafalias, Y.F. 1997. A critical state two-surface plasticity model for sands. Géotechnique, 47(2): 255–272.

Martin, C.M. 2004. ABC – Analysis of Bearing Capacity. http://www.eng.ox.ac.uk/civil/people/cmm/software.

Fig. 8 Integrated bearing capacity factor Nq�

0 100 200 300 400 500 600 700 8000

20

40

60

80

FEM result Eq. 13U

ltim

ate

coup

led

bear

ing

capa

city

fact

or, N

q

E/D

Cricular plate penetration into uniform sand==30.6o, rough condition

Riemer, M.F. and Seed, R.B. 1997. Factors affecting apparent position of steady-state line. Journal of Geotechnical and Geoenviormental engineering, 123(3): 281-287.

Richard F., Wendell, H., Michael, M. and Gioacchino, V. Strain localization and undrained steady state of sand. Journal of Geotechnical and Geoenvironmental Engineering, 122(6): 462-473.

Samieh, A.M. and R.C.K. Wong. 1997. Deformation of Athabasca oil sand in triaxial compression tests at low effective stresses under varying boundary conditions. Canadian Geotech. J., 34: 985-990.

Taylor, D.W. 1948. Fundamentals of soil mechanics. Wiley. New York. Verdugo, R., and Ishihara, K. 1996. The steady state of sandy soils.

Soils Foundation, 36(2): 81-91. Wang, Z.L., Dafalias, Y.F. and Shen, C.K. 1990. Bounding surface

hypoplasticity model for sand. Journal of Engineering Mechanics, ASCE, 116(5): 983-1001.

Salgado R., Bandini, P. and Karim, A. 2000. Shear strength and stiffness of slity sand. Journal of Geotechnical and Geoenvironmental Engineering, 126: 451-461.

Yu, L., Hu, Y.X., Liu, J., Randolph, M. and Kong, X.J. 2012. Numerical study of spudcan penetration in loose sand overlying clay. Computers and Geotechnics, 46: 1-12 Fig. 9 Ultimate value of integrated bearing capacity factor Nq

Carraro, H. Prezzi, M. and Salgado, R. 2009. Shear strength and Stiffness of sands containing Plastic or Nonplastic Fines. Journal of Geotech. and Geoenviromental Engineering, 135(9): 1167-1178.

g

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Plugging Effect of Open-Ended Displacement Piles

Prise en compte de l’effet de bouchon pour les pieux battus ouverts

Lüking J. HOCHTIEF Solutions AG, Civil Engineering Marine and Offshore, Hamburg, Germany

Kempfert H.-G. Institute of Geotechnics and Geohydraulics, University of Kassel, Kassel, Germany

ABSTRACT: During jacking an open-ended displacement pile the soil is entering through the pile toe into the profile. This plug can close up the pile toe completely. Because of this the pile can be treated approximately as a fully closed-ended displacement pile and is able to mobilize an additional base resistance. Indeed the soil-mechanical processes and the different factors of influence on the plugging effect are mostly unknown. This report is based on research work and investigated the influence of different factors on theplugging effect and hence the change in the load-bearing behaviour mainly in non-cohesive soils using experimental, numerical and statistical methods. All investigations show that a fully plugged soil inside the pile could not be identified and disproved the classicalmodel representation of a fully plugged pile toe. The load transfer in the plug takes place by compression arches, which are mainlyinfluenced by the pile diameter and the soil density. Finally, based on these results a practical calculation method is suggested.

RÉSUMÉ : Lors de la mise en place d’un pieu battu ouvert, le terrain est susceptible de pénétrer dans le pieu par son pied de manièreplus ou moins importante. Suivant le degré de pénétration du sol dans le pieu, celui-ci peut être considéré comme ouvert ou fermé et une résistance supplémentaire peut alors être mobilisée. Ce papier propose une étude des processus de pénétration du terrain dans lespieux battus ouverts pour des sols non cohésifs. La variation de capacité portante des pieux induite par ces processus est analyséeselon des points de vue expérimentaux, numériques et statistiques. Toutes les investigations réalisées montrent que l’effet de bouchoncomplet n’existe pas et qu’un pieu battu ouvert ne peut pas être considéré comme véritablement fermé. L’effet de bouchon correspondà la formation de « voûtes » à l’intérieur du pieu. Enfin, une méthode de prévision de la capacité portante intégrant ces processus est proposée.

KEYWORDS: open-ended displacement pile, plugging effect, pile bearing capacity, pile foundation.

1 INTRODUCTION

Open-ended displacement piles are piles, which are open at the pile toe like pipe piles, H-profiles or composed of sheeting piles. During the piling process (jacking, impact driving, vibrating or pressing) the soil is entering into the pile tube. Between the opposite inner shaft areas a plug can occur, which is able to mobilize an additional toe resistance. This toe resistance depends on the soil parameters, the pile geometry and the stress distribution.

Open-ended displacement piles are often used in harbour constructions or as foundations for offshore wind plants (i.e. monopiles or jackets).

Technical standards like API or others assume a fully plugged open-ended displacement pile and treat this plug in a monolithic way. However the soil-mechanical process and the different factors of influence on the plugging effect are mostly unknown.

Starting with a short state of the art this paper summarizes laboratory tests, numerical and statistical calculations and recommends new experience values for the bearing capacity of open-ended displacement piles.

These research results are based on the works described in Lüking 2010 and also Lüking and Kempfert 2012.

2 STATE OF THE ART

The bearing capacity of the plug can be evaluated by the values IFR (Incremental Filling Ratio) after Brucy et al. 1991 or the PLR (Plug Length Ratio) after Paik and Lee 1993, see Eq.1 and Eq.2.

ep dhIFR (1)

ep dhPLR (2)

These values describe the incremental and the absolute ratio

of the height of the plug hp to the pile embedded depth de. An IFR = 1 means that the surface of the plug does not

penetrate into the soil during driving in comparison to the last measurement. Only the pile penetrates into the soil. This means that no plugging effect takes place.

In contrast an IFR = 0 means that the surface of the plug penetrates into the soil with the same value as the pile. In this case the pile is fully plugged and all the soil has to be displaced sideways.

The IFR will be measured during driving by a sounding line. The PLR is only measured after finishing the driving and gives only an average value for the plug development. This is problematic in layered soils.

The highest radial displacement uR and radial stresses 'R occur by an IFR = 0. In this case the soil is fully plugged which means that the soil resistance is the same like the toe resistance of the profile. Then the plug could be treated like a monolith and is comparable to a closed-ended pile. With an increasing IFR the radial displacement and the radial stresses are decreasing. If the IFR lies between 0 and 1 the soil is partially plugged. The changeover from a full plug to a partial plug and no plugging is steady and the statuses cannot easily be distinguished. Figure 1 gives an overview of the described context after White et al. 2005. The maximum pile diameter in which a plugging effect could occur is about 1.5 m, see Jardine et al. 2005.

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Figure 1. Distribution of the radial displacement uR and the radial stress 'R on the pile shaft depending on different IFR after White et al. 2005

3 EXPERIMENTAL INVESTIGATIONS

3.1 General

Different experimental investigations were carried out. The next section gives a short overview of the laboratory test program before the results are discussed. A detailed description and a documentation of all test results is given in Lüking 2010.

3.2 Model Tests and Particle Image Velocimetry Tests

In the first test series a test pile of two pipe piles was constructed. Both piles were only connected at the top. In all the test pile had the following geometry: outer pile diameter 19 cm and inner pile diameter 16 cm. The pile embedded depth after driving the test pile into a sand box was about 140 cm. After this a static pile test loading was carried out.

This test pile was equipped with different strain gauges. Based on the measured strain the inner shaft friction qis, the outer shaft friction qs and the pile toe pressure qb could be calculated. By means of a special constructed cone-penetration-test (lab-CPT) the change in density and the displacement effect of the pile installation could be examined.

In the second test series Particle Image Velocimetry (PIV) tests were carried out. The PIV method is a contact free measurement, in which displacement vectors can be identified. Basics to this method can be found in Raffel et al. 2007.

The test pile in the second test series had an outer pile diameter of 60 mm and a wall thickness of 2 mm. It was driven behind an acrylic glass to an embedded depth of 50 cm. Figure 2 gives a perspective view of both test series which were mainly carried out in non-cohesive soils.

Figure 2. Perspective view of a) test pile of the first test series and b) test pile of the second test series (PIV)

3.3 Results of the experimental test series

In general the experience on the pile bearing behaviour regarding different influence factors could be confirmed. With

an increasing relative density and increasing stress level the pile bearing capacity is also increasing.

The change in density around the test pile was lower in dense sands than in loose sands, which could be identified by different tests with the lab-CPT. The base resistance of the lab-CPT inside the soil plug was up to 80 MPa. A higher density of the soil tends to a higher IFR. Nevertheless the IFR does not converge to a fixed value. It was increasing and also decreasing during driving which means that the soil inside was plugging and loosening again. This phenomenon was also identified during the static pile test loading. However during both test series the value never reached IFR = 0. The minimum was IFR = 0.2. This means that only a partially plugged soil could occur and based on this the concept of a monolithical soil plug should be analyzed critically.

Figure 3 shows the distribution of the inner and outer shaft friction at different load levels from the first test series. The outer shaft friction is increasing with higher pile length as expected. In contrary the inner shaft friction is very high on a length which approximate two pile diameters. Above this the inner shaft friction in section 1 and 2 is very low and it looks approximately independent of the load level. The increasing of the inner shaft friction in section 3 is an indication for a (partial) plugging effect of the soil.

0 50 100 150 200 250 300 350

outer skin friction qs [kN/m2]

load levels30.1 kN

60.2 kN

90.2 kN

120.3 kN

135.3 kN

160

120

80

40

0

pile

leng

th L

[cm

]

350 300 250 200 150 100 50

inner skin friction qis [kN/m2]

pile wall

section a

section b

section 1

section 2

sect

ion

3

Figure 3. Distribution of the inner and outer shaft friction qis and qs for different load levels in non-cohesive soils.

Figure 4 shows the vertical displacement of the soil on the lowest two pile diameters exemplary for the second test series.

Figure 4. Distribution of the vertical displacements in the soil at the pile toe in the second test series in non-cohesive soils.

There an inhomogeneous distribution could be identified. Near the inner pile shaft the vertical displacement is much higher than in the middle of the soil plug. This distribution occurs during driving independently of all investigated boundary conditions in the second test series. It is another indication that the load transfer takes place by the inner shaft friction and not by an additional base resistance underneath the soil plug. This assumption can also be supported by the comparable distribution of the inner and outer shaft frictions, see also in Figure 3. For a monolithic approach the vertical displacement had to be more constant which could not be observed. Furthermore the tests show that these results in non-cohesive soils cannot be transferred easily to cohesive soils. It

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looks like that two different mechanisms are active which are not comparable.

4 NUMERICAL INVESTIGATIONS

4.1 General

The experimental works were further investigated by finite element calculations. The experimental test loadings as well as test loadings with bigger pile diameter were recalculated. The numerical calculation software PLAXIS 2D - Version 9.0 was used. A rotation-symmetric, 2-dimensional FE-model was built. The simulation of the soil displacement during jacking was considered by the method of Dijkstra et al. 2006. A detailed calculation description and the verification of the numerical model is given in Lüking 2010. Finally the numerical results confirmed the results of the experimental tests quantitatively and qualitatively.

4.2 Results of the numerical calculations

Figure 5 shows the distribution of the inner skin friction for different inner pile diameters (Di = 0.45 m up to Di = 3.95 m) of a pile which is embedded in the soil of about de=10 m. The soil is non-cohesive and has a “dense” relative density (cone penetration resistance of about qc ≈ 20 MPa). The settlement for the mobilization of the skin friction was about s = 4.2 cm.

10

8

6

4

2

0

pile

leng

th [m

]

0 100 200 300 400 500

inner skin friction qis [kN/m2]

Di = 0.45 m

Di = 0.95 m

Di = 1.45 m

Di = 1.95 m

Di = 2.95 m

Di = 3.95 m

Figure 5. Distribution of the inner skin friction qis under variation of the inner pile diameter Di at a pile embedded length of de = 10 m and a pile settlement of about s = 4.2 cm

At low pile diameter (Di = 0.45 m) the results show a good agreement in the distribution to the experimental works, compare Figure 3 with Figure 5.

Furthermore the results show that the inner skin friction for lower pile diameters is significantly higher at a length of approximately two pile diameters. On the upper part of the pile length no skin friction was mobilized. With an increasing pile diameter the peak value of the skin friction is reduced and is transferred to the upper part of the pile. At pile diameters of about 3 m or 4 m the distribution of the inner skin friction is comparable to the outer skin friction. The changeover from a raised inner skin friction to a more constant inner skin friction is continuous. Calculations show that this changeover depends mainly on the pile diameter and the relative density of the soil, see Lüking 2010. The distribution of the inner skin friction is also valid at “loose” relative density (qc ≈ 10 MPa).

Figure 6 shows the numerical results for the orientation of the stress trajectories and for the load transfer depending on the pile diameter in a “dense” relative density of a non-cohesive soil. The left part of each pile shows the derived load transfer based on the stress trajectories which are shown on the right part. In general all results show a rotation of the stress trajectories near the pile toe and also at the pile wall. With increasing distance from the pile wall to the middle of the soil plug the rotation is reducing. Also this depends mainly on the pile diameter.

Figure 6. Numerical results for the orientation of the stress trajectories (right part of each pile) and the derived load transfer (left part of each pile) for different pile diameters in a "dense" relative density of a non-cohesive soil

The orientation of the stress trajectories suggests a compression arch, which is in analogy to the load transfer mechanism of the outer skin friction, see Kempfert 2009.

At low pile diameters these compression arches can be overlapped and results in another support. Because of this the inner skin friction can increase significantly which is also shown in the numerical and experimental results, compare Figure 5 and Figure 3. With increasing pile diameter the height of the compression arches is also increasing. This load transfer could also be identified in “loose” relative density.

Finally the results suggest that the load transfer takes place over an inner skin friction which is based on compression arches inside the soil. No fully plugged soil inside an open-ended displacement pile could be identified which would legitimate to treat the plug in a monolithic way.

5 CALCULATION METHODS

5.1 General

Based on the new knowledge two feasable methods for calculating the bearing capacity of open-ended displacement piles are suggested. The values were verified statistically to a large extend with calculation method 1 up to a pile diameter of D = 1.6 m in cohesive and non cohesive soils and with calculation method 2 up to a pile diameter of D = 1.2 m in non-cohesive soils. All histograms of the statistical verifications can be found in Lüking 2010.

5.2 Calculation Method 1

Calculation method 1 is based on an analysis of 28 static and 59 dynamic pile loading tests with pile diameters up to D = 1.6 m. This method derived new adaptation factors which are linked to the values of experience of the EA-Pfähle 2012. The basic equation for calculating the pile resistance is given in Eq.3.

skssbkbbk AqAqR ,, (3)

Rk: characteristic pile resistance b: adaptation factor for the pile toe, see Eq.4 qb,k: characteristic pile toe pressure after EA-Pfähle 2012 Ab: pile base area (contact area of the pile and the

bottom area of the soil plug) s: adaptation factor for the pile skin, see Eq.5 qs,k: characteristic pile skin friction after EA-Pfähle 2012 As: outer shaft area of the pile

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The best compliance for the adaption factors was found by a hyperbolic correlation, see Eq.4 and 5.

5.4 Comparable Calculations

Figure 7 gives an overview of the caclulation results of both methods compared with the results of the pile load tests.

aD

b e 2.1 95.0

D 63.0

(4)

0 2 4 6 8 10Rm [MN], static pile load test

0

2

4

6

8

10

Rca

l [M

N],

calc

ulat

ion

met

hods

calculation method 1,10 % quantile




as e 1.1 (5)

Da: outer pile diameter

5.3 Calculation Method 2

Calculation method 2 is based on an analysis of 28 static pile loading tests with pile diameters up to D = 1.2 m. In contrary to calculation method 1 this method derived new values of experience for each part of pile resistance for the 10 % and the 50 % quantile. Eq.6 gives the basic equation.

sksakaiskisk AqAqAqR ,,, (6)

Rk: characteristic pile resistance qis,k: characteristic inner pile skin friction after Table 1 qa,k: characteristic pile toe pressure of the pile contact

area after Table 2

Figure 7. Calculation results for the characteristic pile resistance Rcal of both calculation methods in comparison to results Rm of static pile load tests

Figure 7 shows that the requirements of the calculation methods for the 10 % and 50 % quantile are fully accomplished. Further calculations and variations of parameters are given in Lüking 2010.

qs,k: characteristic outer pile skin friction after Table 3 Ais: inner shaft area of the pile Aa: contact area of the pile As: outer shaft area of the pile This method is valid for pile diameters from 0.3 m up to 1.2

m only in non-cohesive soils. The first values of the experiences in the following tables are the 10 % quantile and the second are the 50 % quantile.

6 SUMMARY

Table 1. Values of experience for the characteristic inner shaft friction

is,k depending on the pile settlement and the resistance of the CPT

The load transfer inside a plug of an open-ended displacement pile was investigated by experimental, numerical and statistical methods. It was shown that the load transfer takes place by compression arches. A fully plugged soil could not be identified.

q

Characteristic inner shaft friction qis,k [kN/m2]at a cone penetration resistance qc [MN/m2]Settlement s

7.5 15 ≥ 25

s = 0.035•Da 15/÷ 35/ 35/÷ 55/ 50/÷ 67.5/

s = 0.1•Da 30/÷ 50/ 60/÷ 80/ 90/÷ 100/

with = 2•PLR, see Eq. 2

7 REFERENCES

API RP 2A-WSD 2007. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms - Working Stress Design, 21st Edition, American Petroleum Institute, Washington

Brucy F., Meunier J. and Nauroy J.-F. 1991. Behavior of Pile Plug in Sandy Soils during and after Driving. Proceedings of the 23rd Offshore Technology Conference, OTC 6514, Vol. 1, pp 145-154

Table 2. Values of experience for the characteristic pile toe pressure qa,k epending on the pile settlement and the resistance of the CPT

Dijkstra J., Broere W., and van Tol A. F. 2006. Numerical Investigation into Stress and Strain Development around a Displacement Pile in Sand. Proceedings of the 6th European Conference on Numerical Methods in Geotechnical Engineering. NUMGE 06. pp 595-600 d

Characteristic pile toe pressure qa,k [kN/m2] at a cone penetration resistance qc [MN/m2]Settlement s

7.5 15 ≥ 25

s = 0.035•Da 650÷ 1.200 1.300÷ 1.750 1.750÷ 2.800

s = 0.1•Da 1.100÷ 2.000 2.000÷ 3.000 2.800÷ 4.800

Empfehlungen des Arbeitskreises „Pfähle“ EA-Pfähle 2012. Empfehlungen des Arbeitskreises „Pfähle“; 2. Edition, Ed. Arbeitskreis „Pfähle“ of the German Society of Geotechnics. Ernst & Sohn. Berlin

Jardine R. J., Chow F. C., Overy R. F. and Standing J. R. 2005. ICP Design Methods for Driven Piles in Sands and Clays. Thomas Telford, London

Kempfert H.-G. 2009. Pfahlgründungen. Chapter 3.2 in: Grundbau-Taschenbuch. 7th edition. Part 3. Ernst & Sohn. Berlin. pp 73-277

Table 3. Values of experience for the characteristic outer shaft friction s,k depending on the pile settlement and the resistance of the CPT

Lüking J. 2010. Tragverhalten von offenen Verdrängungspfählen unter Berücksichtigung der Pfropfenbildung in nichtbindigen Böden. Schriftenreihe Geotechnik, University of Kassel, Issue 23. q

Lüking J. and Kempfert H.-G. 2012. Untersuchung der Pfropfenbildung an offenen Verdrängungspfählen. Bautechnik 89, Issue 4, pp 264-274.

Characteristic outer shaft friction qs,k [kN/m2] at a cone penetration resistance qc [MN/m2]Settlement s

7.5 15 ≥ 25

sg* 15÷ 25 30÷ 50 50÷ 70

s = 0.1•Da 20÷ 30 35÷ 60 55÷ 75

with sg* [cm] = 0.5•Rs,k [MN] ≤ 1 [cm]

Paik K.-H. and Lee S.-R. 1993. Behavior of Soil Plugs in Open-Ended Model Piles Driven into Sands. Marine Georesources and Geotechnology, Vol. 11, pp 353-373

Raffel M., Willert C., Wereley S. and Kompenhans J. 2007. Particle Image Velocimetry - A Practical Guide. Second Edition, Springer-Verlag, Berlin Heidelberg New York

White D. J., Schneider J. A. and Lehane B. M. 2005. The Influence of Effective Area Ratio on Shaft Friction of Displacement Piles in Sand. Proceedings of the International Symposium on Frontiers in Offshore Geotechnics, Balkema, Rotterdam, pp 741-747

2367

A simplified procedure to assess the dynamic stability of a caisson breakwater

Une procédure simplifiée pour évaluer la stabilité dynamique d’une digue en caissons

Madrid R., Gens A., Alonso E., Tarrago D. Dep. of Geotechnical Engineering and Geosciences, Technical University of Catalonia, Barcelona, Spain

ABSTRACT: The paper describes a simplified method of analysis used to evaluate the stability of a caisson breakwater to sea waveactions. An intensive laboratory program was performed in order to evaluate the static and dynamic characteristics of the foundation soil. Anisotropic and isotropic consolidated cyclic triaxial tests and cyclic simple shear tests were used to define the cyclic interactiondiagram for the foundation soil. The possibility of foundation cyclic mobility due to wave loading and their effect on the breakwaterstability was examined combining the cyclic interaction diagram with the results of finite element analysis. The potential reduction insoil strength is then incorporated into a conventional stability analysis. The procedure is illustrated by a specific application to acaisson breakwater that is part of the extension works of the Barcelona Harbour.

RÉSUMÉ : L’article décrit une méthode simplifiée pour évaluer la stabilité d’une digue verticale sous l’action de la houle. Lescaractéristiques statiques et dynamiques de la fondation ont été évaluées à l’aide d’un programme intensif de tests en laboratoire, quiinclut des essais triaxiaux cycliques isotrope et anisotrope et des essais de cisaillement simple cycliques dans le but d’établir le diagramme d’interaction cyclique du sol. La possibilité d’une mobilité cyclique de la fondation sous l’action de la houle et son effetsur la stabilité de la digue ont été examinés en combinant le diagramme d’interaction cyclique ainsi obtenu avec une analysenumérique par Éléments Finis. La réduction potentielle de la résistance du sol est ensuite incorporée dans une analyse de stabilitéconventionnelle. La procédure est illustrée par une application spécifique à une digue en caissons qui fait partie des travauxd'extension du port de Barcelone.

KEYWORDS: cyclic tests, interaction diagrams, liquefaction, caisson breakwater, wave loading, stability.

1 INTRODUCTION

Two new breakwaters and a large container area, immediate to a new quay, are the main development works of the ongoing extension of Barcelona harbour. A plan view of the new breakwaters and quays is shown in Figure 1.

Breakwaters have a total length of 6.8 km. The East breakwater is of a rubble mound type whereas the South breakwater involves two different types: rubble mound and vertical caissons. This paper refers to the caissons section that has a total length of 1.7 km constructed in water depths that range from 20m to 25m. Most of the foundation soil immediately under the breakwaters consists of weak sediments of clayey silts and silty clays belonging to the pro-deltaic deposits of the Llobregat River.

The paper describes summarily the main geotechnical features of the foundation ground with special attention given to undrained strength parameters. The bases for the static design of the breakwater are then briefly presented. Finally, a description of the cyclic resistance of the foundation soil is described in terms of an interaction diagram; this information is then used in a simplified assessment of the stability of the breakwater under storm conditions incorporating the potential strength reduction due to cyclic loading.

2 SOIL PROFILE CHARACTERISITCS

A representative soil profile at the location of breakwaters is shown in Figure 2. It consists of: i) upper silts and clays, brown and grey in colour, although dark colours occasionally appear when organic matter content increases. The thickness of this deposit underneath the breakwaters is about 50 m. Sandy and silty sand inter-stratifications, were often found, specially in the

upper levels of the layer. ii) an intermediate layer of gravels and sands, whose thickness is about 7 m; some silt partings were also detected. iii) a lower level of clays whose identification properties are similar to the upper clay unit, although it is a denser soil. The maximum thickness of this layer is 14 m. iv) a lower layer of gravels and sands; it includes several clays and sands stratifications.

Figure 1. Plan view of the new breakwaters and new container areas of the Barcelona harbour. The location of the caisson breakwater is indicated.

Closer to the coast line, an upper deltaic sand deposit of increasing thickness, laid on top of the upper stratum of soft silty clays, appears. As it would be expected from a deltaic environment the transitions between this sand deposit and the upper clays are neither sharp nor regular. This sand deposit is 15 m thick at the shore line but it practically disappears at the breakwater location and it is not considered further in this paper.

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Some of the geotechnical indices and properties obtained during the site investigations are shown in Figure 3. The fine grained materials classify mainly as CL (low plasticity clays) and ML (low plasticity silts). Water content commonly exceeds the liquid limit in the upper part of the soft silty clay unit, but at lower levels it is close to the plastic limit, an indication of the self weight consolidation of the sediments. Void ratios range between 0.8 and 1.0 in the upper clay stratum. Dry densities vary from 1.2 Mg/m3 at the upper clay levels to 1.8 Mg/m3 at deeper locations.

AS-6AS-5

S-6AS-4 S-5 Apz-14 AS-3

Upper level of silt and clays

Intermediate level of gravels and sands

Lower level of clays

Lower level ofgravels and sands

0

-110

-100

-90

-80

-70

-60

-50

-40

-30

-20

-10

Figure 2. Soil profile under the caisson breakwater

10 20 30 40 50w (%)

120

100

80

60

40

20

0

Dep

th (m

)

20 30 40 50wL (%)

0 20 40 60wP (%)

0.8 1.2 1.6 2d (Mg/m3)

0 1 2IL (%)

120

100

80

60

40

20

0

Dep

th (m

)

0.4 0.8 1.2e

Figure 3. Basic soil properties

In the low permeability foundation soils, stability is controlled by the undrained shear strength (cu). In the normally

consolidated range, this parameter is largely proportional to the consolidation effective vertical stress. Undrained shear strength has been examined by means of laboratory and in situ tests.

Unconfined compression tests of clay samples provided a value of cu=0.215’v. However, sample disturbance and suction loss may lead to an underestimation of the real value (e.g. Tsuchida, 2000). Simple shear tests performed by NGI provided a value of cu=0.25’v. quite consistent with the results of CPTU tests. Anisotropically consolidated triaxial tests (compression and extension) yielded a range of cu=0.21 – 0.33 ’v, the larger values associated with compression tests. A summary of results obtained is presented in Figure 4. The unusually large values of undrained strength obtained in some vane tests were probably due the occasional presence of sand lenses or laminations.

0 100 200 300 400 500 600 700'v: kPa

0

100

200

300

c u: k

Pa

Unconfined comp.Triaxial testVane testCPTU tests

cu=0.215'v

cu=0.32'v

cu=0.25'v

Figure 4. Undrained shear strength. Summary of results

It was also found that specimens sheared under normal effective stresses reproducing in situ stress conditions showed somewhat higher strength ratios than specimens consolidated to higher effective stress values. This is an indication of some modest overconsolidation/structure effects due to natural creep or aging phenomena. However, the additional stresses applied by the caissons and fills will take the soil in situ to a normally consolidated state. Therefore, a conservative attitude is favoured for the selection of the undrained stress ratio. The static design of the breakwater was eventually performed using a value of cu=0.25’v.

3 BREAKWATER DESIGN

The conventional breakwater design was performed using finite element analysis as the most efficient method to consider automatically the variation of undrained shear strength throughout all stages of construction. The following phases were considered: i) dredging and bench construction on the new soil surface, ii) caisson placement and filling, iii) construction of the superstructure, and iv) backfill behind the caissons to create a new quay zone. Although all potential limit states were considered, it should be pointed out that the use of finite element analysis readily identifies the most critical failure mechanism at every stage of the analysis. It should also be noted that the gain in undrained shear strength during each one of the construction phases was a critical feature with respect to the stability of the subsequent construction phase.

The wave and uplift forces due to storm loading in the different phases of construction are listed in Table 1. They were derived from physical model tests using the specific breakwater design. Wave forces depend on two factors: the height of the superstructure that provides the surface on which the wave impact acts and the wave height that in turn depends on the intensity of the storm. It can be observed that the wave height (and hence the storm intensity) is lower in Phase II. This is due to the temporary character of this Phase that makes it less likely

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that an extremely large storm will occur during that limited period. Probability analysis based on available time series provides the design storm to be used in each particular stage.

For each construction stage, a variety of factors of safety were used to assess the degree of stability of the breakwater affecting either loads or soil strength parameters. In the former case, wave caisson weight and storm wave loads were considered both jointly and separately. The values of safety factors were assessed in relation with the perceived uncertainty of the parameters involved. Thus, a higher factor of safety was demanded when only the wave action was considered due to the much larger uncertainty of the load magnitude associated with the storm. In fact, uncertainty affects both storm intensity and the actual effect on the caisson. The final design of the breakwater is depicted in Figure 5. Note the wide rockfill bench required for stability. An example of the failure mechanism in a particular instance of the analysis is shown in Figure 6.

Table 1. Wave and uplift forces acting on the caissons breakwater at different phases of construction

PhaseShoulder

height(m)

Wave height

(m)

Wave period

(s)

Wave force

(kN/m)

Forceheight

(m)

Dynamic uplift

(kN/m)

II No 5 9 1036.3 9.48 525.1

III +6 5.91 12.7 1436.1 10.36 878.2

IV +11 8.04 12.7 748.9 6.10 766.2

CAISSON

RUBBLE MOUND

172.27

-26.00

RIP RAP 300kg

SEAWARD SIDE

RIP RAP 4 tonCONCRETE BLOCK

RIP RAP 300kg

0.00

-21.00-23.00

-13.75-15.00-13.00

-22.00

-18.00

0.00

+11.00

+3.00

Figure 5. Design of the caisson breakwater.

Figure 6. Failure mechanism for Phase III under storm loading.

4 CYCLIC SHEAR STRENGTH

However, breakwaters are also subjected to cyclic wave loading. Storms are the primary source of energy that may cause cyclic mobility or, in extreme conditions, liquefaction of foundation soils. Even if such extreme events do not occur, undrained shear strength may be lower after a severe episode of cyclic loading. Consequently, clay behaviour under cyclic loading was also investigated in the laboratory by performing cyclic simple shear and triaxial tests (on isotropically and anisotropically consolidated specimens). Data from simple shear tests were favoured because they appear to correspond more closely to the actual breakwater foundation conditions during storm loading.

Results from these tests can be usefully summarized using interaction diagrams such as that shown in Figure 7. This diagram shows a relationship between the normalized average shear stress a/’vc, normalized cyclic shear stress cy/’vc and the number of cycles to reach the cyclic mobility criteria. Also, results obtained from simple shear testing on the plastic Drammen clay (Goulois et al, 1985) are shown for reference.

Failure occurs for a given combination of normalized cyclic and average shear stress. Figure 7 shows the approximate bounds of these combinations for two different loading conditions (40 impacts and 1000 impacts). The normalised cyclic shear stress cy/’vc , for low values of the normalized average shear stress, is close to 0.17 for 40 cycles and to 0.10 for 1000 cycles. A second static bound is provided by the relationship cy/’vc+a/’vc=0.25, which is based on the previous discussion on static undrained strength.

40 impacts

1000 impacts

Figure 7. Interaction diagram from direct simple shear tests (NGI, 2002).

5 SIMPLIFIED ANALYSIS USING THE INTERACTION DIAGRAM

An example of the simplified stability analysis concerning Phase III of construction is presented in this section. The design storm established for this Phase is summarised in Table 2 that contains the number of waves of different heights corresponding to a succession of storm intensities with different durations and different significant wave heights.

It is assumed that the design storm can be represented as the application of a number of wave impacts of a certain magnitude. Then, a static analysis can be used to identify areas in the foundation soil where the stress state exceed the criteria of unstable stress defined by the interaction diagram. Naturally, to use the information contained in the interaction diagram, it is necessary to transform the variable wave loads of the storm into a series of cycles of uniform magnitude. This transformation always involves, to a certain extent, a degree of uncertainty and approximation. It is therefore advisable to adopt a measure of conservatism.

With this approach, two loads intensities were selected from the wave magnitudes shown in Table 2: a large load of 1011.5 kN/m and a smaller load of 341.6 kN/m. The former is assumed to act forty times and the latter five thousand times. The limit criteria corresponding to those two numbers of cycles have been indicated in Figure 7.

Now, it is possible to compute, using a conventional static finite element analysis and applying the corresponding wave loads, the points at which such criteria are exceeded, indicating the possibility that, in those zones, a degree of cyclic mobility occurs with a potential reduction of the undrained shear strength. A quite conservative assumption is that the operational undrained shear strength reduces to the residual value of cu. The foundation zones affected are shown in Figures 8 and 9. They

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are quite similar in the two cases and affect only a quite limited area of the foundation soil.

Table 3. Computed factors of safety (on strength parameters)

Factor of safety

Consideration of cyclic loading

Wave force = 341.6 kN/m

1000 impacts

Wave force = 1011.6 kN/m40 impacts

Yes 1.48 1.18

No 1.55 1.40

Table 2. Characteristics of the design storm

Number of waves

Wave height

(m)

Hs =3m

(24 h.)

Hs =4m

(24 h.)

Hs =5m

(24 h.)

Hs =5.9m

(24 h.)

Wave force

(kN/m)

Forceheight

(m)

1-2 3124 1661 593 150 213.4 9.77

2-3 2203 1693 720 199 341.6 10.28

3-4 848 1133 626 198 475.0 10.62

4-5 195 545 427 161 685.0 10.65

5-6 28 194 236 111 825.4 10.74

6-7 2 52 108 66 870-5 10.80

7-8 0 11 41 34 920.0 10.00

8-9 0 2 13 16 1011.5 10.36

9-10 0 0 3 6 1410.1 11.46

10-11 0 0 1 2 1528.0 11.15

11-12 0 0 0 1 1559.3 11.38

6 CONCLUDING REMARKS

Finally, a new stability analysis is performed with the new distribution of undrained shear strength of the foundation soil for the two cases considered. The analysis also considers the influence of the dynamic uplift caused by the storm loading, derived from the physical model tests carried out for this particular breakwater design. The results, in terms of factor of safety for strength reduction, are shown in Table 3. It can be seen that consideration of cyclic loading has a moderate but noticeable impact on the factor of safety. Given the exceptional character of the design storm and the conservative assumptions made in the analysis, the factor of safety obtained was considered adequate for accepting the design.

A key design feature of a breakwater is the assessment of the stability of the breakwater when subject to extreme storms. This is particularly the case for caisson breakwaters in which the effects of wave action are significantly stronger than for the classical rubble mound type. A proper consideration of the dynamic effects would require the performance of a full dynamic analysis. Here, a simplified stability analysis is proposed that takes into account the potential reduction of the shear strength of the soil due to cyclic loading. It is based on the experimental determination of the interaction diagrams that provide criteria to identify the conditions for which the soil can undergo cyclic mobility and strength degradation. The corresponding strength reduction is then taken into account in conventional stability analyses. The procedure has been illustrated by a specific application to a caisson breakwater that is part of the extension works of the Barcelona Harbour.

7 ACKNOWLEDGEMENTS

The authors are grateful for the technical and financial support provided by APB (Autoritat Portuaria de Barcelona).

8 REFERENCES (TNR 8)

-80

-60

-40

-20

0

20

-100 -80 -60 -40 -20 0 20 40 60 80 100 120

Goulois, A.M., Whitman, R.V., and Hoeg, K. (1985). Effect of sustained shear stress on the cyclic degradation of clay. Proceeding, Symposium on Strength Testing of Marine Sediments, R.C. Chaney and K.R. Demars, eds., ASTM STP 883, ASTM, Philadelphia, 336-351.

NGI. (2002). Laboratory Testing. Geotechnical Testing Report. March 2002.

Tsuchida, T. (2000). Evaluation of undrained shear strength of soft clay with consideration of sample quality. Soil & Foundations 40, No 3, 29-42.

Figure 8. Foundation zones exceeding the interaction diagram criterion. Wave load = 341.6 kN/m and 1000 cycles.

-80

-60

-40

-20

0

20

-100 -80 -60 -40 -20 0 20 40 60 80 100 120

Figure 9. Foundation zones exceeding the interaction diagram criterion. Wave load = 1011.5 kN/m and 40 cycles.

2371

The new remediation technique for buried pipelines under permanent ground deformation

Une nouvelle technique de pose des conduites enterrées soumises à des déformationspermanentes du sol

Moradi M., Galandarzadeh A., Rojhani M. Department of Soil Mechanics and Foundation Engineering of Collage of Civil Engineering, University of Tehran

ABSTRACT: One part of lifelines is buried pipelines such as gas, water and oil pipelines. Permanent ground deformation such as fault crossing and lateral spreads is one of the more important threats for pipelines. In this research, a new remediation technique forburied pipeline system subject to permanent ground deformation is proposed. Also this new technique has been evaluated bycentrifuge modeling of buried pipelines subjected to concentrated PGD. In proposed technique, the high porosity gravels are used aslow-density backfill to fill the trench around the pipe near the susceptible area to PGD, thereby reducing soil resistance and soil-pipe interaction forces and also pipeline strains. Previously, the expanded polystyrene (EPS) geofoam proposed to reduce density ofpipelines backfill. However, the high porosity gravel is better than expanded polystyrene geofoam from many cases such asworkability to construct, environmental effect, durability and cost. In this technical paper, described the proposed technique and also two centrifuge modeling have been done to evaluate its performance. The comparisons of responses of remediated pipeline with unremediated pipeline have been shown that the proposed technique is effective considerably.

RÉSUMÉ : Une partie des réseaux nécessaires au transport du gaz, de l’eau et du pétrole est constituée de conduites enterrées. Lesdéformations permanentes du sol dues à des tassements ou à des mouvements latéraux sont l'une des menaces les plus importantespour les conduites enterrées. Dans cet article, une nouvelle technique de pose des conduites enterrées soumises à des déformationspermanentes du sol est proposée. Cette nouvelle technique a été évaluée par des essais en centrifugeuse sur des canalisations enterréessoumises à des déformations permanentes du sol. Pour la technique proposée, des matériaux sableux dont la porosité est élevée sont utilisés pour le remplissage des tranchées. Ils réduisent les efforts induits par l’interaction sol-tuyau. Auparavant, c’est le polystyrène expansé geofoam qui était utilisé. Le matériau proposé est meilleur que le polystyrène expansé geofoam en ce qui concerne la mise en œuvre, l'effet sur l'environnement, la durabilité et le coût. Dans ce papier, la technique proposée est décrite ainsi que deux modèles encentrifugeuse réalisés pour évaluer sa performance. Les résultats obtenus montrent que la technique proposée est plus efficace quecelle utilisée précédemment.

KEYWORDS: Centrifuge Modeling, Faulting, Lifelines, Pipeline, Earthquake

1 INTRODUCTON

Buried pipelines often serve as lifelines in that they may carry resources that are essential to the support of human life and this is the reason to retain them in serviceable condition in every situation. Among various kinds of natural hazards, earthquakes happen to be the most serious threats for lifelines serviceability. They can damage lifelines through faulting, permanent ground deformation (PGD) and deformations due to seismic wave’s propagation. Faulting can affect pipelines in various ways (Fig. 1) and cause severe damages (Fig. 2) depending on faulting movement direction.

Considering mentioned hazards, lots of statistical, analytical and numerical studies have been conducted since 1970s in order to predict pipelines response and vulnerability level and also to investigate methods of damage mitigation; but it has been a difficult and somehow impossible way to evaluate theoretical and analytical research results due to loss of accurate and efficient records about pipelines response to faulting in actual case histories of earthquakes (Choo et al. 2007). In order to compensate such a gap, studies turned towards applying experimental and physical modeling of this phenomenon. Since 2003, significant researches have been started in U.S.A. and Japan with support of companies and institutes such as Tokyo Gas Company, US lifelines Agency, National Science Foundation in U.S.A, Earthquake Engineering Research Center and etc. Most of mentioned conducted studies have been focused on strike-slip faulting. So, still there is lack of studies

on normal and reverse faultings’ effects and this puts them in prime importance of research priority.

Herein, the authors investigated the pipeline response due to reverse faulting and also investigated the use of high porosity gravel as low-density backfill in pipeline response. It is expected that low-density backfill for will reduce soil-pipe interaction and reduce the pipe train.

Table 1. Centrifuge Facility Properties

Property Unit Quantity

Exerted acceleration g 5 – 130 Acceleration accuracy g + 0.2 Rotational velocity range rpm 38 – 208 Rotation radius m 3Maximum model weight (up to 100 g) kg 1500Maximum model weight (up to 150 g) kg 500

1.1 Faulting simulator split box

Experimental setup provision in order to use in centrifuge instrument has its own limitations; for instance, weight and dimensions of the box is thoroughly tied to the used centrifuge facility properties and it is of prime importance for the box to have the minimum weight and dimensions possible together with having enough strength for high magnitude forces caused due to high exerted accelerations. Regarding these limitations, the group-7000 aluminum alloy which has low density and high strength is used to build up the faulting simulator split box in

2372


this study. Outer dimensions of the box are 102×76×68 cm (l×w×h) and the inner dimensions are 96×70×23 cm. The split line of the box which is the faulting line itself, makes the angle of 30˚ from the vertical direction. The box setup is assembled and fixed on a 4 cm thick aluminum block of 15 cm width that can bear the hydraulic jack caused 5 ton horizontal force and high magnitude vertical force which is exerted due to high accelerations. Holes have been cut in the two ending walls of the box as the backrests for studied structures such as pipelines. Regarding lack of space in the centrifuge basket, the motivating system and the other constituents of the simulator must occupy the minimum space possible.

Moving mechanism has been designed to be enough stable during the faulting movement and also can bear the high magnitude unbalanced forces derived from soil-structure friction.

A wedge-sliding mechanism has been applied for the box movement to direct the faulting through the 30˚ specified direction and prevent form any strike between fixed and moving parts of the split box. The wedge-sliding mechanism is consisted of two rails installed with the angle of 30˚ from the vertical direction and high level force tolerating ball bearings to guide the movement as desired. Sliding the wedge forward and backward, the moving part of the box would have an upward-downward movement (Fig. 4). Considering the high magnitude forces and weight increase in high order accelerations, the moving system has been chosen of hydraulic type to be strong enough and less space occupying. The velocity and displacement control can be done by means of electronic hydraulic valves with a satisfactory level of accuracy and reliability. The hydraulic pressure generator is installed out of the centrifuge basket to save a significant amount of space and is connected to the inside basket moving system by means of hydraulic pipe and rotary joints.

Figure 4. General View

1.2 Scaling laws

The scaling laws used for this modeling are indicated as below (Table 2).

Table 2. Scaling laws for centrifuge testing

Parameter Model / Prototype Dimensions

Length 1/N LStrain 1 1Stress 1 ML-1T-2Acceleration N LT-2Axial Rigidity 1/N2 MLT-2Flexural Rigidity 1/N4 ML3T-2

1.3 Soil properties

Soil material used in first test is chosen to be the granular soil of standard Firoozkouh 161 sand. Soil material used in second test is high porosity gravel with low density. The density

of low-density soil is equal to 50% of Firoozkouh soil density. (Table 3)

Table 3. Properties of Firoozkouh and low-density Soil

CcCuFCD50(mm)eminemaxGsSand type

0.88 2.58 1 % 0.27 0.548 0.874 2.65 Firoozkouh 161

--~0%3--1.3 Low-Density

1.4 Instrumentation

Two types of instruments containing strain gauge and linear variable differential transformers (LVDTs) were installed in the model. The strain gauges are installed in axial and circumferential directions on the

pipelines with the number of 26 in 7 stations. Strain gauges are placed in a way that axial and bending strains could be measures separately. Strain gauges are of the high strain type and are connected in the quarter bridge form.

Three LVDTs of the whole 5 ones are installed on the surface of the pipeline to record the deformation profile and the 2 other ones measure the axial displacement of the two endings of the pipeline. Apart from above, colorful grids were being used on the surface and between the soil layers.

2 RESULTS

Two tests were conducted in this study. In the first one, a stainless steel pipe with diameter of 8.0 mm and wall thickness of 0.4 mm which buried in Firoozkouh sand was subjected to a 70 mm reverse faulting with the acceleration of 40g. In the second experiment, the stainless steel pipe with 8.0 mm diameter and 0.4 mm wall thickness which buried in low-density gravel was subjected to the reverse faulting with 40g acceleration. The properties of model and prototype are indicated in Table 4.

Table 4. Properties of model/prototype for conducted tests

1st Test 2nd Test

Model Prototype Model Prototype

Pipeline Diameter (m) 0.008 0.320 0.008 0.32

Pipeleine WallThickness (m) 0.0004 0.016 0.0004 0.016

Faulting Magnitude (m) 0.070 2.8 0.070 2.8

Backfill Firoozkouh 161 High Porosity Sand (Low Density)

Faulting Type Reverse (60%) Reverse (60%)

Following figures illustrate the deformations of pipeline and soil during the faulting process. In Figs. 9 and 10 bending and axial strains before pipe failure versus distance from the faulting in 2nd test are presented.

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Figure 5. Surface Observation of 1st test

Figure 6. Surface Observation of 2nd test

Figure 7. Section Observation of 1st test

Figure 8. Section Observation of 2nd test

Figure 9. Bending strain during faulting- (Top: 1st test, Down: 2nd test)

Figure 10. Axial strain during faulting-(Top: 1st test, Down: 2nd test)

3 CONCLUSIONS

In this article the report of establishment of the first geotechnical centrifuge in Iran and its initial application in buried pipelines modeling subjected to faulting are presented. Also, a brief summary of the modeling details, related scaling laws and used facilities and instruments are described. Reported in this experimental study are the axial and bending strains diagrams of steel pipe versus distance from the normal faulting before pipe failure for the first time in the literature. Pipe failure happened almost at 3 cm in model or 1.2 m in prototype offset.

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The use of light weight material to fill pipeline trench is an affective technique to improve pipeline response to PGD. This technique changes the pipeline response from wrinkling to beam buckling which is a better deformation mechanism. In this mechanism the deformation of pipeline is distributed along the pipeline, despite of wrinkling mechanism which the deformation concentrated on two points. Also in beam buckling mechanism, the axial strain of pipeline is very small and the maximum bending moment reduced and transferred to the middle of pipeline. Choo et al. (2007) also investigated the use of light weight material (polystyrene blocks) to remediate the behavior of buried pipeline under normal faulting. They found that this technique improved the performance of pipeline under PGD condition.

4 REFERENCES

O’ Rourke, M., Gadicherla, V., and Abdoun, T., (2003). “Centrifuge Modelling of Buried Pipelines”, ASCE, Earthquake Engineering;

Rojhani, M., Ebrahimi, M.H., Moradi, M., and Ghalandarzadeh, A., (2010). “Building the Faulting Simulator Split Box for Geotechnical Centrifuge Modeling”, 4th International Congress of Geotechnical Engineering and Soil Mechanics in Iran, Tehran, Iran.

Taylor, R. N., (1995). Geotechnical Centrifuge Technology, Chapman & Hall Press.

Woo Choo, Y., Abdoun, T. H., O’Rourke, M., and Ha, D., (2007). “Remediation for buried pipeline systems under permanent ground deformation”, Soil Dynamics and Earthquake Engineering.

Wood, D. M., (2004

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Site investigation and geotechnical design strategy for offshore wind development

Investigation géotechniques et stratégie de conception pour le développement d’éoliennes maritimes

Muir Wood A. DONG Energy

Knight P. Parsons Brinckerhoff

ABSTRACT: The development of multi billion euro Offshore Wind Farms presents geotechnical engineers with the opportunity tocreate comprehensive detailed ground models incorporating a large variety of geotechnical hazards. However the political structure ofrenewable energy projects often leads to a fragmented development team, with no one party appointed for the whole design process.Inexperienced clients are often commissioning surveys because they think that they want to do a survey rather than for an engineeringreason. This leads to unclear specification, and a resulting survey that does not add the expected value to the project. The authorsdemonstrate in this paper how site investigation and ground modelling practices that are followed as routine in the design cycle ofonshore projects can be adapted and applied to add significant value to offshore renewable projects. This paper seeks to set out astructure for development of the ground model for offshore wind projects, and demonstrates how clients can ensure their surveys are adding value to the design strategy for their projects.

RÉSUMÉ : Grâce au développement très coûteux de fermes éoliennes en mer, les ingénieurs géotechniciens, ont l’opportunité de concevoir des modèles de sol détaillés et exhaustifs qui peuvent rendre compte d’un large éventail de risques géotechniques. Cependant la structure politique conduit au fait que les projets concernant les énergies renouvelables se trouvent bien souvent menéspar diverses équipes de conception sans véritable coordination. Des clients inexpérimentés ont souvent recours à des sondages dans leseul but de faire des sondages et non pour des raisons techniques. Cela conduit à avoir des cahiers des charges souvent imprécis quin’apportent rien à la valeur attendue du projet. Les auteurs démontrent dans cet article comment les études de terrain et les pratiquesde modélisation du terrain utilisées systématiquement dans les projets sur terre peuvent être adaptées et appliquées aux projets en meret peuvent augmenter leur valeur considérablement. Cet article propose une structure de développement des modèles terrestres pourles champs d’éolienne en mer et démontre comment les clients peuvent s’assurer que leurs sondages valorisent les stratégies de conception de leurs projets.

KEYWORDS: geotechnical hazard management, site investigation strategy, ground model development, offshore wind farm

1 INTRODUCTION

The offshore wind industry in Northern Europe started with the development of small demonstration projects. In recent years these have significantly increased in size, and many of the current projects are now multi billion euro investments, with development times in excess of five years. The industry can demonstrate many examples of how lessons learnt and knowledge gained from the earlier projects have been incorporated into the recent larger projects, ultimately leading to lower capital and operational expenditure per MW of power generated.

Through this process the design methodology and codes (e.g. DNV 2011) which were originally predominately based on the experience of offshore oil and gas infrastructure have also continued to develop and there are many examples of good practice in current projects. However, for geotechnical site investigations this learning process has not been completely positive. It leads to a tendency to base the scope of the investigation on a specification for a previous projects, rather than on what is most appropriate for the site and specific development.

The authors have been involved in the design of over 15 projects, which when built will total more than 8GW of power. A review of these projects shows that they have all spent comparable money on site investigation - typically approximately 1% of the project capital costs (which on the current large projects leads to investigations costing in excess of €30M). However the success of the investigations in managing

the geotechnical risks and bringing value to the projects is extremely variable.

2 EXAMPLE PROJECTS

Typical problems that the authors have experienced are: investigations not planned to mitigate project specific geotechnical hazards; poor recording and interpretation of geological information; planning of surveys not based on the results of proceeding investigations/studies; and surveys not specified by the foundation designers.

Some specific examples of good and poor practice on projects are detailed in the following sections. The project names have not been stated however the approximate construction cost of each project is given.

2.1 Poor practice

2.1.1 Project 1 – value ~€2bn The site investigation comprised site wide sub-bottom geophysical survey with three rounds of geotechnical works: (1) met mast borehole; (2) site wide boreholes; (3) seabed CPTs on every foundation location. Extensive advanced lab testing and down hole geophysical methods and pressuremeter tests were undertaken.

Issues included: the detailed designer was appointed after the site investigation was completed; insufficient time was allowed for the interpretation of the geophysical and geotechnical

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surveys; early CPT refusal and lack of boreholes on location meant that insufficient information was collected at many locations; and, the advanced cyclic testing undertaken was not optimised for the design load being applied to the soil.

The above issues meant that stratigraphy was not certain for the full foundation depth leading to very conservative design and uncertain driveability.

2.1.2 Project 2 – value ~€0.3bn The site investigation comprised a site wide sub-bottom geophysical survey with several rounds of geotechnical works comprising CPTs and boreholes.

The primary issue was the detailed designer being appointed after the site investigation was specified. The boreholes were not deep enough and insufficient information was gathered about the bedrock.

The developer had to commission another round of geotechnical investigation to gather further information for design and risk mitigation, which resulted in project delays and additional costs.

2.1.3 Project 3 – value ~€1.5bn The site investigation comprised a site wide sub-bottom geophysical survey with staged geotechnical works comprising CPTs and boreholes. Extensive advanced lab testing and down- hole geophysical methods and pressuremeter tests were undertaken.

Again, the detailed designer was appointed after site investigation had been undertaken, resulting in an insufficient number of boreholes to allow an efficient design.

2.1.4 Project 4 – value ~€0.1bn The site investigation comprised a site wide sub-bottom geophysical survey with staged geotechnical works comprising CPTs and boreholes.

The specification for the surveys was directly copied by the client from one of their previous projects in very different geology, and the detailed designer was appointed after the site investigations had been undertaken. This resulted in the geophysical techniques not being suitable for the geology and the developer had to repeat the geophysical survey to an enhanced specification.

2.2 Mediocre practice

2.2.1 Project 5 – value ~€1bn The site investigation comprised sub-bottom geophysical survey with one round of geotechnical investigation comprising full depth CPTs and boreholes.

The geophysics was incorrectly scoped, so when planning the geotechnical campaign, the detailed designer could not utilise the geophysics. The project programme was also compressed, meaning the geotechnical campaign was carried out in one stage.

More value, cost savings and mitigation of geotechnical risks could have been achieved with a collaborative geophysical and geotechnical interpretation and a staged geotechnical campaign with ground model development.

2.2.2 Project 6 – value ~€1bn The site investigation comprised sub-bottom geophysical survey with one round of geotechnical investigation comprising full depth CPTs and boreholes with geophysical borehole logging.

Again, the geotechnical campaign was carried out in one stage with the scope being developed whilst offshore. A more cost effective survey could have been achieved with a staged geotechnical campaign with ground model development.

2.3 Good practice

2.3.1 Project 7 – value ~€2bn The site investigation comprised sub-bottom geophysical survey with multi-stage geotechnical investigation comprising CPTs and boreholes. Extensive advanced laboratory testing was undertaken.

The foundation designer was involved in the specification of the surveys, and a ground model was developed based on collaborative geophysical and geotechnical interpretations. The geotechnical investigations were specified considering the confidence in the ground model and the foundation design.

2.3.2 Project 8 – value €1.5bn The site investigation comprised sub-bottom geophysical survey with multi-stage geotechnical investigation comprising CPTs and boreholes.

The early geophysics identified a considerable geological hazard which was successfully mitigated (Liingaard et al. 2012) through ground model development based on collaborative geophysical and geotechnical interpretations. A multi-stage geotechnical investigation allowed the detailed designer to develop a specific design method and scope the investigations to verify the variations from design codes.

3 SUGGESTED APPROACH

The above projects highlight the issues regarding quality and more importantly the effectiveness of the geotechnical investigations. There are common problems with scoping and management of site investigations. The offshore wind industry should look to the experience gained on projects in other fields of engineering, especially the lessons learnt by large onshore infrastructure projects. In particular they should: consider more formal approaches to the management of geotechnical risk; accept that a staged site investigation will deliver the most cost effective results; and, consider a contracting structure that brings the detailed designer onto the project from a very early stage.

3.1 Management of geotechnical risk

Managing ground risks solely through the traditional practice of thorough site investigation will not lead to the most cost effective project. Simply undertaking the most detailed site investigation that the budget will pay for leads to investigations being undertaken that are not specifically targeting the unknowns that are truly affecting the project, and also does not allow for the fact that the acceptable level of risk at construction is very different depending on the project developer and their attitude towards risk.

Developers should therefore seek to apply a more formal approach of managing geotechnical risk so that the site investigation specifically targets the unique project hazards. Clayton (Clayton, 2001) suggests formal processes for managing geotechnical risk, these have been adapted to form the recommended process detailed in Figure 1.

Figure 1 shows a design team led approach to planning, undertaking and reviewing effective site investigations. The design team identifies all hazards affecting the project (design, installation, operation and decommissioning). An assessment is made of the significance of those hazards in consultation with all project stakeholders. The site studies/investigations are planned to specifically target the hazards. The results are studied and the residual hazards reviewed. The consultation is repeated with the stakeholders and the need for further investigation evaluated.

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Figure 1. Risk based approach to managing geotechnical hazards

In managing this risk based approach it is essential that the designers appreciate that developers will have different acceptable levels of risk (Figure 2). This level of risk will depend on developers’ sources of funding, project timescale and contracting/procurement strategy. The acceptable level of risk will also change as a project develops.

Figure 2. Acceptable level of geotechnical risk to a developer

3.2 Staged site investigations

All of the projects reviewed in Section 2 had separate stages of site investigation comprising a geophysical followed by a geotechnical campaign. However the good practice projects had a multistage geotechnical campaign. Eurocode 7 (EC7) (BS EN 1997-2:2007) states that “Ground investigations should normally be performed in phases depending on the questions raised during planning, design and construction”. The code goes on to define two distinct phases of investigation – preliminary investigation and design investigation.

The multistage geotechnical investigation allows for a comparatively cheap ground-truthing geotechnical campaign early in the design process. This campaign gives the designer sufficient information to confirm the expected ground model and develop a preliminary design solution. The design investigation will then be significantly more effective, since it can utilise the most effective methods of investigation for the ground conditions, and fit the foundation type and dimensions that are being designed.

The strategy outlined in Figure 1 will be most effective when repeated several times; designers should be wary of trying to understand all ground based hazards in one stage of geotechnical investigation. A simple preliminary investigation

combined with the existing understanding of the site will often provide the best value to the developer at an early stage.

3.3 Develop contracting structure that allows early appointment of the designer

Many of the example projects detailed in Section 2 had ineffective site investigations because the detailed designer for the foundations was not appointed until after all site investigations had been completed. These projects appointed engineers solely to organise and supervise a single phase of site investigation who were not otherwise engaged in planning or design of the project. “This practice inhibits continuity or integration of project development, with the prospect that the site investigation will not be coupled with other features of the project planning process and will, in consequence, not provide adequate answers to vital questions and will not allow consideration of innovative methods of working”. This quote from Muir Wood (Muir Wood 2000) is just as applicable to offshore wind farm developments as it is to the tunnelling projects it was originally written to describe.

Instead developers should seek to engage a designer at the start of the development phase, who has the competence and ability to provide the geotechnical engineering for the whole project. This allows the foundation design and the scope of the site investigation to develop with the project, which mitigates project risks and delivers cost effective site investigations.

This is the approach taken by most infrastructure projects of equal size and complexity to these offshore wind developments. As these projects have demonstrated, this does not prevent the use of different project procurement mechanisms, provided that mechanisms such as novation are available to the developer.

4 APPLIED EXAMPLE

The strategy detailed in Section 3 is being applied to the first project of the Hornsea Round 3 development area as follows.

4.1 Engineering contract setup

For this project, the developer – DONG Energy is using in-house engineers to undertake the geotechnical design and engineering. This capability is supplemented by the use of additional consultants appointed to the projects for long term positions. This setup gives an integrated approach to engineering and development. Different parts of the project will have different procurement strategies; where the developer chooses not to progress the detailed design in house the design for that element will be contracted to another party, with the original in-house design team maintained in a reviewing capacity.

4.2 Desk study

Prior to the desk study, no geotechnical design parameters were available. The principal geotechnical risks at this stage are shown in Table 1. At this point, the site investigations cannot be specified as the foundation type, ground conditions and design requirements are not understood.

Table 1. Design parameters and primary geotechnical risks prior to desk tudy s

Design parameter Geotechnical risks

None Unknown site constraints

Unknown design constraints

Unknown construction constraints

Following completion of a desk study which includes a review of geological information, design requirements and possible foundation types and construction considerations,

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geotechnical design information is available and project specific geotechnical risks can be identified, as shown in Table 2. At this point the design parameters have been selected based on geological memoirs and nearby site investigations. The undrained shear strength of the main clay unit has been used as an example to show how the design inputs change as more information is collected.

Table 2. Design undrained shear strength of main clay unit and primary eotechnical risks following desk study g


125 to 350 kPa No wind farm specific ground information

Presence of regional geohazards

Several foundation types possible

The design information and project risks are reviewed at this stage. Referring to Figure 2, this level of design information and risk may be acceptable to some clients, in which case the design and construction risks will be managed with a conservative and uncertain design, or passed onto a third party with large financial consequences. However in this case, to confirm the ground conditions and add confidence to the design, site investigation works were specified.

4.3 Geophysical survey

A site wide sub-bottom geophysical survey was carried out. The primary objectives were to: identify the geological units present; identify any unexpected geological features; and, collect information to allow preliminary geotechnical boreholes to be positioned.

4.4 Preliminary geotechnical survey

A preliminary geotechnical survey consisting of seabed CPT and composite CPT and sampling boreholes was carried out. The primary objectives were to: validate the desk study and geophysical survey; collect information for ground model development including in situ testing and sampling in the primary units; and, identify unexpected ground conditions and risks.

Following completion of the preliminary geotechnical site investigation, the results were interpreted and a ground model produced.

4.5 Ground model development

The ground model was developed using a collaborative geophysical and geotechnical approach. The engineering units were identified and geotechnical parameters selected using the CPT and laboratory testing results. Wind farm specific ground information is now known, and the regional geohazards understood, allowing preliminary design to commence. The design parameters and risks (Table 3) have been refined, resulting in a more efficient foundation.

Table 3. Design undrained shear strength of main clay unit and primary geotechnical risks following preliminary site investigation and ground

odel development m


Low: 145 kPa Local variation in site geology

Mean: 240 kPa Advanced design parameters unknown

High: 325 kPa

4.6 Design team and stakeholder review

The confidence and expected accuracy of the geotechnical interpretation and risks were reviewed by the design team and stakeholders. It was decided that there was sufficient knowledge

of the site to progress the design and construction planning and manage the geotechnical risks.

Some of the project infrastructure will be designed and developed by others. There is now sufficient knowledge of the site to be able to define a contract ground model baseline, and enable the risk to third parties of unknown ground conditions to be appropriately handled.

4.7 Future geotechnical survey

The residual project risks are not acceptable for the finalisation of detailed design. To obtain this information, a small number of sampling boreholes (to obtain samples for advanced laboratory testing) complemented by one CPT to the expected foundation depth (to identify local variability and confirm the units present) will be carried out. Subsequent investigation may be required depending on the results of this investigation and the developer’s acceptable risk.

5 CONCLUSIONS

Offshore wind projects in their early stages present considerable challenges to engineers managing the geotechnical hazards given the large complexity, cost and time scale of the projects.

A review of the site investigations undertaken for various projects highlights repeated mistakes leading to inefficient site investigations; namely the site investigations not being planned to mitigate project specific geotechnical hazards and surveys not being specified by the foundation designers.

Lessons must be learnt from the wider construction industry, including: the need for formal approaches to the management of geotechnical risk; acceptance that a staged site investigation will deliver the most cost effective results; and, considering a contracting structure that brings the detailed designer onto the project from a very early stage.

The authors have proposed an engineer led approach for managing geotechnical risks, where the design team and stakeholders are actively involved in assessing and mitigating the geotechnical risks, allowing the geotechnical foundation design and site investigation to develop with the project and ensuring that survey work is specified to directly control project risks.

These principles are being applied to the first project of the Hornsea Round 3 development area, where design and construction risks are being mitigated early with a relatively small amount of site investigation works.

6 REFERENCES

BS EN 1997-2:2007. Eurocode 7 Geotechnical design – Part 2: Ground Investigation and testing

Clayton C.R.I. 2001. Managing geotechnical risk: improving productivity in UK building and construction. Institution of Civil Engineers and Thomas Telford, London.

Det Norske Veritas AS. 2011. Offshore Standard DNV-OS-J101 Design of Offshore Wind Turbine Structures. Det Norske Veritas, Oslo.

Liingaard M.A. Mygind M. Thomas S. Clare M. and Pickles A. 2012 Evidence of tertiary intrusive rock at the West of Duddon sands offshore wind farm. Society for Underwater Technology Proceedings of the 7th International Conference, London 2012, 145-152.

Muir Wood A.M. 2001. Tunnelling: management by design. E&FN Spon, London.

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Diagrammes de stabilité cyclique de pieux dans les sables

Cyclic stability diagrams for piles in sands

Puech A., Benzaria O. Fugro GeoConsulting, Nanterre, France

Thorel L., Garnier J. IFSTTAR, Nantes, France

Foray P., Silva M. 3S-R, Grenoble, France

Jardine R. Imperial College, London, UK

RÉSUMÉ: Cette communication rassemble des diagrammes de stabilité cyclique obtenus dans des sables siliceux denses et par desmoyens expérimentaux variés : essais in situ sur pieux réels, essais sur pieux modèles en grande chambre d’étalonnage et essais sur pieux modèles centrifugés. Elle couvre le cas des pieux battus en traction et celui des pieux forés en compression. Les diagrammescycliques de stabilité sont des outils précieux pour une première estimation de l’effet des chargements cycliques sur le comportement axial des pieux.

ABSTRACT: This paper gathers cyclic stability diagrams obtained from various experimental sources: in situ tests on actual piles,laboratory tests on model piles in a large calibration chamber and model piles in a centrifuge. Driven piles in tension and bored pilesin compression are addressed. Cyclic stability diagrams are useful tools for a preliminary assessment of the effects of cyclic loadingson the behaviour of piles.

MOTS-CLÉS: pieu battu, pieu foré, chargement cyclique axial, diagramme de stabilité cyclique

KEYWORDS: driven pile, bored pile, axial cyclic loading, cyclic stability diagrams

1 INTRODUCTION

Le concept de diagramme de stabilité cyclique pour représenter de manière synthétique la réponse des pieux soumis à des chargements cycliques axiaux a été introduit dans les années 80 par Karlsrud et al. (1986) pour les argiles et par Poulos (1988) pour les sables.

Ce concept se révèle particulièrement utile pour juger en première estimation de l’effet potentiel des chargements cycliques sur la réponse des pieux (Jardine et al., 2012)

On dispose en pratique de peu d’éléments dans les sables. La présente communication propose des diagrammes cycliques applicables aux pieux battus et aux pieux forés dans les sables denses.

2 ESSAIS EN TRACTION

2.1 Essais de Dunkerque (ICL)

Sept pieux tubulaires en acier à base ouverte non instrumentés ont été testés par l’Imperial College de Londres (ICL) sur un site de sable marin dans le secteur Industriel Ouest du port de Dunkerque, France (Jardine & Standing, 2000). Le profil du sol est caractérisé par 3 m de remblai hydraulique sur du sable des Flandres. Le sable est composé principalement de quartz (84%), d’albite et microcline (8%) et de débris de coquillages CaCO3 (8%). Les profils pénétrométriques (qc) du site varient entre 10 et 35 MPa selon la profondeur et l’emplacement. La densité relative est en moyenne d’environ 75%. Des essais de cisaillement direct et triaxiaux indiquent un angle de frottement de pic φ΄ de 35 à 40º et une valeur d’état critique φ’cv ~ 32º. Des renseignements supplémentaires sur les caractéristiques du site et les essais de laboratoire réalisés sont présentés par Jardine et al. (2006).

Les essais cycliques ont utilisé les installations du projet GOPAL (Parker et al., 1999). Au total 21 essais statiques et 14 essais cycliques ont été effectués sur les six pieux de réaction du projet GOPAL. Ces pieux ont été battus jusqu’à une profondeur d’environ 19 m avec une relation espacement/diamètre du pieu d’environ 15. Leur diamètre est de 457 mm.

Plusieurs séries de chargements cycliques ont été appliquées, entrecoupées d’essais statiques référentiels en traction, la plupart d’entre eux avec des cycles uniquement en traction pour mieux individualiser la distribution des charges entre la pointe et le frottement latéral.

La réponse des pieux au chargement cyclique est décrite dans Jardine & Standing, 2012 ; Tsuha et al., 2012; Rimoy et al., 2013.

2.2 Essais en chambre d’étalonnage (3SR-ICL-SOLCYP)

Dans le cadre d’une coopération entre l’Imperial College de Londres (ICL), le Laboratoire 3SR de l’Université de Grenoble et le Projet SOLCYP, plusieurs séries d’essais on été réalisées sur un pieu modèle instrumenté dans la chambre d’étalonnage du laboratoire 3SR de Grenoble. L’objectif initial était de tenter de reproduire à une échelle de laboratoire et dans un environnement contrôlé les résultats obtenus par Jardine et Standing, 2000 sur les pieux battus dans le sable de Dunkerque.

La chambre de calibration a une hauteur de 1,5m et un diamètre de 1,2m. Dans sa paroi intérieure, une membrane en latex et une couche de graisse en silicone ont été mises comme système d’anti frettage pour avoir un meilleur contrôle des conditions Ko du sol. Un système d’isolation permet d’effectuer les essais à une température constante comprise entre 18 et 19º.

Le sable utilisé est le sable de Fontainebleau NE34, (d50 = 0,2 mm, max = 17,2 kN/m3 et min = 14,2 kN/m3). Pour l’ensemble des essais présentés, ce sable été mis en place par pluviation en obtenant un indice de densité relative compris entre 0,65 et 0,70. Le massif de sable est soumis à une pression

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verticale appliquée sur la surface du massif de 150 kPa, correspondant à une résistance de pointe pénétrométrique qc de 20 à 23 MPa.

Le pieu modèle instrumenté a été développé par ICL et est décrit en détail dans Jardine et al (2009). C’est un pieu à base fermée de 36 mm de diamètre, instrumenté en 3 niveaux le long du fût pour mesurer les contraintes tangentielles et radiales dans la surface du pieu. Chaque niveau d’instrumentation inclut aussi un capteur de force. Pour les derniers essais réalisés, il a été équipé d’un capteur de force en pointe.

L’installation initiale du pieu a été effectuée jusqu’à une profondeur d’environ 1 m par des cycles d’enfoncement successifs de 5 à 20 mm à une vitesse de 0,2 mm/s, suivis d’une décharge complète, de façon à simuler les effets d’un processus de battage.

Le programme expérimental a inclus quatorze essais cycliques autant sous contrôle de déplacement que sous contrôle en charge pour des cycles alternés et non-alternés (seulement en traction).

Une première partie des résultats concernant les essais de chargement cyclique a été présentée par Tsuha et al. (2012). On trouvera de plus amples informations dans Rimoy et al., 2013

2.3 Critères de rupture

L’application de cycles sur un pieu installé dans du sable provoque une succession de petits glissements relatifs sol-pieu dont le cumul détermine le déplacement global. La vitesse de déplacement initiale est fonction de l’amplitude Qc et du niveau de chargement maximal Qmax mais ces mêmes paramètres conditionnent également l’évolution du frottement qui peut se détériorer pour des cycles de grande amplitude ou s’améliorer pour des cycles de faible amplitude (Tsuha et al., 2012). Sur un pieu sollicité en traction, la vitesse initiale se modifie pour : - soit conduire vers la rupture : dans ce cas la vitesse

s’accroît. La rupture peut être définie de manière conventionnelle (par exemple pour un déplacement de la tête du pieu de 0,1D) ou lorsque se produit une accélération brutale du taux de déplacement ;

- soit conduire vers la stabilisation : la vitesse décroît continument jusqu’ à passer en dessous d’un seuil où on peut considérer que les déplacements cumulés deviennent non significatifs.

2.4 Diagrammes de stabilité cyclique

La caractérisation complète d’un essai de chargement cyclique suppose la définition des paramètres suivants :

Qm: valeur moyenne de la charge sous chargement cyclique, Qc : demi-amplitude du chargement cyclique, Nf : nombre de cycles conduisant à la rupture, N : nombre de cycles appliqués en l’absence de rupture, f : fréquence des cycles (en général 0,5Hz)

L’essai est dit répété (one-way) si Qc< Qm et alterné (two-way) si Qc>Qm. On définit de plus :

Qu : capacité statique ultime selon le mode considéré (Qut en traction et Quc en compression).

Il est pratique de présenter les résultats d’essais de chargements cycliques dans un diagramme où chaque série de chargement est identifiée par le couple de paramètres normalisés Qm/Qu et Qc/Qu. Ce type de représentation permet de bien visualiser les zones de fonctionnement sous chargements répétés et sous chargements alternés. Si on affecte chaque point du nombre de cycles Nf ayant provoqué la rupture ou du nombre de cycles N appliqué sans provoquer la rupture on peut définir des zones de « stabilité » ou d’« instabilité » du pieu. On voit que la taille de ces zones dépend du (des) critère(s) de rupture choisis.

On reproduit sur les Figures 1 et 2 les diagrammes cycliques obtenus pour les essais de Dunkerque (Jardine and Standing,

2012) et pour les essais en chambre de calibration (Tsuha et al., 2012).

-0,2 0,0 0,2 0,4 0,6 0,8 1,00,0

0,2

0,4

0,6

0,8

1,0 No cyclic failure First failure Cyclic failure after previous cyclic or static failure

3

20624

13

1

1241 1

927

345

>221>200

Qcy

clic/Q

TQmean/QT

>1000

Unstable

Metastable

Stable

Figure 1 : Diagramme de stabilité cyclique des pieux battus de Dunkerque (d’après Jardine and Standing, 2012)

-0,2 0,0 0,2 0,4 0,6 0,8 1,00,0

0,2

0,4

0,6

0,8

1,0

One way

Qcy

clic/Q

T

Qmean/QT

Two way

Stable

Unstable

>1000

Nf=1

10100

1410

4

1000

500

5

Nf=number of cycles to failure

Meta-Stable

66165

580

Figure 2 : Diagramme de stabilité cyclique des pieux modèles en chambre de calibration (d’après Tsuha et al., 2012)

Pour ces deux diagrammes, la rupture est atteinte lorsque le déplacement de la tête du pieu atteint 0,1D. La zone instable caractérise les essais ayant atteint le critère de rupture avant 100 cycles. La zone stable correspond à une zone de faible amplitude de chargement cyclique dans laquelle les pieux ont été soumis à plus de 1000 cycles sans accumuler de déplacements significatifs (pour les deux types de pieux) ou pour lesquels la vitesse de déplacement était inférieure à 1mm pour 1000 cycles (pieux de Dunkerque). Entre ces deux zones se situe une zone qualifiée de métastable dans laquelle les pieux atteignent la rupture entre 100 et 1000 cycles ou développent des taux de déplacement pouvant faire craindre des ruptures au-delà de 1000 cycles.

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Il est intéressant de mentionner que les trois zones de stabilité resteraient inchangées si on adoptait un critère de rupture à 0,03 D comme pour les pieux en compression ci-après.

3 ESSAIS EN COMPRESSION

3.1 Essais de Loon-Plage (SOLCYP)

Dans le cadre du projet national SOLCYP (Puech et al., 2012 des essais sur pieux réels ont été conduits sur le site de Loon-Plage constitué de sables denses. Cinq pieux forés et deux pieux métalliques battus ont été installés et soumis à des séries de chargements statiques et cycliques. Les résultats obtenus sur les pieux forés sont présentés dans Benzaria et al. (2013).

Le site expérimental se situe sur la commune de Loon-Plage (59) près de Dunkerque dans le Nord de la France. Il se caractérise par une couverture de remblais récents (0-0,6m) et d’argile sableuse (0,6-2,2m) sous laquelle on rencontre la formation de sable des Flandres.

Le sable est un sable siliceux très fin (D50 voisin de 0,15mm) et mal gradué (coefficient d’uniformité CU=0,98) très proche de celui rencontré sur le site voisin des essais ICL (même origine). La formation est latéralement homogène et se caractérise par des valeurs de résistance au cône qn croissant de 5 à 40 MPa vers 8 m de profondeur pour se stabiliser ensuite entre 30 et 50 MPa jusque vers 11,5m. L’interprétation des CPT conduit à un indice de densité ID compris entre 0,7 et 0,9 (sable dense à très dense).

Une série d’essais triaxiaux monotones a donné un angle de frottement interne φ’cv voisin de 31° en bon accord avec les valeurs trouvées sur le sable de Dunkerque (Jardine et Standing, 2000)

Les deux pieux F4 et F5 sont géométriquement identiques (D=420mm, fiche 8m). Ils ont été exécutés à l’aide d’une tarière à axe creux vissée dans le sol sans extraction notable de matériau puis extraite sans dévissage tandis que le béton est déversé simultanément par l'axe creux. Les pieux sont équipés d’un train d’extensomètres amovibles de type LCPC introduits dans un tube de réservation positionné entre les armatures.

Les pieux ont été testés trois mois environ après leur mise en place. Les programmes de chargement comportaient des essais statiques de référence à paliers d’une heure selon la norme NF P 94-150, des essais de chargement rapides (réduction des paliers à 3mn) et des essais de chargement cycliques axiaux de type répété à la fréquence de 0.5Hz. Une description plus précise des modes de chargement est indiquée dans Benzaria et al. (2012).

3.2 Essais sur modèles réduits centrifugés (SOLCYP)

Une campagne d’expérimentations sur modèles réduits centrifugés a été réalisée sur la centrifugeuse géotechnique de l’IFSTTAR à Nantes (Guefrech et al., 2012). Les pieux d’élancement 31 sont réalisés à l’échelle 1/23ème. Leur diamètre est de 18 mm. Leur surface est parfaitement rugueuse. Ils sont mis en place selon un procédé non refoulant consistant essentiellement à mettre le sable en place par pluviation alors que le pieu est déjà pré-positionné dans le conteneur. Cette technique simule un pieu moulé en place comme les pieux forés à la tarière creuse du site de Loon-Plage.

Le sable de Fontainebleau NE34 sec est en tout point identique à celui utilisé au laboratoire 3SR pour les essais en chambre d’étalonnage et présente des propriétés physiques et mécaniques très voisines de celle du sable des Flandres.

On s’intéresse dans ce qui suit à une série d’essais réalisés dans un massif à forte densité (ID ~ 0.7) à la fréquence de 1Hz. Seuls les essais de chargement cyclique en compression répétée sont analysés.

3.3 Critère de rupture

La définition de critères de rupture en compression est plus délicate qu’en traction. En effet, quelle que soit la vitesse de déplacement initiale (sur les premiers cycles), les déplacements tendent globalement vers la stabilisation. Cette observation est commune aux pieux in situ (Benzaria et al., 2013) et aux pieux modèles (Guefrech et al.,2012). En effet, même s’il y a dégradation rapide du frottement, le déplacement du pieu provoque une mobilisation progressive de l’effort de pointe qui ralentit progressivement les tassements.

Le critère de rupture ne peut alors être défini de manière conventionnelle (e.g. 0,1D) mais doit s’exprimer en termes de déplacement cyclique acceptable. Ce critère pourra être franchi sur les tous premiers cycles en cas de chargement très sévère (avec une vitesse de déplacement forte voire croissante) mais plus généralement au bout d’un nombre de cycles plus ou moins important et avec une vitesse de déplacement décroissante.

Figure 3 : Vitesses de déplacements des pieux en compression pour les essais en centrifugeuse

L’analyse des vitesses de déplacement effectuée sur les

pieux modèles centrifugés illustre les phénomènes en jeu (Figure 3). On distingue trois familles d’essai : a) les essais pour lesquels un tassement de 0,1D est atteint en moins de 500 cycles et qui présentent une décroissance permanente de la vitesse de tassement ; b) ceux pour lesquels ce tassement est atteint entre 1000 et 5000 cycles avec une vitesse de tassement qui semble se stabiliser ; c) ceux pour lesquels la vitesse de tassement devient rapidement très faible (< 0,5 mm pour 1000 cycles). L’évolution vers des déplacements importants est alors improbable.

Figure 4 : Vitesses de déplacement des pieux en compression pour les essais de Loon-Plage

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La Figure 4 montre le même type d’évolution des vitesses de déplacement pour les pieux de Loon-Plage.

3.4 Diagrammes de stabilité cyclique

On présente sur les Figures 5 et 6 les diagrammes de stabilité cyclique obtenus pour les essais cycliques en compression sur pieux forés de Loon-Plage et les pieux moulés en centrifugeuse.

Pour ces deux diagrammes, la rupture est définie pour un déplacement du pieu de 0,03 D. La zone instable caractérise les essais ayant atteint le critère de rupture avant 100 cycles. La zone stable correspond à une zone de faible amplitude de chargement cyclique dans laquelle les pieux n’ont pas atteint le critère de rupture et les vitesses de déplacement sont faibles. Entre ces deux zones se situe la zone qualifiée de métastable dans laquelle les pieux atteignent la rupture entre 100 et 1000 cycles.

Figure 5 : Diagramme de stabilité cyclique pour les pieux forés de Loon-Plage

Figure 6 : Diagramme de stabilité cyclique pour les pieux moulés en centrifugeuse

On constate une bonne concordance entre les deux diagrammes. Il est recommandé de ne pas extrapoler les données dans le domaine des essais alternés. Les données disponibles (non montrées ici) indiquent en effet une forte réduction des zones métastable et stable dans le domaine alterné.

4 CONCLUSION

Cette communication rassemble des diagrammes de stabilité cyclique obtenus dans des sables siliceux denses et par des

moyens variés : essais in situ sur pieux réels, essais sur pieux modèles en chambre d’étalonnage et essais sur pieux modèles centrifugés. Elle couvre le cas des pieux battus en traction et celui des pieux forés en compression.

Les diagrammes cycliques de stabilité sont des outils précieux pour juger de la sévérité des chargements cycliques sur le comportement axial des pieux.

L’attention est attirée sur la difficulté de définir des critères de rupture significatifs, notamment en compression. L’interprétation des diagrammes ne peut être dissociée des critères choisis pour les élaborer.

5 REMERCIEMENTS

La rédaction de cette communication a été rendue possible par la collaboration de nombreux chercheurs et organismes. Les auteurs remercient tous leurs collègues impliqués dans ces recherches et les différents organismes ayant autorisé la publication des résultats.

6 REFERENCES

AFNOR.1999. NF P 94-150. Norme Française. Sols: Reconnaissance et Essais – Essai statique de pieu sous effort axial – Partie 1: en compression et Partie 2: en traction

Benzaria O., Puech A., and Le Kouby A. 2012. Cyclic axial load-tests on driven piles in overconsolidated clay, Offshore Site Investigation and Geotechnics, SUT, London

Benzaria A., Puech A. et Le Kouby A. 2013. Essais cycliques axiaux sur des pieux forés dans des sables denses. Proceedings 18th ICSMGE, Paris

Guefrech A., Rault G, Chenaf N., Thorel L., Garnier J., Puech A. 2012 Stability of cast in place piles in sand under axial cyclic loading . Proc. 7th Int. Conf. Offshore Site investigation and Geotechnics. London. 12-14 sept. pp.329-334.

Jardine, R.J. and Standing, J.R. 2000. Pile load testing performed for HSE cyclic loading study at Dunkirk, France. Two Volumes. Offshore Technology Report OTO 2000 007; Health and Safety Executive, London. 60p and 200p.

.Jardine, R.J., Standing, J.R. & Chow, F.C. 2006. Some observations of the effects of time on the capacity of piles driven in sand. Geotechnique, 56 (4), 227-244.

Jardine, R., Bitang, Z., Foray, P., & Dalton, C. 2009. Experimental Arrangements for Investigation of Soil Stresses Developed around a Displacement Pile. Soils and Foundations, 49(5), 661–673.

Jardine, R.J. and Standing. 2012. Field axial cyclic loading experiments on piles driven in sand. Soils and Foundations, 52(4), 723–736.

Jardine R, Puech A and Andersen K. 2012. Cyclic loading of offshore piles: potential effects and practical design. Proc.7th Int. Conf. on Offshore Site Investigations and Geotechnics, SUT,. London.

Karlsrud K., Nadim F. and Haugen, T. 1986. Piles in clay under cyclic axial loading - Field tests and computational modeling. Proc. 3rd

Int. Conf. on Numerical Methods in Offshore Piling, Nantes, France Parker, E. J., Jardine, R.J., Standing, J.R. and Xavier, J. 1999, Jet

grouting to improve offshore pile capacity. Offshore Technology Conference, Houston, OTC 10828.

Poulos H.G. 1988 Cyclic stability diagram for axially loaded piles. Journal of Geot. and Geoenv. Eng. 114 (8): 877-895 .

Puech A., Canou J., Bernardini C., Pecker A., Jardine R., and Holeyman A. 2012. SOLCYP: a four year JIP on the behavior of piles under cyclic loading. Offshore Site Investigation and Geotechnics, SUT, London

Rimoy S., Jardine R. and Standing J. 2013. Displacement response to axial cyclic loading of driven piles in sand. Proceedings 18th ICSMGE, Paris

Tsuha, C., Foray, P., Jardine, R., Z.X., Y., Silva, M., & Rimoy, S. 2012. Behaviour of displacement piles in sand under cyclic axial loading. Soils and foundations, 52(3), 393–410.

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Utilisation des essais d'expansion cyclique pour définir des modules élastiques en petites déformations

Determining small strain elastic modulus using cyclic expansion tests

Reiffsteck P., Fanelli S., Tacita J.-L. Univ Paris Est, IFSTTAR GER, Paris, France

Dupla J.-C. Univ Paris Est, Marne-la-Vallée, Navier Géotechnique

Desanneaux G. CETE de l’Ouest, LRPC Saint Brieuc, France

RÉSUMÉ : Depuis trente ans, la réalisation d'essais d'expansion cycliques réalisés en trous préforés ou forés à l'avancement surdifférents sites expérimentaux a permis de disposer d'une base assez importante de cas. La qualité de ces résultats permet de dériver des paramètres de déformabilité à des taux de déformation faibles. Ces essais cycliques ont été réalisés au pressiomètre Ménard et au pressiomètre autoforeur. Un peu moins d'une dizaine de sites ont été étudiés permettant d’observer le comportement de matériauxsableux et argileux normalement consolidés et surconsolidés. Cette communication présente le matériel utilisé ainsi que les procédures suivies. Le programme d'essai composé de plusieurs phases de cycles d’amplitude variable a été proposé dans les annéesquatre-vingt. On observe une évolution du module en fonction du nombre de cycles, de la nature du matériau et du rapport de l'amplitude et de la position moyenne avec la contrainte horizontale en place. Une synthèse des résultats des essais obtenus sur cessites est présentée.

ABSTRACT: For thirty years, realization of cyclic expansion tests carried out in borehole pockets drilled using a separate tool or integrate in the probe, on different experimental sites allowed to have a rich database. The quality of these results allows to derivestress-strain parameters at low strain level. These cyclic tests were carried out using Menard and self-boring pressuremeters. A little less than ten sites were studied covering the behavior of sandy and clayey materials normally consolidated or overconsolidated. Thispaper will present the equipment used and procedures applied. Usually, the test program, which was proposed in the eighties, consists of several phases of cycles of variable amplitude. One can observe a shift of the module depending on the number of cycles, the nature of the material and the ratio of the amplitude and the mean position compared to the at rest horizontal stress. A summary of testresults obtained on these sites is presented.

MOTS-CLÉS : comportement cyclique, essais d’expansion de cavité, pressiomètre autoforeur, pressiomètre Ménard KEYWORDS: cyclic behavior, cavity expansion test, Self-boring pressuremeter, Ménard pressuremeter.

1 INTRODUCTION

Avec le matériel d’essai pressiométrique Ménard, il est possible de réaliser des essais d’expansion par palier (norme NF P94-110-1) et des essais cycliques (norme NF P94-110-2) (AFNOR, 1999 et 2000). Ces derniers essais comportent un cycle réalisé par paliers, dans les mêmes conditions que l’essai pressiométrique Ménard objet de la norme NF P 94-110-1. L’essai d’expansion classique, dans les conditions de forages préconisées par la norme NF P 94-110-1 et avec le protocole de chargement proposé, ne donne pas de résultats utilisables directement dans une étude de la déformabilité des ouvrages notamment lorsque la connaissance des modules en petite déformation est nécessaire (Combarieu et Canépa, 2001).

Les essais avec boucle de déchargement-rechargement permettent de déterminer un module cyclique de déformation. Les valeurs obtenues sont intermédiaires entre les modules en petites déformations obtenus au laboratoire, ou avec des essais de propagation d’ondes in situ, et les modules Ménard usuels (Les pressiomètres Louis Menard, 1960, Borel et Reiffsteck, 2006). Toutefois, un seul cycle est insuffisant pour cerner l’évolution des caractéristiques du sol sous chargement cyclique (Dupla et Canou, 2003). L’étude présentée, réalisée dans le cadre du projet national SOLCYP (projet de recherche sur le comportement des pieux soumis à des sollicitations cycliques, voir le site www.pnsolcyp.org pour plus d’information), comporte des essais multi-cycles réalisés avec la technique de la sonde mise en place dans un trou pré-foré ou foré à l’avancement.

2 DISPOSITIF EXPÉRIMENTAL

Le principe de l’essai consiste à mesurer l’évolution du volume injecté lors de l’application de cycles de pression.

2.1 Matériel

La mesure de la variation de volume en fonction des cycles se fait soit par mesure du volume d’eau injecté, soit par mesure du déplacement d’un palpeur (Figure 1a et b). L’idée étant de pouvoir réaliser des essais avec une sonde pressiométrique mise en œuvre par autoforage ou dans un pré-forage de type Ménard (AFNOR, 2000).

Le matériel utilisé développé par l’entreprise Jean Lutz SA est un contrôleur pression volume (CPV) (de type PREVO), capable de piloter des électrovannes par un ordinateur de type PC via une application spécifique.

Utilisation des essais d'expansion cyclique pour définir des modules élastiques en petites déformations

Determining small strain elastic modulus using cyclic expansion tests

Reiffsteck P., Fanelli S., Tacita J.-L. Univ Paris Est, IFSTTAR GER, Paris, France

Dupla J.-C. Univ Paris Est, Marne-la-Vallée, Navier Géotechnique

Desanneaux G. CETE de l’Ouest, LRPC Saint Brieuc, France

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a)

cellule de garde

cellule de garde

cellule de mesure

gazeau

PC

b)

cellule de garde

cellule de garde

cellule de mesure

gaz

PCacquisition

Figure 1. Architecture de l’essai cyclique (a) au pressiomètre Ménard, (b) au pressiomètre autoforeur

Le principe de fonctionnement est le suivant. Les différentes opérations manuelles sont réalisées, soit directement sur le CPV, soit par le programme. Le pilotage en cyclique est réalisé sur la base d’un fichier d’essai acceptant tout type de signal, harmonique ou multifréquence. Le suivi se fait en temps réel sur un onglet graphique ou sur le tableau de valeur.

2.2 Méthode d’essai

Durant les années 70, l’Association pour la Recherche en Géotechnique Marine rassemblant différentes entreprises, bureaux d’étude et établissements de recherche dans le domaine a mené une campagne, sur plusieurs sites, d’essais cycliques au pressiomètre. Les détails des expérimentations sont rassemblés dans plusieurs rapports et articles du Symposium sur la pressiométrie et ses applications en mer tenu en 1982 à Paris (Jézéquel et le Méhauté, 1982 ; Puech et al., 1982). Trois types d’essais furent réalisés ; nous présentons les deux mis en œuvre dans la présente étude :

- essai de chargement cyclique entre deux bornes de pression pM et pm (Figure 2 a),

- essais de chargement entre deux bornes de pression variables, dont la moyenne est cependant constante, la borne inférieure restant supérieure à po la pression des terre au repos (Figure 2 b).

Figure 2. Différents types d’essais cycliques

Les paramètres retenus pour les essais cycliques découlent de la méthode proposée (cf. ci avant) : essai à pression contrôlée ; adaptation de la fréquence au type de sol pour tenter de rester drainé, niveau de sollicitation : 0,8 ( Rc = ∆pcyc /p’0) ; fréquence : 0,01 à 0,05 Hz et nombre de cycles égal à 50.

La pression initiale pm utilisée pour démarrer l’essai est définie comme la contrainte horizontale en place (pression des terres de repos) (effective de préférence) et la pression maximale pM est égale à (1+0,8)pm (Dupla et Canou, 2003). La pression pm, qui a été prise égale à p0, a été définie à partir des résultats d’essais d’expansion de type Ménard antérieurs par la méthode proposée par Briaud (1992).

2.3 Définition des modules

L’intérêt de réaliser des cycles avec le pressiomètre pour obtenir des modules en petites déformations est apparu très tôt (Les pressiomètres Louis Menard, 1960). Dès l’origine, plusieurs modules ont été définis à partir des courbes expérimentales.

Dans la première zone désignée comme « élastique », le module atteint une valeur quasi indépendante du niveau de déformation. Il est appelé module « initial » G0.

Les courbes en partie monotone sont décrites par un module « sécant » (Gs,1) défini par la pente de la droite reliant l’origine au point actuel et en partie cyclique, un autre module sécant (Gp,N) déterminé par la pente de la droite reliant les deux points d’inversion du cycle N. Les modules maxima des cycles sont calculés avec la relation (avec les notations de la Figure 3) :

La boucle parcourue dans ces séquences de

déchargement/remise en charge est de forme ellipsoïdale. Elle représente l’énergie dissipée en déformation plastique. L’évolution de l’inclinaison des cycles ou module au cours des cycles permettra d’évaluer le comportement du sol. On peut évaluer le durcissement ou l’adoucissement cyclique et l’accumulation de déformation, la stabilisation ou la relaxation ou l’effet rochet.

Figure 3. Calcul des modules des cycles

Le mode d’interprétation est basé sur l’évolution de l’aire caractéristique des boucles de chargement déchargement ainsi que du module sécant des boucles d’hystérésis (Figure 3).

3 ESSAIS CYCLIQUES

3.1 Essais du LRPC de Saint Brieuc

À la fin des années 70, le laboratoire des Ponts et Chaussée de Saint Brieuc a réalisé plusieurs campagnes d’essais d’expansion cyclique au pressiomètre autoforeur (modèle PAF76 de diamètre 132 mm). Ces essais comptaient une centaine de cycles voire plusieurs milliers de cycles (durée de 24 à 72 heures). Ils ont été réalisés sur deux sites principaux Cran et Plancoët (Le Méhauté et Jézequel, 1980).

3.1.1 Site de Plancoët Le site est constitué d’une parcelle plane en bordure de la rivière Arguenon. Le sol est constitué de sols fins très lâches : silts en surface (0 à 4 m), sables fins ensuite (surtout de 6 à 9 m) puis des argiles (de 10 à 12 m) avec quelques inclusions de graviers et de sables. Le substratum et à 15 m. La nappe fluctue en fonction des saisons entre 0,30 et 1,50 m.

3.1.2 Site de Cran La plaine alluviale de la Vilaine en aval de Redon est une vallée sédimentaire de près de 2 km de large. On y rencontre un dépôt d’argile sur une épaisseur de 10 à 20 m, reposant sur une couche de sable et de galets qui recouvre le substratum rocheux. À Cran, la rive droite est constituée par un dépôt d’argile molle marine de 17 m d’épaisseur reposant sur un substratum rocheux (schiste et phtanites).

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3.2 Essais du projet Solcyp

Des tests de validation ont été effectués sur trois sites : Gosier, Cran et Merville. Les deux derniers sites ont fait l’objet de nombreuses études dans le cadre de recherches programmées par les Laboratoires des Ponts et Chaussées. Les caractéristiques qui ont présidé au choix, ont été une relative homogénéité d’ensemble sur une profondeur minimale de 5 à 10 m.

3.2.1 Site de Gosier Les premiers essais avec le nouveau matériel ont été entrepris sur le site de Gosier en Guadeloupe situé dans une zone potentiellement liquéfiable, instrumentée et étudiée dans le cadre du projet ANR Belle Plaine. Des essais pressiométriques Ménard (par paliers) ont été réalisés pour compléter les profils obtenus au pénétromètre statique à pointe électrique et définir les pressions p’0 à utiliser puis deux sondages pressiométriques cycliques ont été réalisés.

La figure 4a présente les courbes de la pression imposée corrigée en fonction de la variation volumique obtenue pour 4 essais de la série MC2. Après une première partie qui correspond à la montée à la charge moyenne en monotone, la phase cyclique entre pM et pm montre la tendance à la stabilisation de quasiment tous les essais, même si celle-ci n’a jamais été atteinte. Apparemment, l’essai à la profondeur 7 m montre une accumulation importante de déformation volumique (couche d’argile molle).

0

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cycle 1cycle 16cycle 36cycle 52cycle 1cycle 16cycle 36cycle 52

Figure 4. a et b Essais d’expansion cycliques pré-forés site de Gosier

3.2.2 Site de Cran Les essais réalisés en 2011 ont été placés à proximité des séries réalisées en 1979.

À la profondeur 2 m, la courbe d’essai présentée sur la Figure 5a montre une accumulation de volume élevée (de l’ordre de 900 cm3) conduisant à la conclusion que le test a été effectué dans la couche molle et que la pression initiale déduite des études précédentes a été surestimée. Le signal obtenu représenté sur la Figure 5a est assez bruité car l’amplitude de la plage de pression est faible, et une interaction entre l’asservissement de l’eau et l’air n’a pu être corrigée à temps dans le pilotage.

3.2.3 Site de Merville Sur le site expérimental de Merville (Nord), on rencontre à 1,5 m environ de la surface, une couverture de limons peu plastiques, affectée par le battement de la nappe et de 1,5 à 42 m de profondeur, l’argile (surconsolidée) des Flandres de l’Yprésien.

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1m cycle 10cycle 1

2m cycle 1

3m cycle 10cycle 20

4m

Figure 5. a et b Essais d’expansion cycliques autoforés site de Cran

Un sondage au pressiomètre autoforeur avec des essais cycliques à 6, 8 et 12 m, en alternance avec des essais d’expansion monotone croissante à 5, 7 et 11 m, a été réalisé (Figure 6 et Figure 7a). Les trois premiers essais cycliques ont été réalisés avec la même amplitude fixée à partir des essais d’expansion monotones. Le dernier essai a été réalisé avec une amplitude basée sur l’essai à 11 m.

0

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0 1 2 3 4 5 6 7 8 9

vom

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e (m

l)

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6m

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Figure 6. Essais d’expansion cycliques autoforés multi-amplitude site de Merville

Les amplitudes de 60 kPa ont abouti à des résultats peu précis car d'une part la source de pression était réglée à une pression trop forte et de ce fait l'électrovanne de la chambre tampon avait du mal à réguler et d'autre part cette valeur est faible et de l'ordre de grandeur de la précision de l'asservissement. Il est donc nécessaire d’adapter les amplitudes aux profondeurs et de réévaluer la méthode de détermination de celle-ci.

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Profondeur 9 m

Profondeur 7 m

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6m

8m

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Figure 7. a et b Essais d’expansion cycliques (a) autoforés et (b) pré-forés site de Merville

À proximité immédiate un sondage au pressiomètre Ménard avec un perforage d’un mètre à la tarière hélicoïdale avec des essais cycliques à 5, 7 et 11 m permet de disposer de résultats comparables (figure 7).

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4 DISCUSSIONS

Un premier constat est que le signal obtenu représenté sur les Figure 4 à 7 est plus bruité que celui donné dans les articles de Jézéquel et Le Méhauté (1982) car l’asservissement par contrôleur pression volume à vis à bille en vitesse de volume (2%/mn) est plus stable qu’un asservissement en pression par électrovanne. Les cycles obtenus présentent des points singuliers à méplats et non de rebroussement.

Les différentes expérimentations ont montré qu’il faut réaliser un minimum de 50 cycles pour obtenir une évolution claire du module sécant et qu’il n’est pas possible de se limiter à quelques cycles – 3 à 10 par exemple – pour obtenir des résultats représentatifs. A noter que l’évolution du module sécant de 50 à 500 cycles est de 10% en moyenne et de 500 à 5000 de 3%.

Même si les différences de protocole dues à la mise au point de ces matériels et essais ne permettent pas une comparaison détaillée, les accumulations volumiques constatées sont du même ordre de grandeur avec les différents modes de mise en œuvre. La qualité du pré-forage est essentielle pour permettre un essai avec un volume injecté initialement minimal. Sur deux sites (Gosier et Cran) indépendamment du mode de mise en œuvre, l’accumulation importante de déformation à certains niveaux a permis de localiser les couches susceptibles d’une chute de caractéristiques importante lors de l’application de sollicitations cycliques (figures 4 et 5).

Tableau 1. Caractéristiques des essais Site Forage Sol z

(m)Gp,1

(105 Pa)aM0(%)

Gp,50 /Gp,1

Plancoët 8-3 silt 2 5,19 0,5 1,6 16-1 3 3,46 1 1,8 7-3 2 1,08 5 3 8-3 sable 7 9,98 0,5 2,2 16-1 7 6,49 1 2 7-3 7 2,69 5 3,3 8-3 argile 11 8,58 0,5 1,43 16-1 11 6,47 1 1,9 7-3 11 3,75 5 2,1 Cran 1 C1 argile moy 29,9 1,67 1,15 C2 argile moy 20 0,99 1,09 Gosier C2-PMT sable 7 51,6 0,5 1,60 9 13,9 3,5 1,41 Cran 2 A0-PAF argile 1 60,8 10 2,71 2 25,6 12 1,59 Merville PAF argile 6 426 0,8 2,01 12 294 3,5 1,93 Merville PMT argile 5 145 0,6 1,37 11 255 0,9 1,21

Lors de tous les essais, une stabilisation des déformations

moyennes des cycles en fonction du nombre de cycle a pu être observée. Selon le type de sol, les cycles tendent à se redresser plus ou moins fortement, comme cela semble être le cas pour les sables de Plancoët et l’argile de Merville.

Les séries d'essais multi-amplitude permettent d'obtenir des courbes d'évolution du module cyclique en fonction de la profondeur et de l'amplitude des cycles pour différentes natures de sol (figure 6 et tableau 2).

Tableau 2. Évolution du module sur plusieurs amplitudes

Site Outil z (m) Gp,50 / Gp,1 de la phase 1 2 3 4 Plancoët PAF 1 1.83 1.08 0.8 1.27 Merville PAF 12 1,93 0,99 0,87 0,98 PMT 11 1,21 0,94 1,42 1,06

On observe une évolution très similaire sur le site de sols lâches de Plancoët et sur le site d’argile raide de Merville : une évolution importante pour la première amplitude, puis une quasi stabilisation pour les autres amplitudes

5 CONCLUSION

Les différentes campagnes de sondages avec essais cycliques mono ou multi amplitudes ont montré l’intérêt de cet essai pour cerner l’évolution du module de cisaillement en fonction du nombre de cycle et le potentiel de l’essai à localiser les horizons susceptibles de liquéfaction.

Il reste à mieux préciser les conditions d’essais pour avoir des jeux de données comparables et si possible disposer de la mesure de la pression interstitielle au niveau de la membrane pour estimer les accumulations potentielles de pression ou adapter la vitesse d’essais afin de rester drainé.

6 REMERCIEMENT

Les auteurs désirent remercier le projet national SOLCYP ainsi que le ministère de l’Environnement de l’Énergie, du Développement Durable et de la Mer pour le financement de cette action de recherche ainsi que leurs collègues O. Malassingne et A. le Kouby.

7 RÉFÉRENCES

AFNOR (1999) Essai pressiométrique Ménard – partie 2 Essai avec cycle, NF P94-110-2, Reconnaissance et essais, pp. 43.

AFNOR (2000) Essai pressiométrique Ménard – partie 1 Essai sans cycle, NF P94-110-1, Reconnaissance et essais, pp. 43.

Borel S., Reiffsteck Ph., (2006) Caractérisation de la déformabilité des sols au moyen d’essais en place. LCPC Paris, pp. 132.

Briaud, J.L., (1992). The Pressuremeter, A. A. Balkema, Rotterdam, Netherlands.

Combarieu O., Canépa Y. (2001) L’essai cyclique au pressiomètre, BLPC, 233, 37-65.

Dupla, J.C., Canou J. (2003). Cyclic pressuremeter loading and liquefaction properties of sands, Soils and Foundations, Vol. 43(2), 17-31.

Jézéquel J.F., Le Méhauté A. (1982) Essais cycliques au pressiomètre autoforeur, Symposium sur la pressiométrie et ses applications en mer, Paris, Éditions Technip, 221-233.

Le Méhauté A. Jézéquel J.F., (1980) Essais cycliques au pressiomètre autoforeur, Rapports des LPC, FAER 1-05-09-22, 29 pages

Les Pressiomètres Louis Ménard (1960) Phase de déchargement des essais pressiométriques, Etude théorique et applications, Circulaire 3 pages

Puech A., Brucy F., Ma E., (1982) Calcul de la capacité axiale des pieux de fondations marines à partir du pressiomètre autoforeur, Symposium sur la pressiométrie et ses applications en mer, Paris, Éditions Technip, 373-388.

2387

Displacement response to axial cyclic loading of driven piles in sand

Réponse en déplacement au chargement cyclique axial de pieux battus dans le sable

Rimoy S., Jardine R., Standing J. Imperial College London

ABSTRACT: Interactive axial cyclic loading stability charts have been developed to guide the assessment of axial cyclic capacitydegradation of piles driven in sands. Less guidance is available regarding displacement accumulation and cyclic stiffness response at full scale. This paper focuses on axial cycling experiments of six full–scale steel open–ended pipe–piles at a marine sand site in Dunkerque, France. Multiple suites of cyclic loading were applied, interspersed with reference static tension capacity tests. The piles’stable, meta-stable and unstable capacity responses are identified with reference to a site-specific normalised cyclic interaction stability diagram. The stiffness response and rates of accumulated displacement associated with each style of cycling are reported. It is shown that under stable loading, the piles’ cyclic stiffnesses remain constant or decline marginally. Similar trends are observed withmeta-stable tests up to onset of an eventual cyclic failure, after which stiffness degrades rapidly. Unstable tests displayed shorterperiods of modest change before marked losses of cyclic stiffness. The patterns of accumulated displacement growth show morecomplex relationships with the cyclic loading parameters that can be expressed in multi-surface 3-D plots.

RÉSUMÉ : Des diagrammes interactifs de stabilité cyclique ont été développés afin d’évaluer la dégradation cyclique des pieux battusdans les sables. Peu de données sont disponibles à échelle réelle en ce qui concerne les déplacements. Cet article s’intéresse aux essaiscycliques axiaux de six pieux tubulaires en acier à base ouverte dans un site de sable marin à Dunkerque. Plusieurs séries dechargement cyclique ont été appliquées, entrecoupées d’essais statiques référentiels en traction. Les réponses stable, méta-stable et instable de capacité des pieux sont identifiées en relation avec un diagramme normalisé de stabilité cyclique. La réponse en termes derigidité et de taux de déplacement accumulé associée à chaque type de chargement cyclique est ensuite présentée. On montre que sous un chargement stable, la rigidité cyclique reste constante ou diminue légèrement. On observe des tendances similaires dans les essaisméta-stables jusqu'à l'apparition d'une éventuelle rupture cyclique, après laquelle la rigidité se dégrade rapidement. Les essais instables ont montré de courtes périodes de léger changement avant de fortes pertes de rigidité cyclique. Les schémas de croissancedes déplacements cumulés montrent des relations avec les paramètres de charge cyclique plus complexes qui peuvent être exprimées dans des représentations 3-D.

KEYWORDS: axial cyclic loading/ pile stiffness/ accumulated displacements/ offshore engineering/ renewable energy MOTS-CLÉS: chargement cyclique axial/rigidité du pieu/déplacements cumulés/ingénierie offshore/énergies renouvelables

1 INTRODUCTION

The axial cyclic response of driven pile foundations can be critical in the design of offshore oil and gas platforms, and multi-piled wind turbines, towers and pylons. Lateral and moment loads imposed by wind or wave action can be large compared to self-weights, leading to multiple modes of axial and lateral cyclic loading on the foundation piles. Lateral loading model tests have been reported that tracked the gradual rotation and stiffness of monopiles (Leblanc et al. 2010); however less guidance is available on full-scale displacement accumulation and stiffness responses under axial cycling.

Jardine et al. (2012) reviewed the potential effects of cyclic loading on offshore pile foundations and considered how these may be addressed in practical design for a range of geomaterials. They note that loads vary with platform weight, water depth, metocean environment and structural form. Of the 15 field research studies they identified, only one concerned silica sands, that at Dunkerque, France reported by Jardine & Standing (2000, 2012). Merritt et al. (2012) describe how the most severe tens or hundreds of cycles imposed in storms are the most critical to pile performance. The Jardine & Standing (2000) field study investigated behaviour up to 1000 cycles.

Karlsrud et al. (1986), Poulos (1988) and Jardine & Standing (2000) have used cyclic stability diagrams to guide the assessment of pile axial cyclic behaviour. These consider the interaction effects of cyclic and mean loads (normalised by static capacity before cycling) and the number of cycles applied. Such interaction diagrams may be zoned to identify a cyclically stable (S) region where there is no reduction of load capacity after N cycles, a meta-stable (MS) region where some reduction of load capacity occurs after N cycles and an unstable (US) region where cyclic failure develops within a small specified number of cycles. Jardine & Standing (2012) used a similar

scheme in interpretation of their field tests at Dunkerque (Figure 1) where multiple cyclic loading tests performed that were interspersed with reference static tension capacity (QT) tests. This paper focuses on further interpretation of the same axial cycling experiments. The axial static and cyclic stiffness responses are discussed and the accumulated cyclic displacement trends associated with each mode of cycling are examined, referring to the site specific normalised cyclic interaction stability diagram.

-0.2 0.0 0.2 0.4 0.6 0.8 1.00.0

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Figure 1. Axial cyclic interaction diagram for the full–scale pile tests in Dunkerque silica marine sands (Jardine & Standing 2012).

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2 SCOPE OF STUDY

Seven full-scale 457mm diameter, 13.5mm wall thickness (increased to 20mm over top 2.5m), open-ended steel pipe-piles, six (R1 – R6) with embedded lengths around 19m and one (C1) driven to 10m were installed as part of the Grouted Offshore Piles for Alternating Loading (GOPAL) project (Parker et al. 1999) in a flat area close to Dunkerque Port Ouest Industrial Zone. The site has a relatively deep profile of dense sand, Figure 2. Chow (1997) reported static and cyclic pile tests incorporating pore pressure measurements that showed a fully drained response over the loading rates applied. The piles’ cyclic capacity trends have been reported by Jardine and Standing (2000, 2012), while Jardine et al. (2006) reported the static tension capacity–time trends. 3 TEST PROGRAMME

Jardine et al. (2006) detail the testing arrangements, pile head load control and displacement measurements. The cyclic test programme is detailed on Table 1; load cycles were performed with periods between 1 to 2 minutes depending on the pile response. The axial cyclic load was applied in approximately sine wave forms as defined in Figure 3. The load-controlled tests involving only tensile pile head loads are termed ‘one–way’ while cycles ranging from tension to compression are referred to as ‘two–way’; tension loads and upward displacement responses are taken as positive throughout. Reference static tension tests to failure were conducted after most of the cyclic tests to assess the effects on the applied axial cyclic loading on the operational static tension (shaft) capacity and isolate any effects of previous (static or cyclic) loading phases from the current axial cyclic behaviour.

Figure 2. Typical site profile for Imperial College test site (Chow 1997)

4 RESULTS AND INTERPRETATION

4.1 Cyclic failure criteria

The axial cycling displacement response is classified as stable (S), meta–stable (MS) or unstable (US) according to the following criteria. Stable response signifies low and stabilising cyclic displacements that remain below 0.01 the pile diameter, D, and show slow rates of change ≤ 1mm/1000 cycles (N) up to N ≥ 1000 without causing loss in operational static shaft capacity. Tests with meta–stable responses accumulate > 0.01D displacements but < 0.1D with moderate rates (1mm/1000 cycles < rates ≤ 1mm/10 cycles) potentially leading to some degradation of the operational static shaft capacity but not causing failure within 100 cycles. Unstable responses lead to cyclic failure within 100 cycles, involving either accumulated permanent cyclic displacements > 0.1D or rates of accumulation of permanent cyclic displacements that exceed 1mm/10cycles with potentially very significant shaft degradation.

The fourteen cyclic tests gave a range of outcomes with one stable (set 1), four meta–stable (set 2) and nine unstable (set 3) responses indicated in Figure 1. The following sections analyse the cyclic stiffnesses and accumulated displacements seen in these three modes using the terms defined in Figure 4.

Figure 3. Load–controlled axial cycling illustrated (Tsuha et al. 2012)

4.2 Pile axial cyclic stiffness

The variations of the piles’ secant stiffness, k = ΔQ/Δs, under first-time static tension loading are shown on Figure 5 represented by the stiffnesses ratio, k/kRef, against the load ratio, Q/QRef, where kRef is the pile stiffness at the first monotonic load step, QRef, in the first-time tension tests. The 19m long piles (R2 to R6) follow common trends although one ‘younger’ and lower ultimate capacity 19m long pile R1 degraded more rapidly than the others as did the shorter (10m long) pile C1. Table 1. Axial cyclic loading test programme: after Jardine & Standing 2000) (

Test mode Test code Qcyclic(kN)

Qmean (kN)

QT(kN) Nf

US 3.R2.CY2 1000 1000 2500 9 MS 2.R3.CY2 700 700 2315 200+ US 2.R3.CY3 950 950 2050 13 MS 2.R4.CY2 1000 1000 2960 221+ US 2.R4.CY4 750 1250 2000 3 S 3.R4.CY6 400 405 2110 1000+ MS 2.R5.CY2 750 1250 2465 345 US 2.R5.CY3 700 700 2000 27 US 2.R6.CY2 750 1250 2000 1 US 2.R6.CY4 700 700 1585 24

One–way

MS 3.R6.CY6 700 700 1650 206 US 2.C1.CY3 620 -40 840 41 US 2.C1.CY4 445 165 620 1 Two–way US 2.C1.CY5 410 10 620 12

Test code explanation: XX M.YY.ZZN: XX = Pile response mode (S - Stable, MS – Meta-stable, US – Unstable) M = Testing campaign phase (out of 3) YY = Pile name (C1, R2 – R6) ZZ = Test type (T – Static tension, C - Static compression, CY – Axial cyclic) N = Test number on the pile in sequence from installation

Figure 6 examines the axial cyclic stiffness trends for the

stable and meta–stable (sets 1 & 2 on Figure 1). The initial normalised stiffness values (i.e. kl/kRef at N = 1) generally decreases as the proportion of applied Qmax to QT increases. In the stable loading test 3.R4.CY6, it can be seen that continued cycling leads to only a marginal stiffness decrease (12%) over 1000 cycles, with stiffness values stabilising and then marginally increasing after 200 cycles. Compared with this, the four meta–stable loading tests showed similarly mild stiffness degradation before manifesting sharply accelerating stiffness degradation as the piles approached cyclic failure under the conditions given in Table 1.

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Pile

hea

d lo

ad, Q

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)

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Figure 4. Illustration of the stiffness and displacement parameters used in the analyses

0 5 10 15 200.0

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Figure 5. Pile stiffness from the first–time axial static monotonic tension loadings normalised by the reference stiffnesses against normalised load

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Data not logged forthe first 34 cycles

Qcyc = 0.42QT

Figure 6. Axial cyclic loading stiffness kl responses normalised by kRef against number of cycles for the stable and meta–stable tests.

The loading stiffness kl degradation trends for the unstable tests (Set 3 of Figure 1) are shown on Figure 7. By definition, these tests failed with sudden stiffness loss after relatively few cycles. However, even these piles retained most of their initial stiffnesses until within ~10 cycles of final failure. Seemingly anomalous stiffness behaviour is observed towards failure in some one–way meta-stable and unstable loading tests when stiffnesses are defined from the unloading cycle phase ku, Figure 8. This reversal in normalised stiffness results from an increased opening-up of the load-unload hysteresis loops as

cyclic failure approaches with more plastic displacements accumulating on the loading loops leading to the progressively decreasing secant loading stiffnesses and apparently stiffer behaviour on unloading as cyclic loading approaches failure.

1 10 1000.0

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US 3.R2.CY2 (0.40) US 2.R6.CY2 (0.36) US 2.R3.CY3 (0.46) US 2.R6.CY4 (0.36) US 2.R4.CY4 (0.44) US 2.C1.CY3 (0.74) US 2.R5.CY3 (0.35) US 2.C1.CY4 (0.72) US 2.C1.CY5 (0.66)

All Unstable tests

Figure 7. Axial cyclic loading stiffness kl responses normalised by kRef against number of cycles for the unstable tests.

1 10 100 10000.0

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k u/k

Ref

Cycles, N

MS 2.R5.CY2 Qcyc = 0.30QT

MS 3.R6.CY6 Qcyc = 0.42QT

US 2.R6.CY4 Qcyc = 0.44QT



Figure 8. Axial cyclic unloading stiffness ku responses normalised by kRef against number of cycles for selected metastable and unstable tests.

4.3 Accumulated cyclic displacements

The patterns of pile head displacement accumulation for the stable and meta–stable cyclic tests are shown on Figure 9. Also shown are the reference lines related to the predefined thresholds for stable, metastable and unstable accumulated displacements rates. An almost static accumulated displacement trend was observed in the single fully stable loading test 3.R4.CY6. The meta–stable tests 2.R3.CY2 and 2.R4.CY2 developed higher, but steady displacement rates > 1mm/100cycles while the other two meta–stable tests 2.R5.CY2 and 3.R6.CY6 displaced by > 1mm/10cycles. A range of responses is evident for the unstable loading tests summarised in Figure 10 which develop displacement rates > 1mm/10cycles.

While the cyclic stiffness patterns varied principally as a function of the applied cyclic amplitudes Qcyclic, the accumulated cyclic displacement patterns depended on both the normalised mean (Qmean/QT) and cyclic (Qcyclic/QT) loads. Rimoy et al. (2013) demonstrate the interactive effects of the loading components Qcyclic and Qmean by considering the accumulated displacements developed after 3, 10, 30, 100, 200, and 300 cycles to produce tentative 3D surfaces equivalent to displacements of 2%, 0.2% or 0.02% pile diameter, Figure 11. The accumulated displacement trends flatten progressively as N increases. The zero cyclic effect boundary was set at Qcyclic/QT =

2390


0.1 following centrifuge studies by Julio (2009). No displacements are expected to accrue due to cycling below this level; further full-scale specific investigation of this lower threshold is required.

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0cyc

lesData not loggedfor first 34 cycles

Stable test

Figure 11. 3D plot for accumulated cyclic displacements equivalent to 0.02%D, 0.2%D and 2%D.

6 ACKNOWLEDGEMENTS

Figure 9. Accumulated cyclic displacements for the stable and meta–stable loading tests

1 10-10

0

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30

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a (m

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US 3.R2.CY2 US 2.R3.CY3 US 2.C1.CY3 US 2.R4.CY4 US 2.R5.CY3 US 2.C1.CY4 US 2.R6.CY2 US 2.R6.CY4 US 2.C1.CY5

The above research was funded by the EU (through the GOPAL project) and Health and Safety Executive (HSE) of UK. We gratefully acknowledge the Port Autonome de Dunkerque for providing the test site. The field testing was conducted in conjunction with Precision Monitoring Control Ltd. of Teesside UK. The first author has been supported by the Commonwealth Scholarship Commission during the writing of this paper. 7 REFERENCES

Chow F.C. 1997. Investigations into displacement pile behaviour for offshore foundations, PhD thesis, University of London (Imperial College).

Jardine R.J. & Standing J.R. 2000. Pile load testing performed for HSE cyclic loading study at Dunkirk, France. Two volumes. Offshore Technology Report OTO2000 007; Health and Safety Executive, London. 60p and 200p.

Jardine, R.J., Standing, J.R., & Chow, F.C. 2006. Some observations of the effects of time on the capacity of piles driven in sand.Géotechnique 56(4): 227-244.

Figure 10. Permanent accumulated cyclic displacements response for the unstable tests

Jardine R.J. & Standing J.R. 2012. Field axial cyclic loading experiments on piles driven sand. Soils and foundations 52(4): 723 - 736.

Jardine R.J., Puech A. & Andersen K. H. 2012. Cyclic loading of offshore piles: Potential effects and practical design. Proceedings of the SUT 7th International Conference on Offshore Site Investigation and Geotechnics, London, UK, pp. 59 - 97.

5 SUMMARY AND CONCLUSIONS

The analysis presented of the axial cyclic loading load–displacement, stiffness and accumulated displacements responses seen in tests on steel open-ended pipe piles driven in silica sand indicate the following.

Julio R.M.H. 2009. Comportement des pieux et des groupes de pieuxsous chargement latéral cyclique. These de doctorat, Ecole Nationale des Ponts et Chaussees, Paris, France.

(1) Axial load–displacement behaviour is highly non-linear, even at relatively low levels of loading. (2) The piles’ cyclic stiffnesses generally remained within 20% of those observed under initial static loading until cyclic failure was approached.

Karlsrud K., Nadim F. & Haugen T. 1986. Piles in clay under cyclic axial loading field test and computational modelling. Proceedings of the 3rd International Conference on Numerical Methods in Offshore Piling, Nantes, France, pp. 165 – 190.

Leblanc C., Houlsby G.T., & Bryne B.W. 2010. Response of stiff piles in sand to long-term cyclic lateral loading. Géotechnique 60(2): 79-90.

(3) The patterns of accumulated displacements depended on both the mean and cyclic normalised loading levels. (4) While displacements accumulate rapidly over just a few cycles in the unstable zone, extended cycling in the stable zone led to minimal (and stabilising) accumulated displacements and axial capacity gains (Jardine et al (2006) meta–stable tests showed intermediate behaviour.

Merritt A.S., Schroeder F.C., Jardine R.J., Stuyts B., Cathie D. & Cleverly D. 2012. Development of pile design methodology for an offshore wind farm in the North Sea. Proceedings of the SUT 7th International Conference on Offshore Site Investigation and Geotechnics, London, UK, pp. 439 - 447.

Parker E. J., Jardine R.J., Standing J.R. & Xavier J. 1999. Jet grouting to improve offshore pile capacity. Offshore Technology Conference, Houston, OTC 10828 1: 415 – 420.

Poulos H.G. 1988. Cyclic stability diagram for axially loaded piles. Journal of Geotechnical and Geoenvironmental Engineering, 114(8): 877-895.

Rimoy S., Jardine R., and Standing J. 2013. Displacement response to axial cycling of piles driven in sand. Geotechnical Engineering. 165 (GE1): 1 – 16.

2391

Experimental Testing of Monopiles in Sand Subjected to One-Way Long-Term Cyclic Lateral Loading

Étude expérimentale de monopiles dans le sable soumis à un chargement cyclique transversal non alterné

Roesen H.R., Ibsen L.B., Andersen L.V. Aalborg University, Aalborg, Denmark

ABSTRACT: In the offshore wind turbine industry the most widely used foundation type is the monopile. Due to the wave and windforces the monopile is subjected to a strong cyclic loading with varying amplitude, maximum loading level, and varying loading period. In this paper the soil–pile interaction of a monopile in sand subjected to a long-term cyclic lateral loading is investigated by means of small scale tests. The tests are conducted with a mechanical loading rig capable of applying the cyclic loading as a sine signal with varying amplitude, mean loading level, and loading period for more than 60 000 cycles. The tests are conducted in densesaturated sand. The maximum moment applied in the cyclic tests is varied from 18% to 36% of the ultimate lateral resistance found in a static loading test. The tests reveal that the accumulated rotation can be expressed by use of a power function. Further, static testsconducted post cyclic loading indicate that the static ultimate capacity increases with the magnitude of cyclic loading.

RÉSUMÉ: Dans l'industrie éolienne maritime, les fondations de type monopile sont les plus largement utilisées. En raison de la forcedes vagues et du vent, ces fondations sont soumises à des charges cycliques élevées dont l’amplitude, le niveau maximal et la fréquence varient. Dans cet article, l'interaction sol-pieu d'une fondation de type monopile implantée dans du sable et soumise à unchargement transversal cyclique est étudiée au moyen d'essais à échelle réduite. Les tests sont effectués avec une grue de chargement mécanique capable d'appliquer un chargement cyclique sinusoïdal avec amplitude, niveau moyen et période de chargement variablependant plus de 60 000 cycles. Les tests sont effectués dans un sable dense saturé. Le moment maximal appliqué durant les essaiscycliques varie de 18% à 36% de la résistance transversale ultime obtenue lors d’essais de chargement statique. Les essais montrentque la rotation accumulée peut être exprimée par une fonction puissance. En outre, des essais statiques menés après le chargementcyclique indiquent que la capacité statique ultime augmente avec le niveau du chargement cyclique.

KEYWORDS: Experimental, wind turbine foundation, monopile, long-term cyclic loading, dense sand. 1 INTRODUCTION

In the offshore wind turbine industry, the most widely used foundation type is the monopile, i.e. a large diameter stiff pile. During the lifetime of a wind turbine, the monopile foundation is subjected to few load cycles with large amplitudes, caused by the strong storms, and also to millions of lateral load cycles with low or intermediate amplitudes due to the wave loading. This loading may cause failure in the fatigue or serviceability limit states, FLS and SLS respectively (Wichtmann et al. 2008). The cyclic loading might induce a change in the soil stiffness and a permanent accumulated rocking rotation (tilt) of the turbine. Due to the efficiency of the wind turbine, strict demands for the rotation and the stiffness of the entire structure are normally made and thus, the change in stiffness and rotation becomes key issues in the design. However, the current design guidance, DNV (2011), on this long-term loading is limited and a procedure for designing large diameter piles is yet to be fully expressed and confirmed. The development of a reliable design method requires verification and for that in-situ and large-scale testing is by far the best tool. However, this is also the most expensive and time-consuming tool. Therefore, the recent choice for evaluating the cyclic behaviour has been numerical modelling and small scale testing. Several authors has investigated this e.g. Niemunis et al. (2005), Achmus et al. (2005), Peng et al. (2006), LeBlanc et al. (2010) and Achmus et al. (2011). However, the research has mainly been based on cyclic triaxial tests, FEM-calculations and 1g experimental setups in dry sand.

In this paper, a 1g testing rig for modelling the environmental loading on a stiff monopile foundation in dense saturated sand is described and the results from four one-way cyclic loading tests are presented. The purpose of the cyclic tests is to evaluate the influence of the number of load cycles, N,on the accumulated rocking rotation of the pile at seabed, under

long-term cyclic loading with constant frequency but different loading amplitude and mean loading level.

The characteristic of the cyclic loading can be described by the ratios and as defined by LeBlanc et al (2010). expresses the magnitude of the loading as the ratio between the maximum load in a load cycle and the maximum static lateral

capacity, . will take a value between 0 and 1. The cyclic load ratio defines the direction of the loading on the basis of the minimum and maximum load in a load

cycle, . will take the value 1 for a static test, 0 for one-way loading, and -1 for two-way loading.

2 EXPERIMENTAL MODEL TESTS

The 1g small scale tests are carried out at the geotechnical laboratory at Aalborg University, Denmark. In the tests an open ended aluminium pipe pile is used. The pile is scaled approximately 1:50 in relation to a typical offshore monopile. In Table 1 the dimensions of the model pile are presented.

Table 1. Dimensions of the open ended aluminium pipe pile.

Diameter Embedded length Wall thickness Load eccentricity

(mm) (mm) (mm) (mm)

100 500 5 600The bending stiffness of the model pile is similar to a scaled

prototype steel pile, however, the behaviour of the pile during loading also depends on the stiffness of the surrounding soil. According to Poulus and Hull (1989) a pile behaves flexible if

and rigidly if , where is a critical length


1

Experimental Testing of Monopiles in Sand Subjected to One-Way Long-Term Cyclic Lateral Loading

Test expérimental de monopiles implantées dans du sable et soumises à un chargement latéral cyclique à long terme

H. R. Roesen, L. B. Ibsen and L. V. Andersen Aalborg University, Aalborg, Denmark

ABSTRACT: In the offshore wind turbine industry the most widely used foundation type is the monopile. Due to the wave and wind forces the monopile is subjected to a strong cyclic loading with varying amplitude, maximum loading level, and varying loading period. In this paper the soil–pile interaction of a monopile in sand subjected to a long-term cyclic lateral loading is investigated by means of small scale tests. The tests are conducted with a mechanical loading rig capable of applying the cyclic loading as a sine signal with varying amplitude, mean loading level, and loading period for more than 60 000 cycles. The tests are conducted in dense saturated sand. The maximum moment applied in the cyclic tests is varied from 18% to 36% of the ultimate lateral resistance found in a static loading test. The tests reveal that the accumulated rotation can be expressed by use of a power function. Further, static tests conducted post cyclic loading indicate that the static ultimate capacity increases with the magnitude of cyclic loading.

RÉSUMÉ: Dans l'industrie éolienne offshore, le type de fondation le plus largement utilisé est la monopile. En raison de la force desvagues et du vent, la monopile est soumise à une charge cyclique élevée dont l’amplitude, le niveau de charge maximale et la périodicité varient. Dans cet article, l'interaction sol-pieu d'une monopile implantées dans du sable et soumises à un chargement latéral cyclique est étudiée au moyen d'essais à échelle réduite. Les tests sont effectués avec une grue de chargement mécanique capable d'appliquer un chargement cyclique sinusoïdal avec amplitude, niveau moyen et période de chargement variable pendant plus de 60 000 cycles. Les tests sont effectués dans du sable dense saturé. Le moment maximum appliqué durant les essais cycliques varie de 18% à 36% de la résistance latérale ultime obtenue lors d’essais de chargement statique. Les essais montrent que la rotation accumulée peut être exprimée par l'utilisation d'une fonction de puissance. En outre, des essais statiques menés après le chargement cyclique indiquent que la capacité statique ultime augmente avec le niveau du chargement cyclique.

KEYWORDS: Experimental, wind turbine foundation, monopile, long-term cyclic loading, dense sand.

1 INTRODUCTION

In the offshore wind turbine industry, the most widely used foundation type is the monopile, i.e. a large diameter stiff pile. During the lifetime of a wind turbine, the monopile foundation is subjected to few load cycles with large amplitudes, caused by the strong storms, and also to millions of lateral load cycles with low or intermediate amplitudes due to the wave loading. This loading may cause failure in the fatigue or serviceability limit states, FLS and SLS respectively (Wichtmann et al. 2008). The cyclic loading might induce a change in the soil stiffness and a permanent accumulated rocking rotation (tilt) of the turbine. Due to the efficiency of the wind turbine, strict demands for the rotation and the stiffness of the entire structure are normally made and thus, the change in stiffness and rotation becomes key issues in the design. However, the current design guidance, DNV (2011), on this long-term loading is limited and a procedure for designing large diameter piles is yet to be fully expressed and confirmed. The development of a reliable design method requires verification and for that in-situ and large-scale testing is by far the best tool. However, this is also the most expensive and time-consuming tool. Therefore, the recent choice for evaluating the cyclic behaviour has been numerical modelling and small scale testing. Several authors has investigated this e.g. Niemunis et al. (2005), Achmus et al. (2005), Peng et al. (2006), LeBlanc et al. (2010) and Achmus et al. (2011). However, the research has mainly been based on cyclic triaxial tests, FEM-calculations and 1g experimental setups in dry sand.

In this paper, a 1g testing rig for modelling the environmental loading on a stiff monopile foundation in dense saturated sand is described and the results from four one-way

cyclic loading tests are presented. The purpose of the cyclic tests is to evaluate the influence of the number of load cycles, , on the accumulated rocking rotation of the pile at seabed, under long-term cyclic loading with constant frequency but different loading amplitude and mean loading level.

The characteristic of the cyclic loading can be described by the ratios and as defined by LeBlanc et al (2010). expresses the magnitude of the loading as the ratio between the maximum load in a load cycle and the maximum static lateral capacity, = /. will take a value between 0 and 1. The cyclic load ratio defines the direction of the loading on the basis of the minimum and maximum load in a load cycle, = /. will take the value 1 for a static test, 0 for one-way loading, and -1 for two-way loading.

2 EXPERIMENTAL MODEL TESTS

The 1g small scale tests are carried out at the geotechnical laboratory at Aalborg University, Denmark. In the tests an open ended aluminium pipe pile is used. The pile is scaled approximately 1:50 in relation to a typical offshore monopile. In Table 1 the dimensions of the model pile are presented.

Table 1. Dimensions of the open ended aluminium pipe pile.

Diameter Embedded length Wall thickness Load eccentricity

(mm) (mm) (mm) (mm)

100 500 5 600

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2

The bending stiffness of the model pile is similar to a scaled prototype steel pile, however, the behaviour of the pile during loading also depends on the stiffness of the surrounding soil. According to Poulus and Hull (1989) a pile behaves flexible if and rigidly if /3, where is a critical length defined by Eq. 1. is the bending stiffness of the pile and is Youngs modulus of elasticity of the soil.

= 4.44

. (1)

Due to the low stresses in the soil at 1g small scale testing, the stiffness is also low. From previous testing and numerical modelling an estimated soil stiffness of 4 MPa can be used for the sand in the test setup (Roesen et al. 2010). With use of Eq. 1 the model pile is thereby found to behave rigidly during lateral loading. In comparison a prototype steel monopile with = 5m and = 0.07 m installed in sand with Es = 70 MPa is found to behave rigidly with a slenderness ratio / = 3and behave flexible with / = 9. Thus, for the examined slenderness ratio (/ = 5) the model pile experiences a more rigid behaviour than the prototype pile. Nevertheless, the results obtained in the small scale model tests can be used as underlying basis for understanding the monopile behaviour during lateral cyclic long-term loading.

The test setup consists of a cylindrical sand container with an inner radius of 2.00 m and a height of 1.20 m surrounded by a loading frame equipped for both static and cyclic loading. The setup is an improvement of the system presented in Roesen et al. (2012) which originally is based on the setup presented by LeBlanc et al. (2010). A cross-sectional sketch and a photo of the system are shown in Figure 1 and 2. The pile is installed in the middle of the container by use of a mechanical motor with installation velocity of 0.02 mm/s. The container holds up to 0.90 m dense saturated sand with 0.30 m highly permeable gravel underneath. In the bottom a drainage system with perforated pipes ensures homogeneous in- and outflow of water.

The cyclic loading system is a simple load controlled system based on a lever arm, weight hangers with applied masses, ,, and , wires, and an electric motor controlling the rotation of weight . The rotation causes an oscillating motion on the lever and thereby a cyclic loading on the pile. The system is thereby capable of providing sinusoidal loading to the pile for more than 60 000 load cycles. The rotational frequency of the motor is set to 0.1 Hz to be in agreement with environmental wave loading (Peng et al. 2006).

Initially, when the mass = = 0, the mass is chosen to outbalance the system. Depending on the weights chosen for and the system is capable of providing both one- and two-way loading with varying and, i.e. different direction, amplitude, and mean loading level. The loading is applied through steel wires attached to the pile 600 mm above soil surface. Hence, the pile experiences both horizontal and moment loading. In both sides of the pile a HBM U2A 100 kg load cell is attached measuring the actual force applied to the pile throughout the whole test. The displacement of the pile is measured using three WS10-125-R1K-L10 displacement transducers from ASM GmbH. The transducers, 1, 2, and 3are mounted 600 mm, 375 mm, and 155 mm above soil surface, respectively. The rocking rotation,, and displacement of the pile at soil surface is found by use of linear regression of the three measurements assuming rigid pile behaviour. The data sampling rate is 2 Hz.

Before conducting any cyclic tests a static loading test is performed. The static test is conducted displacement controlled by use of a motor with a loading rate of 0.02 mm/s. The displacement is actuated 600 mm above soil surface, i.e. the same height as the loading in the cyclic loading tests. The pile is loaded to a rotation of 2°, unloaded, and reloaded to failure. The static test is used as a reference for the ultimate lateral

Figure 1. Sketch of the test setup. F1 and F2 refer to the two load cells, D1, D2, and D3 refer to the three displacement transducers and m1, m2, and m3, refer to the weights applied on the load hangers. All measurements are in meters.

Figure 2. Test setup for cyclically long-term loaded monopiles.

Table 2. Test programme with relative soil densities, , loading characteristics, and number of cycles, .

Test No.

Type (%)

Static test after cyclic loading

1 Static 78.56 - - - -

2 Cyclic 87.76 0.18 0.03 50 894 yes

3 Cyclic 85.38 0.24 0.10 51 732 no

4 Cyclic 87.87 0.25 -0.01 50 960 yes

5 Cyclic 91.70 0.36 0.03 60 224 yes

resistance and the maximum resistance obtained is interpreted as the ULS load on the pile.

In total four long-term cyclic loading tests are performed, each with more than 50 000 load cycles. The tests are conducted with = , i.e. one-way loading with the target = 0. The magnitudes of the loading in the cyclic tests are chosen to reflect realistic loading conditions for FLS and SLS loading, which according to LeBlanc et al. (2010) is approximately 30% and 40% of the ultimate limit state loading (ULS), respectively. Thus, the target maximum moment applied in the cyclic loading tests are defined as 20%, 25%, 30% and 40% of the maximum static lateral resistance, i.e. is chosen in the interval 0.2 to 0.4. In Table 2 a summary of the testing programme with the obtained loading characteristics is presented.

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3

In general the magnitude of the loading is seen to be a little less than expected. This result verifies the importance of measuring the actual force on the pile as some of the applied load is lost in the system due to friction. For three of the tests the cyclic load ratio,, is seen to be close to zero which is in agreement with the target loading.

In order to investigate the influence of cyclic loading on the ultimate lateral resistance static loading tests were performed after the cyclic loading.

2.1 Soil Conditions

The tests are conducted using saturated Aalborg University Sand No. 1 (Baskarp Sand No.15). In Table 3 the properties of the sand are summarised.

Table 3. Properties of Aalborg University Sand No. 1

Specific grain density

Maximum void ratio

Minimum void ratio

50%- quantile

Uniformity coefficient

= /(-) (-) (-) (mm) (-)

2.64 0.858 0.549 0.14 1.78

Prior to each test the sand is prepared by use of an initially upward gradient of 0.9 followed by mechanical vibration with a rod vibrator. The obtained homogeneity and compaction of the sand is verified by conducting three cone penetration tests (CPT) with a laboratory cone; one in the middle of the container and two in a distance 400 mm from the centre in the active and passive side of the pile, respectively. The relative densities of the sand,, are derived in accordance to Ibsen et. al (2009) where the laboratory cone is correlated with in-house triaxial tests on the same sand type. The mean values of the relative densities found prior to each experiment are presented in Table 2 together with the characteristic of the tests themselves.

3 TEST RESULTS

Initially, the static loading test is used as a reference test for the ULS moment capacity and thus the choice of maximum moment loading in the cyclic tests. The moment-rotation relationships obtained in both the static and the cyclic tests are presented in Figure 3. The static test clearly defines a maximum moment capacity of 360 Nm which is interpreted as the ULS load. In all the cyclic tests the rotation obtained in the first loading cycle follows the static reference test cf. Figure 3. This verifies the use of the static test as a reference for the loading despite the difference in relative densities of the soil cf. Table 2. Even though the cyclic loading system is an improvement of the system presented in Roesen et al. (2012) the maximum moment loading in the cyclic tests are seen to decrease a little during the test. Therefore, the characteristics of the cyclic loading, and cf. Table 2, are calculated as mean values over the whole test and is seen to be lower than the target value.

In Figure 4 the rotation of the pile, , at soil surface as a function of the number of cycles, , for test no. 2 is presented. The figure shows the cyclic response during loading and the rotation is seen to accumulate throughout the entire test. Similar results are obtained in the three other tests. In the evaluation of the accumulated rotation the maximum values of the rotation are used, i.e. the rotation marked with dark grey in Figure 4. As seen in Figure 3 the rotation in the first loading cycle is equal to the rotation obtained in the static reference test. Thus, in order to evaluate the influence of the cyclic loading only the accumulated rotation, Δ = − , is investigated. is the rotation obtained at the th loading cycle and is the rotation obtained in the first loading cycle.

Figure 3. Moment-rotation relationships of the static reference test and the four cyclic loading tests.

Figure 4. Rotation of the pile at soil surface as a function of the number of cycles in the test with = 0.25. Maximum and minimum values of the rotation are indicated by dark grey and black colouring.

Figure 5. Normalised accumulated rotation as a function of the number of cycles for the four cyclic tests.

In Figure 5 the accumulated rotation obtained in all four cyclic tests are presented. The rotations are normalised with respect to the rotation obtained in the first loading cycle. The accumulated rotations of the stiff pile are fitted with a power function as suggested by several authors, e.g. Long and Vanneste (1994), Peralta and Achmus (2010), and LeBlanc et al. (2010). The fitted expression is given by Eq. 2 and shown as the dotted black lines in Figure 5.

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4

= (2)

and are dimensionless constants determined empirically from the tests. The results from the long-term one-way loading cf. Figure 5 shows a general good agreement with the power function even though deviations in the first 1000 cycles are observed. The values for the power are found to be similar for all the tests with values in the range of 0.11 to 0.18. These values are found to be smaller than the value of 0.31 as presented by LeBlanc et al. (2010). The results for the constant cf. Figure 6 indicates that depends linearly of the magnitude of the loading which is in agreement with the findings in LeBlanc et al. (2010).

Figure 6. Fitted empirical constant as a function of the loading magnitude in the four cyclic tests.

The influence of the cyclic loading on the static lateral capacity is evaluated by means of the results from the three static tests performed post cyclic loading cf. Figure 7. The maximum moments obtained indicates that the lateral capacity depends on the cyclic loading and increases with increasing load magnitude.

Figure 7. Moment-rotation relationships obtained in the static tests post cyclic loading compared with the reference static test.

4 CONCLUSION

This paper presents a description of a 1g laboratory small scale test setup for modelling laterally long-term cyclic loading of a stiff pile in saturated dense sand. A static loading test and four one-way cyclic loading tests with maximum moment loading equal to 18% to 36% of the maximum static capacity are presented. The purpose of the tests is to evaluate the influence of the number of load cycles on the accumulated rocking rotation of the pile at seabed during long-term cyclic loading. In addition the effect of the cyclic loading on the static lateral capacity is evaluated by means of static loading tests conducted post cyclic loading.

All the tests are carried out with an open ended aluminium pipe pile scaled approximately 1:50 in relation to a typical monopile foundation for an offshore wind turbine. In the four

cyclic tests more than 50 000 load cycles are applied to the pile. When evaluating the cyclic tests the accumulated rotation normalised with respect to the rotation obtained in the first loading cycles is used. The results reveal that the accumulation of rotation during long-term cyclic loading can be described by use of a power function. Further, the maximum moments obtained in the static tests conducted post cyclic loading indicates that the lateral capacity depends on the cyclic loading and increases with increasing load magnitude.

The entire test setup is still in the initial phase of testing and can be improved even more. Thus, the findings inhere must be evaluated further and supplemented with additional testing with varied loading characteristics, i.e. varied and for both one- and two-way loading.

5 ACKNOWLEDGEMENTS

This research is associated with the EUDP programme “Monopile cost reduction and demonstration by joint applied research” funded by the Danish energy sector. The financial support is sincerely acknowledged.

6 REFERENCES

Achmus, M., Abdel-Rahman, K. and Peralta, P. 2005. On the design of monopile foundations with respect to static and quasi-static loading. Copenhagen Offshore Wind 2005.

Achmus, M., Albiker, J. and Abdel-Rahman, K. 2011. Investigations on the behaviour of large diameter piles under cyclic lateral loading. In: Frontiers in Offshore Geotechnics II - Gourvenev & White (eds), Taylor & Francis Group, LLC.

DNV 2010. Offshore standard DNV-OS-J101: Design of offshore wind turbine structures, Technical report DNV-OS-J101, Det Norske Veritas.

Ibsen, L. B., Hanson, M. Hjort, T. and Taarup, M. 2009. MC-parameter Calibration of Baskarp Sand No. 15, DCE Technical Report No. 62. Department of Civil Engineering, Aalborg University

LeBlanc, C., Houlsby, G. and Byrne, B. 2010. Response of stiff piles to long-term cyclic lateral load, Géotechnique, 60 (2), pp. 79-90.

Long J. H. and Vanneste G. 1994. Effects of Cyclic Lateral Loads on Piles in Sand. Journal of Geotechnical Engineering, 120 (1), pp. 225-244.

Niemunis,A., Wichtmann, T. and Triantafyllidis, T. 2005. A high-cycle accumulation model for sand, Computer and Geotechnics, 32 (4), pp. 245-263.

Peng, J.-R., Clarke, B. G. and Rouainia, M. 2006. A device to Cyclic Lateral Loaded Model Piles, Geotechnical Testing Journal 29 (4) pp. 1-7.

Peralta, P. and Achmus, M. 2010. An experimental investigation of piles in sand subjected to lateral cyclic loads, 7th International Conference on Physical Modeling in Geotechnics, Zurich, Switzerland.

Poulus H., and Hull T. 1989. The Role of Analytical Geomechanics in Foundation Engineering. Foundation Engineering.: Current Principles and Practices, 2, pp. 1578-1606.

Roesen, H. R., Thomassen, K., Sørensen, S. P. H., and Ibsen, L. B., 2010. Evaluation of Small-Scale Laterally Loaded Non-Slender Monopiles in Sand DCE Technical Report No. 91, Aalborg University. Department of Civil Engineering.

Roesen, H. R., Ibsen, L. B., and Andersen, L. V. 2012. Small-Scale Testing Rig for Long-Term Cyclically Loaded Monopiles in Cohesionless Soil, Proceedings of the 16th Nordic Geotechnical Meeting, Copenhagen, 9-12 May, 2012, vol. 1/2, p.435-442..

Wichtmann, T., Niemunis, A. and Triantafyllidis, T. 2008. Prediction of long-term deformations for monopile foundations of offshore wind power plants. 11th Baltic Sea Geotechnical Conference: Geotechnics in Maritime Engineering, Gdansk, Poland.

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Pieu sous charge latérale : développement de lois de dégradation pour prendre en compte l’effet des cycles

Pile cyclic lateral loading: Development of degradation laws for describing the cyclic effect

Rosquoët F. Laboratoire LTI (AE3899), Amiens, France

Thorel L., Garnier J., Chenaf N. LUNAM Université, IFSTTAR, Nantes, France

RÉSUMÉ : A l’origine, le dimensionnement des pieux sous une charge latérale supposait que le sol est entièrement à l’état de rupture (calcul aux états limites). Les méthodes de calcul ont progressé et le dimensionnement est maintenant réalisé en déplacement maissans possibilité de tenir compte de l’effet des cycles de chargement (sauf dans le cas des ouvrages offshore). Pour corriger cette lacune, nous proposons deux méthodes : une méthode globale basée sur le déplacement en tête de pieu et sur le moment maximum, etune méthode locale basée sur les courbes P-y. Pour la méthode globale, nous montrons que l’effet des cycles sur le déplacement estessentiellement lié au rapport entre l’amplitude de la charge cyclique et la charge maximale. Nous proposons une loi de type logarithme donnant le déplacement relatif en fonction du nombre de cycles. On notera que l’effet des cycles sur le moment maximumest faible. La méthode locale est basée sur l’interaction entre le sol et le pieu permettant de relier directement la réaction latérale du sol P et le déplacement du sol y. Nous introduisons un coefficient d’abattement qui permet de prendre en compte l’effet des cycles enmodifiant la réaction des courbes P-y statiques.

ABSTRACT: Generally, the design of pile under lateral cyclic loads supposed that the soil is completely in the state of failure (limitstates calculation). However, the calculation methods progressed and the design can be executed in displacement and at maximumbending moment. To analyse the cyclic effect we propose two methods: the global method is based on pile head displacement andmaximum bending moments and local method is based on P-y curves. For the global method, we purpose an empirical law to evaluate pile head displacements at application point. A simple power function of DF/F and a logarithm function of the number of cycles are proposed to calculate pile head displacements under cyclic loading from the displacement values under applied monotonic loads. Weshow that the effect of the cycles on the bending moments is weak. Local methods are based on the soil reaction profile, P and the pile lateral displacement, y called P-y curves. We purpose a reduction coefficient then applied to the monotonic P-y curves to take thecyclic effect on the soil degradation into account.

MOTS-CLES : Comportement sous chargements cycliques – Séismes KEYWORDS: Pile under lateral cyclic load – Seism

1 INTRODUCTION

Le chargement latéral cyclique des pieux est généralement le résultat des sollicitations mécaniques engendrées par les vagues, le vent sur des structures offshore, l'amarrage de bateaux sur des quais, des surcharges variables ou des dilatations thermiques. Il est caractérisé par quatre paramètres qui sont la charge maximum appliquée F ; l’amplitude de la variation de la charge DF ; le nombre de cycles n et le type de chargement (non-alterné ou alterné).

L’utilisation des modèles réduits de pieux permet de réaliser des études paramétriques et ainsi améliorer notre compréhension de ces phénomènes à multiples variables. Une meilleure connaissance et quantification de la réponse des pieux sous charge latérale cyclique permettra d’optimiser leurs dimensionnement pour ce type de sollicitation mécanique.

La modélisation physique des structures géotechniques en centrifugeuse est une technique assez répandue (Garnier 2001) et a déjà été appliquée à l’étude de pieux sous charge latérale cyclique dans le sable (Rosquoët 2004, Rakotonindriana 2009) ou dans l’argile (Khemakhem 2012). La centrifugation des modèles réduits est indispensable pour assurer le respect des conditions de similitude qui imposent qu’un modèle à l’échelle 1/n soit testé sous une accélération centrifuge égale à n fois la gravité terrestre.

Un important programme de recherche est en cours en France sur les pieux sous charge cycliques (Projet national SOLCYP). Les travaux décrits ci-dessous s’intègrent dans ce programme et concernent les charges latérales cycliques. Le but final de ces études est de proposer une méthode rationnelle de

dimensionnement des pieux soumis à de telles sollicitations et d’évaluer l’influence du chargement latéral cyclique sur les éléments qui permettent de dimensionner le pieu, c’est-à-dire le déplacement horizontal en tête y, le moment maximum M ou encore les relations entre la réaction du sol P et le déplacement horizontal du pieu y, appelées « courbes P-y ».

L’étude se limite aux cas où les cycles dégradent la réaction du sol et aux charges cycliques de service. Le pieu modélisé peut être considéré comme un pieu souple. La partie du pieu située à une profondeur supérieure à 3 fois la longueur de transfert l0, n’a théoriquement plus d’influence sur la réponse due à une charge en tête (Frank 1999). De fait, dans cette étude, on admettra que l’effet des cycles se manifeste essentiellement dans les couches de surface, à des profondeurs inférieures à 4 ou 5 fois le diamètre du pieu. A titre de comparaison, l’API (2002) fixe cette profondeur limite à z/B = 2,625 pour des sables, soit 1,89 m dans notre cas.

2 DISPOSITIF ET MÉTHODE EXPÉRIMENTALE

Les essais ont été effectués dans des massifs secs homogènes de sable de Fontainebleau NE34 de poids volumique 16 kN/m3

(indice de densité ID = 86 %) reconstitués par pluviation. Le pieu modèle au 1/40ème est mis en place à 1g par battage dans les massifs préalablement reconstitués. Ce pieu modèle est testé sous une accélération de 40 g. Il représente un pieu grandeur réelle de 0,72 m de diamètre, de 12 m de longueur de fiche ayant une rigidité à la flexion de 476 MN.m² (Figure 1). Il est instrumenté par 20 paires de jauges de déformation collées sur

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deux génératrices diamétralement opposées permettant de déterminer à tout moment le profil des moments fléchissants.

Figure 1. Pieu modèle instrumenté de 18mm de diamètre.

Un servo-vérin horizontal placé sur un bâti rigide fixé sur le conteneur d’essai permet d’appliquer le chargement à 40 mm au-dessus du sol. Deux capteurs de déplacement (d’une course de 100 mm) fixés par l’intermédiaire d’une rotule à 20 mm et de 65 mm au dessus du sol permettent de connaitre le déplacement au point d’application de la charge (Figure 2). L'effort est mesuré à l'aide d'un capteur de force, d’une capacité maximale de 500 daN. Pour le chargement unidirectionnel non alterné (traction uniquement), un câble assure la liaison entre le capteur et le pieu modèle.

Figure 2. Détail du dispositif de chargement avec les capteurs de déplacement.

Dans le cas d’un chargement cyclique de type non alterné, l’effort est toujours appliqué dans le même sens, la charge variant entre F et F – DF.

3 MÉTHODE GLOBALE

3.1 Déplacement en tête

L’effet des cycles sur l’évolution du déplacement en tête au point d’application de la charge (figure 3) est fortement dépendant de l’amplitude des cycles (DF) et du chargement maximal (F).

1,00

1,05

1,10

1,15

1,20

1,25

1,30

0 5 10 15 20 25 30 35 40

Dép

lace

men

t rel

atif

(.)

Nombre de cycles (.)

P33 ; DF = 600 N P344 ; DF = 600N

P36 ; DF = 450 N

P347 ; DF = 450 N

P32 ; DF= 300 N

P318 ; DF = 150 N

P346 ; DF = 150 N

Valeur modèle, ID = 86 %

Figure 3. Evolution des déplacements relatifs sous la charge maximale (F = 960 kN) en fonction du nombre de cycles pour différentes amplitudes DF.

Comme déjà observé par d’autres auteurs, une loi logarithmique représente très correctement la relation entre le déplacement relatif et le nombre de cycles (Eq. 1) :

nbyn

yln1

1 (1)

où yn est le déplacement au cycle n, y1 le déplacement à la fin du chargement statique, b un coefficient positif adimensionnel et n le nombre de cycles (Figure 3).

L’interpolation des courbes est réalisée par la méthode des moindres carrés. Pour tous les cas étudiés, les valeurs expérimentales sont proches de la loi logarithmique choisie (coefficient de corrélation R2 = 0,98). Le coefficient « b » dépend de l’amplitude des cycles. Lorsque l’amplitude des cycles DF tend vers 0, nous sommes dans le cas d’un essai de fluage puisque la charge est constante et égale à F. Nous avons observé que le déplacement induit par le fluage peut être négligé, par conséquent, le coefficient « b », est dans ce cas proche de 0.

L’évolution du déplacement relatif en fonction du nombre de cycles, pour chaque essai réalisé, peut être caractérisée par ce coefficient « b ». On note b l’incertitude liée à la mesure de ce déplacement. En supposant que les incertitudes sur le déplacement relatif soient toutes de même amplitude nous pouvons calculer les incertitudes sur la constante « b » de la fonction logarithmique (Rosquoët 2004).

Tableau 1. Estimation de b et de l’incertitude b associée. Essai Nombre de

cyclesF(kN)

DF(kN)

b(.)

b

(.)P33 14 960 960 0,082 0,019 P344 14 960 960 0,081 0,017 P36 18 960 720 0,078 0,017 P347 40 960 720 0,075 0,01 P32 15 960 480 0,071 0,021 P318 25 960 240 0,044 0,017 P346 40 960 240 0,049 0,01

Il apparaît (Tableau 1), que le coefficient « b » est strictement croissant avec l’amplitude DF. Il est possible de représenter les variations de « b » par une fonction puissance. La figure 4 montre la courbe d’évolution du coefficient « b » en fonction du rapport entre l’amplitude et la charge maximale appliquée (DF/F), permettant ainsi de rendre adimensionnelle l’amplitude des cycles.

0

0,02

0,04

0,06

0,08

0,1

0 0,2 0,4 0,6 0,8 1

b (.)

DF/F (.)

R2 = 0,98

Valeur prototypeID = 86 %F = 960 kN

b 0,08DFF

0,35

Figure 4. Evolution du coefficient « b » en fonction de DF/F.

L’expression du coefficient « b » défini sur la figure 4 en fonction de DF/F permet de compléter l’équation (2) pour ainsi donner le déplacement en tête pour un cycle n donné.

35,0

1

ln08,01

F

DFnyyn (2)

Le tableau 2 présente les valeurs proposées dans la littérature pour le paramètre b, que l’on peut appeler paramètre de dégradation.

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Tableau 2. Plages des valeurs de b proposées dans la littérature.

Auteurs Sol Pieux testés Nombrede cycles

Paramètre de dégradation

Hettler(1981)

Sablesec

Rigides / 1g 0,2

Bouafia(1994)

Sablesec

RigidesCentrifugés

5 0,18 < b < 0,25

Lin et Liao (1999)

Diverssables

Pieuxin situ

100 0,02 < b < 0,24

Verdure et al. (2003)

Sablesec dense

FlexiblesCentrifugés

50 0,04< b < 0,18

Rakotonindriana (2009)

Sablesec dense

FlexiblesCentrifugés

500 0,12

Li et al. (2010)

Sablesec dense

RigidesCentrifugés

100 à 1000

0,17< b < 0,25

Peralta(2010)

Sablesec

Flexibles 1-g 10000 0,21

Au regard des expressions de « b » proposées dans la littérature et des résultats des études réalisées par Rosquoët (2004) et Rakotonindriana (2009), nous retenons une valeur moyenne de b de 0,1. Finalement nous proposons pour l’expression de l’évolution du déplacement en tête en fonction du nombre de cycles et de l’amplitude (Eq. 3) :

35,0

1

ln1,01

F

DFnyyn (3)

3.2 Moment maximum

Le moment maximum est l’un des paramètres dimensionnant et il est important d’examiner son évolution lors de chargements cycliques (Figure 5).

0,98

1

1,02

1,04

1,06

1,08

1,1

0 10 20 30 4

Mom

ent m

axim

um re

latf

(.)

Nombre de cycles (.)

P344 ; DF = 600 N

P33 ; DF = 600 N

P36 ; DF = 450 N

P347 ; DF = 450 N

0

P32 ; DF = 300 N

P346 ; DF = 150 NP318 ; DF = 150 N Valeur modèle ID = 86 %

sFigure 5. Evolution des moments maximums relatifs en fonction du nombre de cycles pour différents rapports DF/F.

Comme pour les déplacements relatifs, il est possible d’interpoler l’évolution du moment maximum relatif en fonction du nombre de cycles par une fonction de type logarithmique (Eq. 4) :

naM

nM

ln11

(4)

L’effet des cycles sur le moment maximum s’avère faible, inférieur à 8 % pour 15 cycles (Rosquoët, 2004) et à 12 % pour 75000 cycles (Rakotonindriana, 2009). De plus, nous avons montré que les valeurs du coefficient représentant l’effet des cycles sur le moment maximum sont du même ordre de grandeur que l’incertitude sur ce coefficient (tableau 3). Par conséquent, dans le cas d’un sable sec et dense, nous proposons à ce stade d’appliquer une majoration forfaitaire de 10% au moment maximum observé sous la charge statique F (solution conservative).

Tableau 3 : Estimation de a et de l’incertitude a associée (F = 960 kN, ID = 86 %). Essai Nombre de

cyclesDF/F(.)

a(.)

a

(.)P33 14 1 0 /P344 14 1 0 /P36 18 0,75 0,0047 0,0038 P347 40 0,75 0,0069 0,0052 P32 15 0,5 0,019 0,017 P318 25 0,25 0,026 0,014 P346 40 0,25 0,025 0,006

4 MÉTHODE LOCALE

Comme déjà proposé par plusieurs auteurs et adopté dans les règles API, nous admettons qu’il est possible de modéliser l’effet des cycles, sur les courbes P-y, par une diminution de la réaction P (pour un déplacement y donné) dans les couches de surface. Cette approche présente l’avantage de pouvoir être mise en œuvre avec tout logiciel de calcul de pieu sous charge latérale statique. Pour quantifier l’influence des cycles sur la « dégradation » du sol nous introduisons un coefficient d’abattement rc qui dépend a priori de cinq paramètres : la profondeur z, le déplacement du pieu y, le nombre de cycles n, la charge appliquée F et l’amplitude des cycles DF.

4.1 Méthode itérative pour déterminer un coefficient d’abattement rc

Le coefficient d’abattement rc des courbes P-y statiques va être déterminé à l’aide d’un calcul itératif par calage progressif entre les données expérimentales (état du pieu au 15ème cycle) et les données calculées par le logiciel Pilate (courbe P-y statique abattues d’un coefficient r).

0,0

1,2

2,4

3,6

4,8

6,0

7,2

8,4

9,6

10,8

12,0

-10 10 30 50 70 90 1

Prof

onde

ur (m

)

Déplacement (mm)

Valeur prototypeEssai P36ID = 86 %

10

22 mm =

50 mm = 2

85 mm = 4

Figure 6. Evolution du déplacement en fonction de la profondeur.

Pour une profondeur comprise entre 0 et 3,6 m, si l’on note le déplacement du pieu au niveau de la couche comprise entre 2,4 et 3,6 m, on constate que le déplacement est proche de 2 entre 1,2 et 2,4 m et 4 entre 0 et 1,2 m (Figure 6).

0

2

4

6

8

10

12

-10 40 90 140

Prof

onde

ur (m

)

Déplacement (mm)

Pilate : charge statique

Expérimental : charge statique

Pilate : charge cyclique

Expérimental : charge cyclique

Valeur prototypeP36 ; F = 960 kN ; DF = 720 kN15 cyclesId = 86 %En surface z = 0 m

16,1

16,1

15

15

1

1

Pilate

alExpériment

Pilate

alExpériment

yy

yy


Dans un premier temps, les coefficients d’abattement rc sont déterminés par calage sur les données expérimentales obtenues sur le déplacement de la tête du pieu en fonction du nombre de cycles. Pour représenter l’effet des cycles, les courbes P-y

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statiques de surface (jusqu’à 3,6 m) sont affectées d’un facteur rc sur la réaction P inférieur à 1 et croissant avec la profondeur.

Compte tenu du constat tiré de la figure 6, on impose que le coefficient r soit égal à 1 - 4 entre 0 et 1,2 m ; 1 - 2 entre 1,2 et 2,4 m et 1 - entre 2,4 et 3,6 m (avec un coefficient strictement positif tel que 1 - soit inférieur ou égal à 1). Par itération, nous recherchons la valeur de qui conduit au rapport correct entre le déplacement au 15ème cycle y15 et le déplacement du pieu sous la charge maximum y1 (Figure 7).

La même procédure a été utilisée pour les autres cas, de charge maximum de 960 kN, d’amplitude des cycles variant entre 960 et 240 kN. Il est possible d’exprimer le coefficient r en fonction du rapport d’amplitude de la variation de la charge sur la charge DF / F, pour les couches de surfaces entre 0 et 5B (Figure 8).

0,7

0,75

0,8

0,85

0,9

0,95

1

0,2 0,4 0,6 0,8 1

r (.)

DF / F (.)

3 B < z < 5 B

1,5 B < z < 3 B

0 < z < 1,5 B

Figure 8. Evolution du coefficient rc en fonction de DF / F (pour 15 cycles).

L’extrapolation des courbes de la figure 8 permet de déterminer des expressions simples du coefficient rc en fonction du rapport DF / F pour les trois couches de surfaces (Tableau 4).

Tableau 4 : Extrapolation par des droites de l’évolution du coefficient rcn fonction de DF / F (pour 15 cycles). e

Profondeur z Expression de rc en fonction de DF / F Bz 5,10

FDFr 12,087,0

BzB 35,1 F

DFr 058,094,0

BzB 53 F

DFr 029,097,0

5 COMPARAISON MÉTHODE GLOBALE / MÉTHODE LOCALE POUR LE DÉPLACEMENT EN TÊTE.

Un exemple de comparaison entre les deux méthodes est donné sur la figure 9 pour l’essai sur un pieu dans un sable dense (ID = 86 %) et des cycles caractérisés par F = 960 kN et DF = 720 kN.

-2

0

2

4

6

8

10

12

-10 40 90 140 190 240

Prof

onde

ur (m

)

Déplacement (mm)

Méthode locale : Pilate

Expérimental : charge cyclique

Déplacement en tête (méthode global)

Déplacement en tête (expérimentale)

Valeur prototypeP36 ; F = 960 kN ; DF = 720 kN15 cyclesId = 86 %Abattement courbe P-y (entre 0 et 3,6 m)


L’écart entre la méthode globale déduite de la relation présentée dans le tableau 4 et la courbe expérimentale est de 2% soit de l’ordre de grandeur de l’incertitude sur la mesure du

déplacement (l’incertitude sur le déplacement est de 0,104 mm, l’écart entre méthode locale et expérimentale est de 0,112 mm).

6 CONCLUSION

Une nouvelle analyse de l’étude paramétrique réalisée à l’IFSTTAR sur un pieu soumis à une charge latérale cyclique dans un sable sec et dense (Rosquoët 2004) et complétée par la suite (Rakotonindriana 2009, Khemakhem 2012) permet de proposer de nouvelles méthodes de dimensionnement des pieux sous chargement latéral cyclique. Une méthode globale, basée sur une loi logarithmique en fonction de DF/F permet de donner le déplacement en tête pour un nombre de cycles n donné. On notera que l’effet des cycles sur le moment maximum, dans le cas d’un sable sec et dense, est négligeable.

Une méthode locale basée sur l’abattement des courbes P-y statiques a été réalisée. Des calculs itératifs ont permis de valider la modélisation de l’effet des cycles sur le déplacement et sur les moments par une modification des courbes P-y d’un facteur rc croissant entre 0 et 5B (soit entre 0 et 3,6 m). Des expressions simples ont été proposées permettant de déterminer ce facteur rc à partir du rapport DF / F.

On note, pour finir une bonne concordance des résultats obtenus par les deux méthodes. Toutefois si la méthode globale est utilisable pour un très grand nombre de cycles (> 500), la méthode locale n’est valide que pour une quinzaine de cycles et pour des chargements cycliques non alternés.

7 REFERENCES

API - American Petroleum Institute. 2002. Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms - Working Stress Design. RP 2AWSD, Washington, DC.

Bouafia A. 1994. Etude expérimentale du chargement latéral cyclique répété des pieux isolés dans le sable en centrifugeuse. Canadian Geotechnical Journal, Vol. 31, n°5, 740-748.

Frank R. 1999. Calcul des fondations superficielles et profondes.Editions Technique de l’Ingénieur et Presse des Ponts et Chaussées. Paris.

Garnier J. 2001. Modèles physiques en géotechnique I-Evolution des techniques expérimentales et des domaines d’application. RevueFrançaise de Géotechnique. N°97. 3-29.

Hettler. A. 1981. Verschiebungen starrer und elastischer Gründungskörper in Sand bei monotoner und zyklischer Belastung. Institut für Boden-und Felsmechanik der Universität Karlsruhe,Heft Nr. 90.

Khemakhem M. 2012. Etude expérimentale de la réponse aux charges latérales monotones et cycliques d’un pieu foré dans de l’argile. Thèse de Doctorat. Ecole Centrale de Nantes Université de Nantes.336 p.

Lin S.S. & Liao J.C. 1999. Permanent strains of piles in sand due to cyclic lateral loads. Journal of Geotechnical and Geoenvironemental Engineering, Vol. 125, n°9, 798-802.

Li Z. Haigh S. K. & Bolton M. D. 2010. Centrifuge modelling of mono-pile under cyclic lateral loads. 7th International Conference on Physical Modelling in Geotechnics. Zurich. Vol. 2, 965-970.

Peralta P. and Achmus M. 2010. An experimental investigation of piles in sand subjected to lateral cyclic loads. 7th International Conference on Physical Modelling in Geotechnics. Zurich. Vol. 2, 985-990.

Rakotonindriana M. H. J. 2009. Comportement des pieux et des groupes de pieux sous chargement latéral cyclique. Thèse de doctorat. Ecole Nationale des Ponts et Chaussées. 381p.

Rosquoët F. 2004. Pieu sous charge latérale cyclique. Thèse de Doctorat. Ecole Centrale de Nantes Université de Nantes. 305 p.

Verdure L. Levacher D. & Garnier J. 2003. Effet des cycles sur le comportement d’un pieu isolé dans un sable dense sous chargement latéral. Revue Française de Génie Civil. 7/2003. 1185-1210.

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Behavior of marine silty sand subjected to long term cyclic loading

Comportement du sable limoneux marin soumis à une charge cyclique de longue durée

Safdar M., Kim J.M. Pusan National University, Busan, South Korea

ABSTRACT: The foundations for offshore wind turbines are demanding due to the dynamic nature of the offshore loading. A greater understanding of the behavior of wind turbine foundation soil, will certainly lead to the stable construction of foundations which inturn, will make offshore wind farms a more feasible part of the solution to the global energy problem. This paper presents the resultsof cyclic direct simple shear test (CDSS) to explain the long term cyclic behavior of marine silty sand. Cyclic behavior of marine sandare based on the number of loading cycles, cyclic shear strain amplitude, relative density, and cyclic stress ratio. These results aremodeled and can be applied to design offshore wind turbine foundations.

RÉSUMÉ : Les fondations pour les éoliennes offshore sont principalement exigeante en raison de la nature dynamique du chargement offshore. Une meilleure compréhension du comportement de l'éolienne des sols de fondation, va certainement conduire à laconstruction des fondations stables qui à leur tour, feront de parcs éoliens en mer un rôle plus possible de la solution au problème mondial de l'énergie. Ce document présente les résultats d'essai de cisaillement cyclique directe simple (CDSS) pour expliquer lecomportement cyclique à long terme de sable limoneux marin. Comportement cyclique de sable marin sont basés sur le nombre de cycles de charge, cyclique d'amplitude de contrainte au cisaillement, la densité relative et du taux de contrainte cyclique. Ces résultatssont modélisés et peut être appliquée à la conception fondations d'éoliennes off-shore.

KEYWORDS: Cyclic Loading, Offshore Wind Turbine, CDSS, Cyclic Stress Ratio

1 INTRODUCTION

Understanding the behavior of offshore marine sand subjected to long term cyclic loading is very vital in solving several offshore geotechnical problems. Several researchers have studied behavior of clay and sand subjected to cyclic loading.

(Vucetic et al. 1988) studied the degradation of marine clays under cyclic loading. (D. Wijewickreme et al. 2005) studied the cyclic loading response of loose air-pluviated Fraser river sand. (K.H. Andersen 2009) investigated in detail, the bearing capacity of the soil under cyclic loading, and stated that the cyclic shear strength and the failure mode under cyclic loading depend on the stress path and the combination of average and cyclic shear stresses. Safdar et al., 2013, studied the cyclic behavior of marine silty sand subjected to symmetrical cyclic loading. Different approaches have been made as an attempt to include cyclic loading in the design procedure of offshore wind turbine foundation (Soren et al. 2012).

1.1 Stress controlled CDSS test

Constant volume direct simple shear (DSS) test is a reliable method for measuring undrained shear strength of undisturbed or compacted soil samples. The DSS test is most similar to the CU triaxial test in that samples are consolidated prior to shearing. The simple shear is the test condition that only normal and shear stress acting on top face of a specimen is defined, whereas the displacement constraints exist for the other boundaries: The bottom face of specimen is theoretically fixed, and the radial strain on specimen is zero.

The CDSS test procedure is based on that of a constant-volume direct simple shear testing of soils, which has been studied extensively for half a century and is described in the standard ASTM D6528-07. The sample is consolidated under a

normal load within a wire-reinforced membrane (in this study) or a stack of thin rings that provide lateral confinement.

Once consolidation is complete, a horizontal shear force is applied to one end of the sample. The sample height is continuously maintained during shear to ensure constant volume. Rather than measuring pore pressures, which would require complete saturation of the sample, the pore pressure response is inferred from the change in vertical stress which is monitored throughout the test (Baxter et al 2010). In this way changes in applied vertical stress (Δσv), which are required to keep the sample height constant, are assumed to be equal to the excess pore water pressure (Δu) that would develop if the test were truly undrained with pore pressure measurements (Finn, 1985, Dyvik et al. 1987).

Figure 1 Simple Shear Condition, (Dyvik et al 1987)

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Figure 2 (a). Wire Reinforced Membrane (b). Stacked Rings (Baxter et al 2010)

The cyclic shear response of natural low-plastic Fraser

River silt was investigated using constant-volume direct simple shear (DSS) testing (Wijewickreme 2010). Maria V. Sanin et al., (2011) studied the cyclic shear response of undisturbed and reconstituted Fraser River Silt. A soil can be subjected to many different stress conditions, being purely cyclic stress, static or average stress, or a combination of both. Andersen (2009) shows this clearly in a study on Drammen clays at the NGI. Drammen clays samples are tested in cyclic triaxial and cyclic simple shear conditions for different combinations of static and cyclic shear stresses. In this study cyclic simple shear tests have been performed with static or average shear stress, = 0 or symmetrical cyclic loading.

1.2 Sample Preparation

Air Pluviation with dry compaction approach was developed to produce samples of the silty sand with consistent heights and initial relative densities. The equipment used consists of the shear box having bottom cap, two o-rings, wire-reinforced membrane, top cap, triaxial pressure panel, and compacting hammer. Sample diameter is 63.5mm and height is varied from 20 to 25 mm to maintain height to diameter ratio less than 0.4, in order to fulfill the ASTM D6528-07 criteria. In this study marine silty sand is obtained from the West coast of South Korea. Specific gravity of material tested is Gs =2.65. Marine silty sand has minimum voids ratio of 0.74 and maximum voids ratio of 1.18. Details of properties of soil tested are given in Table 1. T able 1 Properties of marine silty sand.

Figure 3 Stress-path responses of NC Fraser River silt under constant volume cyclic DSS loading (σ’v = 97 kPa; CSR = 0.20; =0; OCR = 1.0) (Maria V. Sanin et al., 2011).

Figure 5 Constant volume cyclic DSS test on undisturbed Fraser River Delta silt. (σ’v = 100 kPa, CSR = 0.14) (Maria V. Sanin et al., 2011).

Figure 5 Stress-strain behavior under different loading conditions (Andersen, 2009)

1.3 Testing Program

The laboratory testing program for this study was designed to analyze the behavior of marine silty sand when subjected to cyclic loads for different combinations of parameters such as cyclic stress ratio, no. of loading cycles and relative density.

For marine silty sand, the tests were performed at a frequency of 0.1 Hz. Effect of Relative Density (Dr %) for 65, and 70 percent is studied for various CSR and no. of loading cycles. Marine silty sand has minimum voids ratio of 0.74 and maximum voids ratio of 1.18. Specific gravity of Gs=2.65.

To produce in-situ (K0) stress conditions, a vertical consolidation stress must be applied to the sample prior to shearing. Applied vertical stresses simulate the loads from overburden material located over the soil sample. For marine

2401


silty sand, a normal consolidation stress of 100 KPa was applied in one step for all the specimens.

1.4 Cyclic Direct Simple Shear test results

Several researchers have used different shear strain failure criteria such as 3%, 4%, 5% and 7.5%. A shear strain failure condition was used for tests performed on marine silty sand and the failure criterion was established as 4% double amplitude shear strain. The results of a test are shown in Figure 6, 7 and 8.Figure 8 shows the applied cyclic shear stress of +/- 12 kPa (CSR = 0.12). Figure 6 shows the development of shear strain throughout the test, which reaches 4% double amplitude shear strain at nearly 205 cycles.

8 CDSS tests were conducted, 4 tests at voids ratio of 0.898 (Dr 65%) and CSR range of 0.10, 0.12, 0.14, and 0.16. 4 tests at voids ratio of 0.847 (Dr 70% and CSR range of 0.10, 0.12, 0.14, and 0.20. All the tests were conducted at nominal initial effective confining stress of 100 kPa to provide a basis for comparison between tests. In case of 70% relative density, Figure 6 shows the degradation curve and development of shear strain with increasing number of cycles. In a general sense, marine silty sand specimens seem to exhibit gradual increase in shear strain and degradation of shear stiffness with increasing number of load cycles. Typical stress paths and stress-strain curves of tests conducted on marine silty sand specimens are presented in Figures 7 and 8 respectively.

Figure 6 Peak-Peak Shear Strain vs No. of Loading cycles for (σ’v = 100 kPa, CSR=0.12 and Dr (%) = 70).

Figure 6 shows that samples having relative density of 70% reached 2% cyclic double amplitide cyclic shear strain after 42 cycles and 205 to reach failure.

Figure 7 Stress Path During constant volume cyclic DSS loading of silty sand for (σ’v = 100 kPa, CSR=0.12 and Dr (%) = 70).

Figure 8 Stress-strain response of marine silty sand under constant volume cyclic DSS loading (σ’v = 100 kPa; Dr (%) = 70, CSR = 0.12; α= 0.0; OCR = 1.0).

In case of 65% relative density, Figure 9 shows the

degradation curve and development of shear strain with increasing number of cycles. Typical stress paths and stress-strain curves of tests conducted with 65% relative density on marine silty sand specimens are presented in Figures 10 and 11 respectively.

It was observed that specimens having higher relative densities require higher no. of loading cycles to reach 4% double amplitude cyclic shear strain and specimens having lower relative density reach to failure in smaller no. of loading cycles. In case of higher cyclic stress ratio (CSR) the soil samples reached the failure criterion in few no. of loading cycles.

Figure 9 Peak-Peak Shear Strain vs No. of Loading cycles for (σ’v = 100 kPa; Dr (%) = 65, CSR = 0.12; α= 0.0; OCR = 1.0).

Figure 9 shows that samples having relative density of 65%

reached 2% cyclic double amplitide cyclic shear strain after 15 cycles and 57cycles to reach failure.

Figure 10 Stress Path During constant volume cyclic DSS loading of silty sand for (σ’v = 100 kPa; Dr (%) = 65, CSR = 0.12; α= 0.0; OCR = 1.0).

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Samples having 70% relative density and subjected to CSR of 0.12 reached cyclic double amplitude of 4% shear strain at nearly 205 cycles. In case of samples having 65% relative density and subjected to CSR of 0.12 reached cyclic double amplitude of 4% at nearly 57 cycles.

2 SUMMARY AND CONCLUSIONS

The constant volume cyclic shear response of marine silty sand was examined using data from CDSS tests. The intent was to compare the shear response of the silty sand specimens under different relative densities.

Figure 11 Stress-strain response of marine silty sand under constant volume cyclic DSS loading (σ’v = 100 kPa; Dr (%)= 65, CSR = 0.12; α= 0.0; OCR = 1.0).

samples having higher value of cyclic stress ratio and/or low relative density fail at a small number of loading cycles. It is found in this particular study that the number of loading cycles required to reach the threshold strain is not much different for two relarive densities after 300 loading cycles. The trends are little different from the previous study reported in which the curves are almost parallel even in high number of cycles.

3 ACKNOWLEDGEMENT

This research is supported by the Ministry of Land, Transport and Maritime Affairs, Korea 2010 research grant (2010 Construction Technology Innovation Program, 10-CTIP-E04).

4 REFERENCES

Andersen K.H 2009. Bearing capacity under cyclic loading- offshore, along the coast, and on land. The 21st Bjerrum Lecture presented in Oslo, 23November 2007”. NRC Research Press Web site (www.cgj.nrc.ca)

Baxter C.D.P, Bradshaw A.S., Ochoa-Lavergne M. and Hankour R. 2010. DSS Test Results using Wire-Reinforced Membranes and Stacked Rings. GeoFlorida 2010 ASCE.

Boulanger R.W & Seed R.B. 1995 . Liquefaction of sands under Bidirectional monotonic and cyclic loading. Journal of Geotechnical Engineering ASCE Vol. 121, No. 12 pp. 870-

Figure 12 CSR versus No. of Loading cycles to reach double amplitude shear strain of 4% for marine silty sand

878. Bjerrum, L., and Landva, A. 1966. Direct simple-shear tests on a

Figure 12 shows the number of loading cycles versus cyclic stress ratio that reach shear strain of 4%. As expected, samples having higher value of cyclic stress ratio and/or low relative density fail at a small number of loading cycles. It is found in this particular study that the number of loading cycles required to reach the threshold strain is not much different for two relarive densities after 300 loading cycles. The trends are little different from the previous study reported (Fig. 13) in which the curves are almost parallel even in high number of cycles.

Norwegian quick clay. Geotechnique 16(1), 1-20. Dyvik, R., Berre, T., Lacasse, S., and Raadim, B. 1987. Comparison of truly undrained and constant volume direct simple shear tests. Geotechnique 37(1), 3-10. Idriss, I. M., and Boulanger, R. W. 2008. Soil liquefaction during Earthquakes. Monograph MNO-12, Earthquake Engineering Research Institute, Oakland, CA. Maria V. Sanin, Wijewickreme D. 2011. Cyclic shear response of undisturbed and reconstituted Fraser River Silt. Pan-Am CGS Geotechnical Conference. Safdar M., Kim, J.M. (2013). “Cyclic Behavior of Marine Silty Sand”

Electronic Journal of Geotechnical Engineering (EJGE)

Vol.18A, pp. 209-218. Soren K. N., Amir Shajarati, K.W.Sorenson, L.B. Ibsen 2012. Behaviour of Dense Frederikshavn sand during cyclic Loading. DCE Technical Memorandum No. 15 Vucetic M. & Dobry R. 1988. Degradation of marine clays

under cyclic loading. Journal of Geotechnical Engineering ASCE Vol. 114, No. 2 pp. 133-149.

Wijewickreme D. 2010. Cyclic shear response of low plastic Fraser River silt. Proceedings of the 9th U.S. National and 10th Canadian conference on Earthquake Engineering Wijewickreme D., Sanin M.V. and Greenaway G.R. (2005)

Cyclic shear response of fine-grained mine tailings. Canadian Geotechnical Journal Vol. 42 pp. 1408-1421

Wijewickreme, D., Sriskandakumar, S., and Byrne, P. 2005. Cyclic loading response of loose air-pluviated Fraser River sand for validation of numerical models simulating centrifuge tests. Canadian Geotechnical Journal Figure 13 Effect of stress densification on cyclic resistance of loose air-

pluviated sand (Wijewickreme et al., 2005) 42(2), 550-561.

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Influence des chargements cycliques axiaux dans le comportement et la réponse de pieux battus dans le sable

Influence of cyclic axial loads in the behaviour and response of driven piles in sand

Silva M., Foray P. Laboratoire 3SR, Grenoble, France

Rimoy S., Jardine R. Imperial College London, London, UK

Tsuha C. University of Sao Paulo, Sao Paulo, Brazil

Yang Z. Zhejiang University, Zhejiang, China

RÉSUMÉ: Cet article présente un nouveau diagramme de stabilité cyclique pour le frottement latéral ainsi que l’effet sur la capacité en traction des pieux modèle instrumentés soumis à des chargements cycliques axiaux suite à une large série d’essais en chambred’étalonnage dans du sable siliceux. Les mesures locales des contraintes dans le sol (verticales, radiales et orthoradiales) à différentes distances de l’axe du pieu, ainsi que les mesures tangentielles et radiales à l'interface sol-pieu, permettent une analyse détaillée de l’évolution des chemins de contraintes locaux autour du pieu et au sein du massif lors des chargements cycliques et post-cycliques.

ABSTRACT: This paper presents a new cyclic stability diagram for the lateral friction and the effect on the tensile capacity ofinstrumented model piles subjected to axial cyclic loading across a wide range of calibration chamber testing in silica sand. Localmeasurements of stresses in the soil mass (vertical, radial and orthoradial) at different distances from the axis of the pile, as well asshear and radial stresses at the soil-pile interface, provide a detailed analysis of the evolution of local stress paths around the pile andthe soil mass during cyclic and post-cyclic loading.

MOTS-CLÉS: Chambre d’étalonnage, chargement cyclique axial, interaction sol-pieu, diagramme de stabilité cyclique.

1 INTRODUCTION

Des fondations sur pieux de plateformes de pétrole/gaz et des turbines éolienne/marémotrice peuvent être soumises à des chargements cycliques de très longue durée en raison des conditions environnementales. Les longs pieux battus, installés par des cycles de charge-décharge, subissent une dégradation sévère du frottement lors de leur installation (Lehane et al. 1993, Kolk et al, 2005). Celle-ci peut être compensée par un phénomène de « cicatrisation » avec le temps (« ageing »), observé dans des expériences in-situ (Chow et al 1998, Jardine et al. 2006). Selon leur finalité (fondation d’ouvrages offshore ou d’éoliennes), ces pieux sont ensuite souvent soumis à des sollicitations cycliques de service (houle, vibrations) qui affectent également leur frottement. La réponse cyclique de l’interface sol-pieu a été étudiée en laboratoire à l’aide de pieux modèle dans des chambres d’étalonnage pressurisées (Chan & Hanna 1980, Al-Douri & Poulos 1994, Chin and Poulos 1996, Le Kouby et al. 2004). Ces essais ont montré que la contrainte de cisaillement décroît d’autant plus que l’amplitude des déplacements tangentiels augmente. Jardine et al. (2006) ont montré à partir des essais in-situ réalisés à Dunkerque (Chow et al. 1998) qu’alors que des cycles de grande amplitude dégradaient la capacité du pieu en frottement, des cycles non alternés de faible amplitude accéléraient les effets bénéfiques du phénomène de “cicatrisation” (augmentation de la capacité avec le temps). Le Kouby et al. (2004) donnent des résultats similaires sur un pieu modèle de 20 mm de diamètre.

L’étude présentée ici a pour but de donner une meilleure compréhension, à une échelle de laboratoire et dans un environnement contrôlé, des résultats obtenus sur des pieux réels à Dunkerque par Jardine & Standing (2000), en mesurant, grâce à une instrumentation exceptionnelle de la chambre d’étalonnage et du pieu, les chemins de contraintes locaux, le long du pieu et au sein du massif lors des différentes phases d’installation, de repos et de chargement statique et cyclique du pieu. Ce travail fait partie d’un programme de recherche

commun entre le laboratoire 3SR et Imperial College London, ainsi que du projet National SOLCYP.

2 DISPOSITIF EXPÉRIMENTAL

2.1 La chambre d’étalonnage

La grande chambre d’étalonnage du Laboratoire 3S-R a été adaptée de façon à maîtriser les conditions environnementales (température, pression, alimentation électrique) sur des périodes de longue durée (plusieurs mois), afin de pouvoir mesurer les phénomènes de « cicatrisation » du frottement après cyclage et après une longue période de repos. Une isolation complète de la cuve ainsi qu’un système de chauffage/refroidissement des parois ont été réalisés de façon à minimiser les variations de température dans la cuve. La pressurisation des membranes a été conçue pour fonctionner en cas de panne du réseau.

La chambre d’étalonnage de Grenoble est composée de trois éléments cylindriques de 50 cm de hauteur et d’un diamètre interne de 1.2 m. Le fond de la chambre et le couvercle supérieur sont constitués de plaques rigides d'une épaisseur de 100 mm. Une pression verticale d’environ 150 kPa est imposée par une membrane en caoutchouc remplie d’eau, qui est fixée sous le couvercle supérieur et contrôlée par un système d’interface eau/air. Une membrane en latex de 2 mm d'épaisseur a été installée sur la paroi interne de la chambre afin de mieux fournir des conditions K0 au bord de la cuve. Une couche de graisse de silicone entre cette membrane et la paroi de la chambre permet de réduire le frottement entre le mur et la masse de sable et d'assurer une meilleure homogénéité des contraintes à l'intérieur du sol.

Un système d’isolation thermique formé de serpentins en cuivre autour de la cuve permet de travailler avec une température presque uniforme entre 18 et 19 degrés.

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9R

26R

46R

Verin electric

100 mm 60 mm

~99

0 m

m

enfo

ncem

ent

18 oC Sable de Fontainebleau

Membrane supérieure Niveau de référence

Cellule de charge

±0.00 m

-0.33 m

-0.52 m

-1.50 m

Niveau supérieur de capteurs, h/R=40

Niveau inférieur de capteurs, h/R=14.4

Système de guidage

Serpentins en cuivre autour de la cuve pour contrôle de la température

-1.05 m

Mini-ICP

-0.79 m

Niveau intermédiaire de capteurs, h/R=29.4

Membrane en latex -1.30 m

Membrane inférieure (non pressurisée)

Figure 1: Modifications apportées à la chambre d’étalonnage du laboratoire 3S-R.

Pour les essais présentés ici, la chambre d'étalonnage a été remplie avec du sable siliceux de Fontainebleau NE34 commercialisé par l’entreprise Sibelco. En utilisant une technique de remplissage par pluviation, nous obtenons une densité du massif moyennement dense à dense, Dr=72%. Le tableau 1 montre les propriétés mécaniques du sable.

Tableau 1: Propriétés mécaniques du sable NE34

Gs D10(mm)

D50(mm)

D60(mm) emax emin

max(kN/m3)

min(kN/m3)

2.65 0.150 0.210 0.230 0.90 0.51 17.2 14.2

Plusieurs essais pénétrométriques ont été effectués sur des échantillons de sable NE34 sous 150 kPa avec une résistance en pointe quasi-constante de 21 ± 2 MPa. Des détails supplémentaires sur les propriétés mécaniques du sable sont décrits dans Yang et al. (2010).

2.2 Le pieu d’essai Mini-ICP

Le pieu utilisé correspond au pieu modèle Mini-ICP développépar l’Imperial College London et décrit par Jardine et al. 2009. C’est un pieu tubulaire en acier à base fermée de 36mm de diamètre avec un cône solide de 60° à sa pointe. Il est composé de trois niveaux d'instrumentation (ou clusters) identifiés selon leur distance à la pointe en: A (9R), B (24) et C(44), R étant le diamètre du pieu. Chacun de ces clusters contient une cellule de charge axiale, qui peut être utilisée pour calculer la moyenne de frottement dans le fût du pieu; un capteur de contraintes en surface (SST: surface stress transducer) pour mesurer la contrainte totale radiale et la contrainte de cisaillement; un capteur de température MEMS et un inclinomètre. À partir de l'essai ICP03, une cellule de charge axiale a été ajoutée près de la pointe du pieu pour mieux séparer les contributions de la pointe et du frottement latéral dans la capacité totale du pieu.

Le pieu a été installé jusqu’à une profondeur finale de 0.98m par fonçage avec des cycles de charge-décharge en simulant la dégradation des efforts radiaux comme dans une installation par battage. La vitesse de pénétration varie entre 0.5 et 2 mm/s et l’amplitude des cycles d’enfoncement successifs est de 5, 10 ou 20 mm. La décharge de chaque cycle a été menée jusqu’à obtenir une charge en tête du pieu nulle.

2.3 Mesure des contraintes dans le massif

36 mini-capteurs Kyowa et TML de capacités variant de 500 kPa à 7 MPa selon leur distance au pieu, ont été mis en place sur trois niveaux comportant chacun douze capteurs, de façon à mesurer les contraintes verticales, radiales et orthoradiales sur des rayons concentriques à des distances de 2R à 16R. Leur étalonnage sous chargement cyclique a été effectué à Imperial College en suivant le protocole établi par Zhu et al 2009. La réponse de chaque capteur est représentée par une série de courbes d’hystérésis selon l’histoire de charge précédente.

3 PROGRAMME EXPÉRIMENTAL

Après une certaine période de cicatrisation post-installation, un premier essai en compression et un essai en traction ont été effectués pour définir la capacité du pieu avant les chargements cycliques. Plusieurs séries d’essais de faible et de haut niveau de chargement cyclique ont ensuite été effectuées, ainsi que des essais de traction pour évaluer l'effet des charges dans la réponse du pieu à l'arrachement.

Les essais cycliques ont été caractérisés selon deux paramètres de charge; Qcyclic et Qmean correspondant respectivement à la moitié de l’amplitude de chaque cycle et à la valeur moyenne de charge. Les essais de faible niveau en charge ont été effectués sous chargement contrôlé (FC), uniquement en traction (non alterné). Ces essais ont mobilisé jusqu'à 60% de la capacité totale du pieu en tension, Qt. Les cycles de charge de haut niveau alternés (compression et traction) ont été conduits en déplacement contrôlé (DC). Les cycles ont été appliqués à une fréquence relativement basse, allant de moins de 0,5 cycles par minute pour les essais alternés, à presque 2,5 cycles par minute pour les essais de faible charge.

Des essais de traction statique post-cyclique ont été réalisés, avec un taux de déplacement de 0,01 mm/s, afin de vérifier les capacités à l'arrachement disponibles après chaque chargement cyclique. Comme indiqué par Jardine et al (2006), les essais préalables peuvent influer sur les capacités développées lors des essais ultérieurs. Le tableau 2 résume les différentes séries d’essais cycliques réalisées.

Tableau 2: Programme d’essais de chargement cyclique

ID N of cycles Description Qcyclic

/QT

Qmean/ QT

ICP1-OW1 (FC) 1000 0 à -4,0 kN 0,22 0,22 ICP1-TW1 (DC) 100 -4~5 mm 0,41 0,06 ICP2-OW1 (FC) 1000 0 à -3,0kN 0,12 0,12 ICP2-OW2 (FC) 1000 0 à -4,8 kN 0,20 0,20 ICP2-OW3 (FC) 500 0 à -6,8 kN 0,28 0,28 ICP2-TW1 (DC) 100 -2,0 ~ 3,0 mm 0,48 0,15 ICP3-OW1 (FC) 100 0 à -9,6 kN 0,38 0,38 ICP3-TW1 (FC) 287 -5,0 à +8,0kN 0,54 0,08 ICP3-TW2 (FC) 199 -5,0 à +5,0kN 0,40 0,06 ICP3-TW3 (FC) 50 -5,0 à +7,0 kN 0,44 0,02 ICP3-TW4 (FC) 37 -5,0 à +10,0kN 0,44 0,02 ICP4-OW1 (FC) 7000 0 à -3,5 kN 0,15 0,15 ICP4-TW1(FC) 600 -4,0 à +4,0kN 0,23 0,06 ICP4-OW2 (FC) 50 -2,3 à -4,6 kN 0,21 0,63

OC = cycles non alternés, et TC = cycles alternés

4 RÉSULTATS DES ESSAIS CYCLIQUES NON ALTERNÉS À FAIBLE AMPLITUDE

4.1 Evolution des contraintes le long du pieu et au sein du massif

Les mesures de contrainte effectuées au sein du massif et le long du pieu permettent de préciser les mécanismes conduisant à l’amélioration ou à la dégradation du frottement. La Figure 2montre l’évolution des contraintes radiales dans le sol à différentes distances du pieu. On constate une décroissance de ces contraintes radiales avec le nombre de cycles, plus accentuée lors de cycles alternés. La même situation est vérifiée avec les contraintes radiales le long du pieu, accentuée lors des grands cycles mais avec des amplitudes plus grandes. On remarque également dans la Figure 2, que le chargement statique effectué entre les deux séries de cycles entraîne une nette augmentation des contraintes près du pieu après les petits cycles (effet de la dilatance).

La distribution des contraintes dans le massif du sol à différentes distances de l’axe du pieu a été interprétée par Jardine et al. 2012 en utilisant l’information des capteurs.

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0 2 4 6 8 10 12 14 1620406080

100120140160180200220

Cont

rain

te (k

Pa)

Temps (heures)

r/R=2 r/R=3 r/R=5 r/R=8 r/R=20

Figure 2: Evolution des contraintes radiales dans le sol : De 0 à 9.6h : 1000 cycles non alternés de faible amplitude (ICP1-OW1) ; de 10,4 à 14,7h : 100 cycles alternés de grande amplitude (ICP01-OW1).

La mesure simultanée des contraintes radiales et tangentielles permet de tracer les chemins de contraintes suivis par le sol au contact du pieu. Les petits cycles non alternés, tels que ceux représentés sur la Figure 3 pour les trois niveaux le long du pieu, provoquent un déplacement des chemins de contraintes vers la gauche, traduisant une tendance à la contractance de l’interface et donc sa densification. C’est cette densification qui produit ensuite une augmentation de la dilatance et l’amélioration de la capacité. Lorsque l’amplitude de ces cycles non alternés augmente, les cycles s’approchent de la droite de rupture en traction (Yang et al 2010).

0 100 200 300 400 500

-200

-100

0

100

200 ClusterA ClusterB ClusterC

Direction dechemin de contraintes

'=27o

Cont

rain

te d

e ci

saill

emen

t rz (k

Pa)

Contrainte radiale r (kPa)Figure 3: Chemins de contraintes à la surface du pieu. Essai Mini-ICP2. Cycles non alternés de faible amplitude.

A l’inverse, les chemins de contraintes des grands cycles alternés traversent l’équivalent d’une « ligne de changement de phase » et produisent des alternances de dilatance/contractance et une désorganisation du sol autour de l’interface. On peut remarquer qu’on retrouve pour ces chemins de contraintes la forme en « ailes de papillon » qu’on observe lors d’essais de cisaillement cyclique à volume constant ou à rigidité normale contrôlée (Fakharian & Evgin 1997, Mortara et al. 2007).

0 100 200 300 400 500

-200

-100

0

100

200 ClusterA ClusterB ClusterC

Direction dechemin de contraintes

'=27o

Cont

rain

te d

e ci

saill

emen

t rz (k

Pa)

Contrainte radiale r (kPa)Figure 4: Chemins de contraintes à la surface du pieu. Essai Mini-ICP4. Cycles alternés de forte amplitude.

4.2 Capacité du pieu à l’arrachement avant et après chargement

Le tableau 3 montre les essais statiques effectués après chaque essai cyclique pour évaluer l’effet de ceux-ci sur la capacité à l’arrachement.

Tableau 3: Programme d’essais de chargement cyclique

Essai Essai précédent (selon Tableau II) QT (kN) Variation entre

essais (%) ICP1-T1 1ère compression 9,2 - ICP1-T2 ICP1-OW1 10,8 17,4% ICP1-T3 ICP1-TW1 4,9 -54,6% ICP2-T1 1ère compression 12,1 - ICP2-T2 ICP2-OW1 13,2 9,1% ICP2-T3 ICP2-OW2 14 6,1% ICP2-T4 ICP2-OW3 13,7 -2,1% ICP2-T5 ICP2-TW1 8,7 -36,5% ICP3-T1 1ère compression 12,5 - ICP3-T2 ICP3-OW1 10,9 -12,8% ICP3-T3 ICP3-TW1,2,3,4 4,8 -56,0% ICP4-T1 1ère compression 11,5 - ICP4-T2 ICP4-OW1 13,9 20,9% ICP4-T3 ICP4-TW1 5,5 -60,4% ICP4-T4 ICP4-OW2 6 9,1%

Comme indiqué, selon l’amplitude de la charge cyclique, il est possible d’obtenir une augmentation de la résistance, ce que proposent Jardine et al (2006) dans des essais sur le terrain.

0.0 0.5 1.0 1.5 2.0 2.5 3.0 3.50

2

4

6

8

10

12

14

16

Forc

e d'

arra

chem

ent (

kN)

LVDT (mm)

avant chargement

après essais de fortdéplacement

aprés essais non-alternésde faible amplitude

Figure 5: Courbes force d’arrachement–déplacement du pieu avant et après application des cycles Essai ICP2.

Les essais d’arrachement avant et après l’application de cycles, présentés dans la Figure 5 pour les essais Mini-ICP2, montrent clairement que l’application des cycles non alternés de

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faible amplitude améliore la capacité du pieu en raison de la densification de l’interface, alors que les cycles de grande amplitude avec passage de la traction à la compression produisent une nette dégradation de cette capacité due à la désorganisation de l’interface.

4.3 Diagramme de stabilité cyclique

Une première partie des résultats concernant les essais de chargement cyclique a été résumée par Tsuha (2012). On a analysé l’évolution du frottement latéral du pieu et sa stabilité selon le nombre et le type de cycles appliqués (réversibles ou non réversibles), l’amplitude et la valeur moyenne de la charge. La Figure 6 montre le diagramme de stabilité cyclique établi à partir de la totalité des essais réalisés. Trois zones ont été identifiées pour le comportement du pieu.

Stable, correspondant à une zone de faible amplitude au niveau de la charge et où le pieu peut être soumis à plus de 1000 cycles sans accumuler une déformation importante.

Métastable, correspondant à une déformation importante ou une rupture entre 100 et 1000 cycles.

Instable, correspondant à une rupture obtenue en moins de 100 cycles.

La rupture cyclique a été définie comme ; i) un déplacement permanent atteignant 10% du diamètre (i.e. 3.6 mm), ou ii) un taux de déplacement montrant une augmentation forte, en considérant comme « lent » un taux inférieur à 1 mm/104 cycle, et « rapide » un taux supérieur à 1 mm/100 cycle.

-0.2 0.0 0.2 0.4 0.6 0.8 1.00.0

0.2

0.4

0.6

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Qcy

clic/Q

T

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Stable

Unstable

>1000

Nf=1

10100

1410

4

1000

500

5

Nf=number of cycles to failure

Meta-Stable

66165

580

Figure 6: Diagramme de stabilité cyclique essais ICP1-4 (d’après Tsuha et al., 2012)

5 CONCLUSION

La mesure des chemins de contraintes le long de pieux instrumentés avec précision, ainsi que ceux au sein du massif de sol au voisinage du pieu, permet de mieux appréhender les mécanismes gouvernant la dégradation ou l’amélioration du frottement lors de sollicitations cycliques. Elle permet en particulier de faire le lien entre les états « stable », « métastable » et « instable » dans les diagrammes classiques de stabilité cycliques et le comportement élémentaire des interfaces par rapport à leur « ligne de transformation de phase » ou « droite caractéristique ».

L’ensemble des essais effectués jusqu’à présent montre que l’application d’un grand nombre de petits cycles non alternés produit une densification de l’interface sol-pieu, favorisant le développement d’une dilatance lors d’un chargement ultérieur et donc une amélioration de la capacité du pieu à l’arrachement.

A l’inverse les grands cycles alternés produisent une désorganisation de l’interface et une diminution de la dilatance

lors d’un chargement ultérieur et une dégradation de la capacité à l’arrachement. Ce phénomène est analogue à celui conduisant à une dégradation du frottement le long des pieux battus lors de leur installation.

6 RÉFÉRENCES

Al-Douri R. and Poulos H.G. 1994. Cyclic Behaviour of Pile Groups in Calcareous Sediments. Soils and Foundations 34, 49-59.

Chan S. and Hanna T. H. 1980. Repeated loading on single piles in sand. American Society of Civil Engineers. Journal of the Geotechnical Engineering Division 106 (2), 171-188.

Chin J. T. and Poulos H.G. 1996. Tests on model jacked piles in calcareous sand. Geotechnical Testing Journal 19(2), 164-180.

Chow F.C., Jardine R.J., Brucy F., Nauroy J.F. 1998. Effects of time on capacity of pipe piles in dense marine sand. Journal of Geotechnical and Geoenvironmental Engineering 124(3), 254-264.

Fakharian K. and Evgin E. 1997. Cyclic simple-shear behavior of sand-steel interfaces under constant normal stiffness condition. Journal of Geotechnical and Geoenvironmental Engineering 123(12), 1096-1105.

Foray P., Tsuha C.H.C, Silva M., Jardine R.J. et Yang Z. 2010. Stress paths measured around a cyclically loaded pile in a calibration chamber. Proc. Int. Conf. on Physical Modelling in Geomechanics, ICPMG 2010, Zurich, Switzerland.

Jardine R.J. and Standing J.R. 2000. Pile load testing performed for HSE cyclic loading study at Dunkirk, France. Two Volumes. Offshore Technology Report OTO 2000 007. Health and Safety Executive, London. 60p and 200p.

Jardine R.J., Chow F., Overy R., Standing J. 2005. ICP Design Methods for Driven Piles in Sands and Clays, Thomas Telford.

Jardine R.J., Standing J.R., Chow F.C. 2006. Some observations of the effects of time on the capacity of piles driven in sand. Géotechnique 56 (4), 227-244.

Jardine R.J., Zhu B.T, Foray P., Dalton C.P. 2009. Experimental arrangements for the investigation of soil stresses developed around a displacement pile. Soil and Foundations 49(5), 661-673.

Jardine R., Zhu B., Foray P., Yang Z., 2012. Interpretation of stress measurements made around closed-ended displacement piles in sand. Géotechnique. Accepté.

Kolk H.J, Baaijens A.E, Vergobbi P., 2005. Results of axial load tests on pipe piles in very dense sands: the EURIPIDES JIP Proc. 1st Int. Symposium on Frontiers in Offshore Geotechnics, ISFOG 2005, Perth, W.A.,pp. 661- 667.

Lehane B. M., Jardine R. J., Bond A. J., Frank R., 1993. Mechanisms of shaft friction in sand from instrumented pile tests. Journal of Geotechnical Engineering 119(1), 19-35.

Le Kouby, A., Canou, J., and Dupla, C. 2004. Behaviour of model piles subjected to cyclic axial loading. Cyclic Behaviour of Soils and Liquefaction Phenomena: 159-166. Triantafyllidis (ed).

Mortara G., Mangiola, A., Ghionna, V.N. 2007. Cyclic shear stress degradation and post-cyclic behaviour from sand-steel interface direct shear tests. Canadian Geotechnical Journal 44(7), 739-752.

Tsuha C.H.C., Foray P., Jardine R.J., Yang Z.X., Silva M., Rimoy S. 2012. Behaviour of displacement piles in sand under cyclic axial loading. Soils and Foundations 52(3), 393-410.

Yang Z.X., Jardine R.J., Zhu B.T., Foray P., Tsuha C.H.C., 2010. Sand grain crushing and interface shearing during displacement pile installation in sand. Géotechnique 60(6), 469-482.

Zhu B.T., Jardine R.J., Foray P., 2009. The use of miniature soil stress measuring sensors in applications involving stress reversals. Soilsand Foundations 49(5), 675-688.

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Characterization of the geotechnical properties of a carbonate clayey silt till for ashallow wind turbine foundation

Caractérisation des propriétés géotechniques d'un silt argileux carbonaté glaciaires pour unefondation superficielle d’éolienne

Tyldesley M., Newson T.Geotechnical Research Centre, Department of Civil Engineering, University of Western Ontario, Ontario, Canada.

Boone S.Golder Associates Ltd., London, Ontario, Canada.

Carriveau R.Entelligence Research Group, Department of Civil Engineering, University of Windsor, Ontario, Canada.

ABSTRACT: Wind energy is a major source of renewable energy and is projected to capture 11% of the energy generation capacityfor Ontario by 2018. A number of problems that the energy industry currently faces stem from a lack of understanding of cyclicloading of Ontario soils and a paucity of regional regulatory guidance for site investigation and design methods for wind turbinefoundations. A multi-disciplinary research project is underway to integrate laboratory testing, field monitoring and numericalmodeling of a commercial wind turbine on a shallow foundation. This paper describes an initial part of the study to characterize thegeotechnical properties of the clayey silt till soils on the site. Emphasis has been placed on comparison of different in situ andlaboratory methods, and correlations for determining key geotechnical parameters for wind turbine foundation design.

RÉSUMÉ : L'énergie éolienne est une source importante d'énergie renouvelable et doit permettre de satisfaire 11 % de la capacité deproduction d'énergie de l'Ontario d'ici 2018. Un certain nombre de problèmes auxquels l'industrie de l'énergie est actuellementconfrontée provient d'un manque de connaissances des sols de l’Ontario sous charges cycliques et de directives réglementairesrégionales pour les méthodes d'investigation et de conception des fondations d’éoliennes. Un projet de recherche multidisciplinaire esten cours pour intégrer les tests en laboratoire, l’instrumentation et la modélisation numérique d'une éolienne commerciale surfondation superficielle. Cet article décrit la partie initiale de l'étude pour caractériser les propriétés géotechniques du silt argileuxglaciaire du site. L'accent a été mis sur la comparaison de différentes méthodes in situ et en laboratoire ainsi que les corrélations pourdéterminer les paramètres géotechniques clés pour la conception de fondation d’éoliennes.

KEYWORDS: wind turbine, clay, till, shallow foundation, soil-structure interaction, elastic, anisotropy, in situ, geophysical.

1 INTRODUCTION

1.1 Wind energy and turbine design in Canada

Wind is a major source of renewable energy and is projected tocapture 11% of the energy generation capacity for Ontario by2018 (CANWEA, 2011). However, to achieve this expansionsome major technical and policy issues must be addressed bythe Canadian wind sector. Some of these issues are associatedwith the construction and design of foundations for windturbines. Foundations for onshore wind turbines usually consistof large gravity bases and monopiles (e.g. DNV/Risø, 2002).The geometry and foundation type depends on the wind climate,power regulation philosophy, physical characteristics of themachine, uplift criteria, required foundation stiffness andgeotechnical characteristics of the site (Bonnett, 2005). Thecritical analyses for design include bearing capacity andoverturning resistance, horizontal and rotational displacements,and dynamic soil-structure interaction (Harte et al., 2012).

Although there has been much recent research associatedwith foundations for offshore wind turbines (e.g. Byrne andHoulsby, 2003), the literature on onshore systems is stillrelatively sparse. Consequently, despite similar issues for windturbine foundations across the industry, there is often diverseinterpretation of design codes and understanding of the behaviorof foundations (Morgan and Ntambakwa, 2008). This can leadto quite different foundation designs on different wind farmswith the same turbines and comparable geotechnical profiles.This issue is exacerbated in Canada, since there is currently noregional regulatory guidance for site investigation and designmethods for wind turbine foundations. Hence it is not surprising

that rather generic approaches have developed for siteinvestigation and design, which are relatively crude and canlead to quite conservative designs. To capture more windenergy, the industry is continuing to develop larger turbines andis considering more marginal sites in terms of geotechnicalcharacteristics, which will only complicate the current situation.

1.2 Project overview and objectives

A number of the above issues are being addressed as part of amulti-disciplinary research project that includes an integratedlaboratory testing, field monitoring and numerical modelingprogram investigating the behaviour of a fully operationalCanadian commercial wind turbine throughout its service life.The equipment installed on the turbine will enable an integrated,life cycle assessment of the wind turbine and its foundation.This paper describes the portion of the study that involvespreliminary characterization of the geotechnical properties ofthe wind farm site. In particular, a comparison between the insitu testing, laboratory testing and commonly used correlationsare presented. It is anticipated this process will guide futureprojects on clayey silt tills in Ontario and provide cost effectivesite investigation and design methods for turbine foundations.

2 SOIL PROFILES & MATERIAL CHARACTERISATION

2.1 Wind farm and geological environment

The wind farm is located in a simple geographical andenvironmental area in the Great Lakes region of SouthernOntario. The farm has horizontal axis 2.3 MW turbines with an

Characterization of the geotechnical properties of a carbonate clayey silt till for a shallow wind turbine foundation

Caractérisation des propriétés géotechniques d'un silt argileux carbonaté glaciaires pour une fondation superficielle d’éolienne

Tyldesley M., Newson T. Geotechnical Research Centre, Department of Civil Engineering, University of Western Ontario, Ontario, Canada

Boone S. Golder Associates Ltd., London, Ontario, Canada

Carriveau R. Entelligence Research Group, Department of Civil Engineering, University of Windsor, Ontario, Canada

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80 m hub height and triple bladed rotors with a 93 m diameter.The tower has a typical tapered tubular steel design and isfounded on a 16 m diameter hexagonal reinforced concreteshallow foundation at 3.6 m depth. The site is underlain bycarbonate-rich clayey silt tills that are a ubiquitous feature of theGreat Lakes basins and is located at the confluence of fourmajor geological deposits. These consist of the Port Stanley andTavistock tills, glaciolacustrine sand and gravel, andglaciolacustrine clayey silt. These materials were laid down inthe Port Bruce Stade (c. 14,800 years bp.) during the re-advanceof the Laurentide Ice Sheet of the Late Wisconsin. Thesesubglacial lodgement tills are calcareous and fine-grained,suggesting that the ice overrode and incorporated fine-grainedglaciolacustrine sediments deposited during the previous ErieInterstade. This has created approximately 40-45 m thickness ofclayey silt tills with interbedded glaciolacustrine sediments. Thebedrock is shale with limestone-dolostone-shale interlayers.

2.2 Overview of site investigation

The site investigation was designed to establish detailedstratigraphic and geotechnical characteristics for the soilsbeneath the wind turbine foundation. Forty metre deepboreholes were drilled on the site to evaluate the soil profile,perform in situ tests and collect high-quality samples forlaboratory testing. A track-mounted drill was used for thedrilling activities. Three boreholes were drilled 10-16 madjacent to the turbine foundation (to ensure minimal stresschange from the foundation). The wash boring method was usedfor two of the holes and the PQ coring method for the otherhole. The boreholes were drilled to depths of twice thefoundation diameter and were spaced at 3 m to allow for latercross-hole geophysical testing. Thin-wall shelby tube samplingwas completed to obtain minimally disturbed samples for thelaboratory testing. In situ testing adjacent/in to the boreholesconsisted of SPT, field shear vane, cross-hole geophysics andseismic SCPTu, and was conducted to depths of 30 m. Tocomplement the in situ test results, laboratory tests wereconducted for soil classification and geotechnical properties.

2.3 Soil description and basic properties

This deposit can be separated into three zones: a heavilyweathered oxidized upper crust from 0–1.5 m, a partiallyweathered lower crust that transitions from an oxidized to anunoxidized state from 1.5–4.5 m and an unweathered clay tillbelow 4.5 m to greater than 40 m depth. The intensity offissuring in the upper crust is very intense and the depositbecomes nearly unfissured below 4.5 m. The fissures arevertically dipping planar joints striking at right-angles. Thefissure spacing at 1.5 m depth is 15 cm and this increases to 0.6-1.2 m at 4.5 m depth. The variation in moisture contents and theAtterberg limits with depth are shown in Figure 1.

The upper crust zone of this deposit is weathered, mottledbrown-grey or brown-green with a stiff to very stiff consistency.This weathered zone generally has higher moisture contents(22-32%) due to the infiltration of surface water into the fissuresof the clay. The underlying lower crust is prevalently brown incolour and has a very stiff consistency and relatively lowernatural moisture content (16-20%). At several locations thislayer has clayey silt, sandy clay and silt seams. A soil colourchange occurs from brown to grey between 3 and 4 m below theground surface. Below the crust, the unweathered till extendsbeyond the maximum depth of sampling. This zone ischaracterized by a uniform grey appearance, a stiff to very stiffconsistency and relatively uniform moisture contents (16-24%).

Atterberg test results (Table 1) indicate that the material canbe classified as CL-ML to CL (silty clay or low plasticity clay).There is an increase in liquid limit and plasticity towards theupper crust and the clay content is also found to increase nearthe surface, leading to little change in activity (0.5).

Figure 1. Moisture contents and Atterberg limits with depth.

The liquidity index (IL) is found to range from 0.2 to 0.4 in thevirgin till, is below zero in the lower crust and ranges from 0.15to 0.25 in the upper crust. The bulk unit weights of the profileare generally uniform and range from 20.3 to 21.6 kN/m3.

Table 1. Atterberg Limits and Particle Size Distributions.

Layer LiquidLimit (%)

PlasticLimit (%)

Clay(%)

Silt(%)

Sand(%)

Upper Crust 46 21 40 45 15

Lower Crust 34 19 29 49 20

Unweathered Till 30 17 31 45 21

Semi-quantitative XRD shows that the unweathered till ispredominantly composed of quartz/feldspar (39%), carbonate(25-35%), mica/illite (16%), chlorite/kaolinite (7%) and traceminerals. In the 2 micron range the minerals are dominated byillite, calcite and chlorite. The lower crust has a similarcomposition, with more quartz/feldspar (49%), lower carbonate(22%), mica/illite (18%), chlorite/kaolinite (7%) and swellingclay (2%) and other trace minerals. In similar deposits (Quigleyand Obunbadejo, 1974) downwards leaching has removedcarbonates from the near surface and redeposited lower in thecrust. Table 2 shows the values of total carbonates, dolmite andcalcite (from the gas evolution method) in the three zones,confirming the removal of carbonates from the near surface.

Table 2. Carbonate Contents in the Soil Profile.

Layer TotalCarbonates (%)

Dolomite(%)

Calcite(%) C/D ratio

Upper Crust 0 0 0 -

Lower Crust 19.9 6.2 13.7 2.2

Unweathered Till 24.8 6.2 18.6 3.0

2.4 Compressibility and strength properties

In common with other tills around the world thecompressibility, permeability and strength characteristics of thismaterial are generally a function of the clay content. Estimatesof undrained shear strength (su) using various methods are

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shown in Figure 2. All of the profiles show that the values of suare relatively constant with depth below 7 m and are in therange of 100-130 kPa. The lower crust material (2-4.5 m)increases in strength rapidly, in excess of 250 kPa and the uppercrust material has a similar strength to the lower till. The usualhierarchy of strengths is seen for the different methods, due tothe different modes of shearing. However, the field vane (FSV)shows higher values than the triaxial compression (CIU) test.This is likely due to partial drainage and problems rotating thevane slowly enough for an undrained state. Two estimates havealso been determined from the CPT (Mayne, 2007):

su = (qt - vo)/Nkt (1)

su = u/Bq.Nkt (2)

where Nkt is a cone factor (taken as 15), u is the excess porepressure and Bq is the ratio of excess pore pressure to the netcone resistance (qt - vo). The approach based on excess porepressures appears to give better estimates for the strengths, butthe cone would be anticipated to provide lower values than CIUtriaxial, since the shearing mode is a complex combination oftriaxial compression/extension and plane strain. The depth ofthe foundation base and one base diameter (B) are also shown.

Figure 2. Undrained shear strength with depth.

From oedometer testing, average compression index (cc) for thethree layers was found to be 0.072 and average recompressionindex (cr) was 0.008, giving a ratio of 0.12, which is in the usualrange in the literature. The values of the two indices are quitelow and are typical for sandy clays/silts, and the values from thecrustal material are lower than those for the weathered till.

The pre-consolidation pressures (vp’) from oedometer testshave been estimated using the method of Boone (2011) and thecorresponding overconsolidation ratio (OCR) is shown inFigure 3. This shows low OCRs in the weathered till, with arelatively small increase in the crustal material, up to an OCR of4. Another estimate of OCR is shown using the relationship ofLadd et al. (1977), equation (3), with m = 0.8 and the ratio ofundrained shear strength (from CIU triaxial testing) to the insitu vertical effective stress [su/vo’]nc = 0.22:

m

ncvou

ocvou OCR/s/s

(3)

This shows similar values of OCR at depths below 15 m, butmuch higher OCR values for shallower depths, up to an OCR of15 at 4 m. Two further estimates of over-consolidation ratiohave been made using the CPT data with expressions for the

preconsolidation pressure (vp’), after Mayne (2007):

vp’ = 0.33.(qt - vo) (4)

vp’ = 0.161.Go0.478.’vo

0.42 (5)

where Go is the small-strain stiffness determined from theseismic cone data, takes a value of 0.85 for silts and qt is thecone tip pressure. These relationships show similarcharacteristics to the previous estimates, with the small-strainexpression closely following the oedometer derived data and theCIU triaxial derived data following the CPT expression.Interestingly, the ratio of undrained shear strength to the in situvertical effective stress in the upper crust [su/vo’] oc shows quitehigh values of 2.7-3.4, dropping to 0.3 at depth. This suggestsvalues of Ko in excess of 1 and as high as 2.4 in the crust.

Figure 3. Overconsolidation ratio with depth.

2.5 Small-strain stiffness properties

Small-strain stiffness (Go) is presumed to be a function of thevoid ratio, stress history and ratio of horizontal (h) to verticalstresses (v). It is also thought to be related to the soil macro-fabric and can often display cross-anisotropic characteristics(where the vertical axis is an axis of radial symmetry). Thecharacterization of cross-anisotropic elastic materials can bereduced to five independent elastic moduli (Eh, Ev, vh, hh andGhh; Pennington et al., 1997). In situ and laboratory estimates ofsmall-strain stiffness often use measurements of shear wavevelocity (Vs) travelling and polarized in different directions todetermine shear modulus. Hence various methods ofdetermining in situ elastic moduli provide often providedifferent components of the elastic stiffness tensor Go(ij).

Estimates of the small-strain stiffness (Go) from different insitu tests are shown in Figure 4. This includes cross-holegeophysics, seismic cone and two correlations; one usingstandard CPT output parameters (Long and Donohue, 2010) andone based on soil properties (Hardin and Black, 1969):

Vs = 1.961.qt0.579.(1+Bq)1.202 (6)

Vs = (159-53.5eo).OCR0.18/2.’vo0.25 (7)

where Vs is the shear wave velocity, eo the in situ void ratio,small-strain shear modulus Go = .Vs

2 and is density. Thevalues of Go appear to generally increase with depth and rangefrom 50 to 350 MPa, with the majority of values being between75 and 150 MPa. The cross-hole measurements were made withan axial hammer system and thus provide estimates of Gohv;these values are generally constant with depth and give the

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highest in situ estimates of shear modulus. The seismic coneprovides estimates of Govh and these values are lower than thoseof the cross-hole testing. The Long and Donohue (2010) CPTcorrelation (from equation 6) shows comparable variations in Gowith depth and falls between the two other in situ test datasets.However the Hardin and Black (1969) method based on OCRand overburden (equation 7) shows much higher estimates of Go(despite using the lower values of OCR from Figure 3).

Figure 4. Small-strain shear modulus with depth.

Similar variations in the elastic shear moduli (Govh and Gohv)have been observed with in situ tests previously (Pennington etal. 1997). These would be expected to be equal for a perfectlycross-anisotropic material. However, the different travel timesmay result from the averaging of the shear wave velocitythrough layered strata (for vertical travel), compared to lateralwave velocity through the stiffest layers. This leads to cross-hole measurements tending to measure the stiffest layers, ratherthan the average stiffness for SCPT measurements.

3 DISCUSSION

The design of gravity base foundations for onshore windturbines requires accurate estimates of strength parameters forbearing capacity and stiffness parameters for displacements,within at least 1B of the founding level. The adoption of themost appropriate methods for site investigation to determinethese parameters is debated in the industry and variouspublished correlations are commonly used. Unfortunately manyof these correlations have been previously developed forgeologically young and relatively simple materials, and theirapplicability beyond their original databases can be uncertain.

The different methods of determining the undrained shearstrengths show the crustal materials are quite strong,particularly near the founding depth, with undrained shearstrength of up to 300 kPa reducing to 100 kPa at the crust base.Given the relatively high cv (and permeability) and field vanevalues, there is a possibility that the CPT and vane estimatesmay be artefacts due to partial drainage. The crustal zone alsohas fissuring related to drying/wetting and frost action, and fieldshearbox tests on similar materials have indicated that bulkstrengths can reduce considerably, and therefore representativevalues may be closer to 60-80 kPa (Lo, 1970). However,whether crustal fissures and associated strength changes aresignificant for such large shallow foundations is questionable.

Since overconsolidation ratio is often used as a component ofcorrelations to determine geotechnical parameters, accurateestimation is important. Overconsolidation in tills is oftenattributed to loads from the overlying ice, however if drainage isinhibited, then only a small degree of consolidation will occur.

The measurement of preconsolidation pressure in tills usinglaboratory testing has been found to be quite difficult due to thehigh pressures often required to fully define compression curvesand the effects of sample disturbance (which lead to under-estimation of vp’). The difficulties with this process are evidentin the wide range of estimates for OCR shown in Figure 3.

Stiffness anisotropy is often evident in soils from in situ andlaboratory measurements. The data in Figure 4 shows thegeneral difficulties in choosing appropriate estimates of thesmall-strain stiffness (Go). Indeed cross-anisotropy in till maybe difficult to justify, since sub-glacial shear and consolidationcould have effects on the anisotropy of the in situ stress andfabric. Rocking stiffness (k) for circular surface loads (radius,R) is estimated using equation 8, (DNV/Risø, 2002):

)1(3GR8k

3

(8)

where is Poisson’s ratio and G is the shear modulusdetermined from the shear modulus ratio G/Go that corrects thestiffness for degradation due to strain level (this is typically 0.25for the presumed strain levels of 10-3 for wind turbines).Manufacturers recommend criteria for rocking stiffness toensure the natural frequency of the turbine remains above themain excitation frequencies. The range of small-strain moduli inFigure 4 indicate rocking stiffnesses from 50 to 170 GNm/rad,which is in excess of typical requirements of 40 GNm/rad, butstill represents quite a significant range of stiffness.

4 CONCLUSIONS

There is currently little guidance for choosing cost effective siteinvestigation methods and interpreting the results for this typeof geotechnical structure on glacial tills in Ontario. It isanticipated that the completion of this project will provide someof the missing knowledge and insight required in this area.

5 ACKNOWLEDGEMENTS

We wish to acknowledge the support of NSERC, GolderAssociates, Michael Cookson, JJ Davis, and Paul Dawson.

6 REFERENCES

Boone S. 2011. A critical reappraisal of preconsolidation pressureinterpretations using the oedometer test. Can Geo J. 47 (3), 281-296.

Bonnett D. 2005. Wind turbine foundations: loading, dynamics anddesign. The Structural Engineer, 83 (3), 41–45.

Byrne B. and Houlsby. G. 2003. Foundations for offshore windturbines. Phil Trans: The Royal Society, 361 (1813), 2909-2930.

CANWEA. 2011. The Economic Impacts of the Wind Energy Sector inOntario 2011–2018. ClearSky Advisors Inc. Report, pp 46.

DNV/Risø. 2002. Guidelines for Design of Wind Turbines, 2nd Edition.Hardin B.O. and Black W.L. 1969. Vibration Modulus of Normally

Consolidated Clay; Closure. J. SMF. ASCE, 95 (SM6), 1531–1537.Harte, M., Basu, B. and Nielsen R. 2012. Dynamic Analysis of Wind

Turbines Including Soil-Structure Interaction. EngineeringStructures, 45, 509-518.

Ladd C.C., Foott R., Ishihara K., Schlosser F. and Poulos H.G. 1977.Stress-deformation and strength characteristics: SOA report. Proc.,9th Int. Conf. on Soil Mech and Found Eng., Tokyo, 2, 421-494.

Lo K.Y. 1970. The operational strength of fissured clays. Geotechnique20 (1), 57-74.

Long M. and Donohue S. 2010. Characterisation of Norwegian marineclays with combined shear wave velocity and CPTU data. Can GeoJ. 47 (7), 709-718.

Mayne P. 2007. NCHRP Synthesis 368. 2007. Cone PenetrationTesting, A Synthesis of Highway Practice. National CooperativeHighway Research Program, TRB: Washington, D.C.

Pennington D. S., Nash D. F. T. and Lings M. L. 1997. Anisotropy ofG0 shear stiffness in Gault clay. Géotechnique. 47 (3), 391 - 398.

Quigley R. M. and Ogunbadejo T. A. 1974. Soil weathering, soilstructure and engineering properties. Soil Microscopy, 165-178.

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Cyclic loading of caisson supported offshore wind structures in sand

Chargement cyclique des éoliennes offshore soutenues par des caissons à succion en sable

Versteele H. Cathie Associates SA/NV, Diegem, Belgium (formerly Université de Liège, Liège, Belgium)

Stuyts B., Cathie D. Cathie Associates SA/NV, Diegem, Belgium Charlier R. Université de Liège, Liège, Belgium

ABSTRACT: With the number of offshore wind turbines in Europe growing rapidly, offshore wind farm developers are looking forsupport structures which are relatively light, easy to produce and install and are suited for water depths in excess of 30m. Suctioncaissons could offer a solution for these requirements. Since cyclic environmental loads form an important part of the loadingconditions, the cyclic degradation of the caisson capacity needs to be evaluated in detail. During storm events, pore pressure build-up inside and around the caisson can lead to degradation of capacity and stiffness. To date, there are no generally accepted materialmodels which combine generation and dissipation of pore pressure with the mechanical response of the sand. Existing methods foranalyzing pore pressure build-up are reviewed. Subsequently, a numerical model is proposed which captures the phenomena of porepressure generation and dissipation around the caisson. Pore pressure increases under storm load cycles are calculated from cycliclaboratory tests and are added to existing pore pressures in the numerical model. The influence of cyclic loading history and drainage effects on the caisson performance is assessed using the 3D FE model. Implications for suction caisson design in sand are outlined.

RÉSUMÉ : Vu la croissance rapide du nombre d'éoliennes offshore en Europe, les développeurs des parcs éoliens offshore sont intéressés par des structures combinant légèreté, facilité de fabrication et qui sont adaptées à des profondeurs d'eau supérieures à 30m.Les caissons à succion répondent à ces critères. Comme les charges environnementales cycliques constituent une partie importante du chargement total, la dégradation cyclique de la capacité portante du caisson doit être évaluée en détail. Lors de tempêtes,l'accumulation de pressions d’eau interstitielle à l’intérieur et autour du caisson peut induire une dégradation de la capacité et de la raideur. A ce jour, il n’existe pas de modèle de matériau unanimement accepté qui combine génération et dissipation de pressioninterstitielle et comportement mécanique du sable. Les méthodes existantes d'analyse de génération de pression interstitielle sont examinées dans un premier temps. Ensuite, un modèle numérique intégrant les principaux mécanismes de génération et dissipation deces surpressions autour du caisson est introduit. L'augmentation de pressions interstitielles résultants des charges cycliques dues aux tempêtes est estimée de manière indirecte sur base des résultats d'essais cycliques en laboratoire; ces surpressions sont ensuiteajoutées aux pressions interstitielles existantes dans le modèle numérique. L’influence de l’historique de chargement cyclique et des conditions de drainage est évaluée à l’aide du modèle éléments finis 3D. Enfin, les implications de ces résultats pour la conception de caissons à succion sont exposées.

KEYWORDS: suction caisson, cyclic loading, liquefaction analysis, offshore wind turbine, marine geotechnics

1 INTRODUCTION

1.1 Suction caisson as foundations for offshore wind turbines

The European Wind Energy Association expects that the installed offshore wind capacity within the EU will increase from 4GW to 40 GW by 2020 (EWEA 2011) requiring the installation of approximately 6000 6MW turbines located ever further offshore in consequently deeper waters. Due the demanding working conditions at sea and the limited availability of offshore installation vessels, the foundation system typically accounts for up to 25-30 % of the total cost of an offshore wind farm. This makes the choice and design of the foundation an important factor in the overall cost effectiveness of offshore wind farms.

Offshore wind farm developers are thus looking for support structures which are relatively light, easy to produce and install and are suited for water depths in excess of 30m. Suction caissons could offer a solution for these requirements.

A suction caisson is a steel structure consisting of a circular top plate with peripheral vertical skirts (Figure 1). In operation it is similar to a skirted gravity foundation, but the skirt length is significant compared to the diameter.

Installation of the caisson is achieved in two phases. After initial penetration under the self-weight of the caisson, water is pumped out. The induced pressure difference pushes the caisson

into the soil, while the induced seepage forces and reduced effective stress near the skirt tips facilitate penetration.

Advantages of the caisson include a potentially lower cost than equivalent piled foundations (Senders 2008) and relatively easy installation and removal, not restricted by water-depth.

Diameter

Skirt

leng

th

Figure 1: Cross-section sketch of a suction caisson and installation principle


1

Cyclic loading of caisson supported offshore wind structures in sand

Chargement cyclique des éoliennes offshore soutenues par des caissons à succion en sable

H. Versteele Cathie Associates SA/NV, Diegem, Belgium (formerly Université de Liège, Liège, Belgium)

B. Stuyts & D. Cathie Cathie Associates SA/NV, Diegem, Belgium

R. CharlierUniversité de Liège, Liège, Belgium

ABSTRACT: With the number of offshore wind turbines in Europe growing rapidly, offshore wind farm developers are looking for support structures which are relatively light, easy to produce and install and are suited for water depths in excess of 30m. Suction caissons could offer a solution for these requirements. Since cyclic environmental loads form an important part of the loading conditions, the cyclic degradation of the caisson capacity needs to be evaluated in detail. During storm events, pore pressure build-up inside and around the caisson can lead to degradation of capacity and stiffness. To date, there are no generally accepted material models which combine generation and dissipation of pore pressure with the mechanical response of the sand. Existing methods for analyzing pore pressure build-up are reviewed. Subsequently, a numerical model is proposed which captures the phenomena of pore pressure generation and dissipation around the caisson. Pore pressure increases under storm load cycles are calculated from cyclic laboratory tests and are added to existing pore pressures in the numerical model. The influence of cyclic loading history and drainage effects on the caisson performance is assessed using the 3D FE model. Implications for suction caisson design in sand are outlined.

RÉSUMÉ : Vu la croissance rapide du nombre d'éoliennes offshore en Europe, les développeurs des parcs éoliens offshore sont intéressés par des structures combinant légèreté, facilité de fabrication et qui sont adaptées à des profondeurs d'eau supérieures à 30m. Les caissons à succion répondent à ces critères. Comme les charges environnementales cycliques constituent une partie importante du chargement total, la dégradation cyclique de la capacité portante du caisson doit être évaluée en détail. Lors de tempêtes, l'accumulation de pressions d’eau interstitielle à l’intérieur et autour du caisson peut induire une dégradation de la capacité et de la raideur. A ce jour, il n’existe pas de modèle de matériau unanimement accepté qui combine génération et dissipation de pression interstitielle et comportement mécanique du sable. Les méthodes existantes d'analyse de génération de pression interstitielle sont examinées dans un premier temps. Ensuite, un modèle numérique intégrant les principaux mécanismes de génération et dissipation de ces surpressions autour du caisson est introduit. L'augmentation de pressions interstitielles résultants des charges cycliques dues aux tempêtes est estimée de manière indirecte sur base des résultats d'essais cycliques en laboratoire; ces surpressions sont ensuite ajoutées aux pressions interstitielles existantes dans le modèle numérique. L’influence de l’historique de chargement cyclique et des conditions de drainage est évaluée à l’aide du modèle éléments finis 3D. Enfin, les implications de ces résultats pour la conception de caissons à succion sont exposées.

KEYWORDS: suction caisson, cyclic loading, liquefaction analysis, offshore wind turbine, marine geotechnics

1 INTRODUCTION

1.1 Suction caisson as foundations for offshore wind turbines

The European Wind Energy Association expects that the installed offshore wind capacity within the EU will increase from 4GW to 40 GW by 2020 (EWEA 2011) requiring the installation of approximately 6000 6MW turbines located ever further offshore in consequently deeper waters. Due the demanding working conditions at sea and the limited availability of offshore installation vessels, the foundation system typically accounts for up to 25-30 % of the total cost of an offshore wind farm. This makes the choice and design of the foundation an important factor in the overall cost effectiveness of offshore wind farms.

Offshore wind farm developers are thus looking for support structures which are relatively light, easy to produce and install and are suited for water depths in excess of 30m. Suction caissons could offer a solution for these requirements.

A suction caisson is a steel structure consisting of a circular top plate with peripheral vertical skirts (Figure 1). In operation it is similar to a skirted gravity foundation, but the skirt length is significant compared to the diameter.

Installation of the caisson is achieved in two phases. After initial penetration under the self-weight of the caisson, water is

pumped out. The induced pressure difference pushes the caisson into the soil, while the induced seepage forces and reduced effective stress near the skirt tips facilitate penetration.

Advantages of the caisson include a potentially lower cost than equivalent piled foundations (Senders 2008) and relatively easy installation and removal, not restricted by water-depth.

Figure 1: Cross-section sketch of a suction caisson and installation principle

Diameter

Skir

t len

gth

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2

1.2 Loads on offshore wind turbines

The foundation must resist loads caused by the weight of the structure, the operation of the turbine, currents, wind and wave action. Incoming waves exert a cyclic horizontal force (and moment) on the foundation, which in the case of offshore wind turbines may be a significant proportion of the weight of the structure. A vertical weight of 6MN and a horizontal wave loading of up to 3MN are realistic values for a 3.5MW turbine (Houlsby et al. 2005).

The offshore design standard DNV-OS-J101 (DNV 2011) specifies that the structure must be able to resist a 50-year design storm (a storm with a probability of occurrence of 1/50 during one year), where not only the peak loads, but the entire history of cyclic loading affects the stability of the structure. For the cyclic loading assessment, the irregular wave loading is usually converted into an idealized, equivalent design storm.

1.3 Structural configuration

Caissons could support offshore wind turbines in two ways, based on mode of load transfer to the soil (Figure 2). A monopod foundation consists of a single caisson and is suited for shallow waters. In deeper water, the increased moments acting on the caisson would require a very large caisson. In that case a tripod (three caissons) or quadripod (four caissons) structure could be economical, as moment loads are converted into a vertical push and pull action on the individual caissons.

Figure 2: The monopod and multipod concept and reaction forces on the caissons

1.4 Scope of work

The aim of this paper is to examine the effect of cyclic loading during a design storm on both the monopod and multipod and to produce a model which is suitable for engineering practice. The presented model is still under development, and is considered a starting point for more sophisticated approaches.

2 CYCLIC DEGRADATION OF SOILS AND FOUNDATIONS

2.1 Pore pressure build up in sand under cyclic loading

Cyclic shearing of sand degrades the soil structure and causes a tendency to densify. This is the case even for very dense sands that are dilative under monotonic loading conditions (Seed and Idriss 1980, Andersen and Berre 1999).

Under undrained conditions, volume changes are prevented by the low compressibility of water, so normal stresses carried by the soil will be transferred to the pore water, thus increasing the pore water pressure in the sample as illustrated in Figure 3. The decrease in effective stress furthermore causes a progressive increase in average shear strain. Failure occurs when the generated pore pressure reaches a critical value umax.

Figure 3: Behaviour of sand under cyclic loading (after Andersen and Berre 1999)

The intensity of cyclic loading is expressed in terms of the cyclic shear stress ratio, the ratio of cyclic deviatoric stress amplitude over mean effective stress. This formulation is convenient for the interpretation of triaxial test results and for implementation in the finite element procedure.

⁄ (1)

Based on several cyclic tests at different CSR, cyclic shear strength curves can be established, expressing the number of cycles required to induce failure Nl as a function of the CSR and Dr.

The cyclic shear strength depends on the relative density and the initial shear stress in the sample. The set of curves used in this study was presented by Lee and Focht (1975) in their investigation of the liquefaction potential at the Ekofisk site, North Sea. The curves for this typical dense North Sea sand are redrawn in Figure 4.

Figure 4: Cyclic shear strength curves for dense North Sea sand at the Ekofisk site (after Lee and Focht 1975)

The build-up of pore pressure in samples can be described by the empirically determined pore pressure generation function given in Eq. 2 and plotted in Figure 5. The empirical constant αdepends on the soil properties and is on average equal to 0.7 (Rahman et al. 1977). As it is cyclically loaded, the soil sample evolves from the initial, undisturbed state at N = 0 to a state of liquefaction at N = Nl and u = umax.

⁄ (2)

2.2 Drainage conditions

In laboratory tests soil samples are brought to failure under undrained conditions. However, in situ loading conditions may be fully or partially drained, depending on the combination of soil permeability, frequency of the loading and drainage conditions.

For offshore turbines founded on sand, the high permeability and relatively slow wave loading results in the dissipation or redistribution of a significant part of the generated pore pressure

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during the cyclic loading itself. This effect becomes more important as the soil permeability increases and the loading frequency diminishes. Not taking into account the simultaneous dissipation leads to overestimation of the generated pore pressure and potentially to overconservative design.

Figure 5: Pore pressure generation function.

2.3 Liquefaction of foundations

The definition of failure of a foundation due to liquefaction requires special attention. Not all parts of the soil under a foundation will fail at the same time or will fail at all. Some intensely loaded zones may liquefy completely or partially, while other zones may still be intact.

Taiebat (1999) discussed the problem and proposed the following definitions. Total failure of a foundation-soil system under cyclic loading is defined as the condition where the soil mass deforms continuously under the ambient and cyclic loads applied to the foundation, resulting in bearing capacity failure. Partial failure involves large permanent displacements during cyclic loading. Some elements of the soil liquefy and lose their strength, but overall, the soil mass remains stable.

Due to the complexity of the problem, numerical analysis is often the preferred method to asses to what extent the foundation capacity is degraded.

3 EXISTING NUMERICAL METHODS

There are at least two approaches to numerical modelling of offshore foundation liquefaction. In the first approach an appropriate constitutive model is used to capture cyclic stress-strain behaviour of the soil. Many such models exist and they can successfully reproduce soil behaviour in laboratory conditions (e.g. bounding surface plasticity, multi-surface plasticity). However, the number of required parameters and calculation time are two obstacles that up to now have limited application of these models to analysis of boundary value problems in engineering practice.

The second approach is simpler and consists of improving a conventional (possibly slightly modified) constitutive model by incorporating the effects of cyclic loading separately, based on a set of laboratory tests. A rigorous review of the work by researchers who followed this approach to analyze offshore foundations subjected to wave loading is given by Taiebat (1999).

4 IMPLEMENTED METHOD

The proposed method follows the second approach and is based on the work by Rahman et al. (1977), Taiebat (1999) and to a lesser extent Lee & Focht (1975) and Verruijt & Song (1991).

The calculation procedure is as follows: undrained pore pressure increases are calculated analytically, at regular time

intervals in the FE analysis. At each node, the pore pressure at the end of the previous interval (which includes effects of all previous loading) is converted into an equivalent number of cycles using Eq. 2. The increase in pore pressure during next interval (containing a number of load cycles) can then be calculated from Eq. 2, assuming the CSR is constant during this interval.

After the pore pressure and effective stress in the FE analysis are updated accordingly, the dissipation analysis continues over the length of the considered time interval. This is done in a coupled Biot-type consolidation analysis in the FE package Abaqus.

The total design storm consists of a number of load parcels, during which the cyclic load (and thus the CSR) is assumed to have a constant average and amplitude. The load parcels are subdivided in a number of steps and the process of updating the pore pressure and subsequent consolidation is repeated for every subdivision, tracing the average pore pressure response (excluding oscillations within each load cycle) over the entire load history of the design storm.

5 APPLICATION TO SUCTION CAISSONS

In two case studies the influence of cyclic loading history and drainage effects on the caisson performance is assessed using the proposed model. Realistic forces acting on the foundation are estimated from the loads outlined in section 1.2 and a simplified load histogram is adopted. Corresponding realistic caisson dimensions are found by applying the bearing capacity equation (DNV 1992) for the tripod caisson and the formula proposed by Byrne and Houlsby (2003) for the monopod caisson. In both cases the sand is represented by an isotropic elastic material model with Mohr-Coulomb plasticity.

5.1 Leeward caisson of a tripod

5.1.1 Model Initially the horizontal load, divided over three caissons, is neglected. The resulting axisymmetric problem only considers vertical cyclic loading on the individual caisson due to weight of the structure and overturning moments as this is the most important load component. The histogram consists of 3 load parcels of 2000 seconds each, applying 200 load cycles at 60% of the maximum load in the first and last parcel and 200 cycles at maximum loading in the middle parcel.

5.1.2 Results An example of calculated pore pressure response within and around a 8x8m caisson is shown in Figure 6. First of all it is clear that the abrupt increases (generation) and gradual decreases (dissipation) are an approximation for the real behaviour.

Figure 6: Example of excess pore pressure history, tripod caisson

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The analysis predicts liquefaction of the soil near the skirt tips and build-up of pore pressure inside the caisson during the second load parcel. Stress redistribution towards the baseplate will cause an additional increase in pore pressure inside the caisson. The abrupt increase at t = 2000s is due to the nonlinear dependency of generated pore pressure on the CSR, which increases at the start of the second load parcel. Much of the pore pressure is dissipated by the end of the last load parcel, even though cyclic loading continues (at 60% of the second parcel). As the pore pressure dissipates, settlements due to the cyclic loading are expected.

The discretization of cyclic loading in load parcels and subsequently in subdivisions affects the accuracy of the analysis, but the results seem to converge as the number of subdivisions is increased. Where short drainage paths or high CSR values are involved, sufficiently short steps are required. The rate of pore pressure dissipation is affected by the length of the skirts. Longer skirts result in slower dissipation and higher potential for pore pressure accumulation inside the caisson.

5.2 Monopod

5.2.1 Model The monopod caisson (20x10m) is subjected to three degree of freedom loading, including a horizontal and moment load. A 3D FE model of half the caisson is sufficient, taking advantage of the plane of symmetry formed by the vertical and the direction of aligned wind and wave loading. A six hour design storm, consisting of 2160 waves in five load parcels, was adopted.

5.2.2 Results The five load parcels are distinguishable in the pore pressure response plotted in Figure 7 and peak pore pressure occurs right after the peak of the storm. The permanent horizontal load due to wind and/or current causes an asymmetric cyclic shearing in the example, so the observed peak pore pressure (4 kPa) does not occur on the center line. The consequences, such as potential differential settlements and tilting of the turbine, should be examined in a more advanced analysis.

Figure 7: Example of excess pore pressure history, monopod caisson

6 CONCLUSIONS AND FURTHER DEVELOPMENTS

A pore-pressure generation and dissipation model has been developed to study the effect of cyclic loading on suction caissons in sand. Example analyses have shown that the proposed model can be successfully applied to the study of suction caissons, both in 2D and in 3D. However, the model needs further improvement to allow prediction of the complete liquefaction behaviour, including settlements, of a caisson.

The model can be used to predict which areas are prone to pore pressure build-up, estimate the rate of pore pressure build-up and to some extent how fast this pore pressure is dissipated.

Analysis of the type presented here may be useful to assess the geotechnical and structural risks related to cyclic loading of caissons in sand such as:

• reduction in caisson bearing capacity due to generated pore pressures;

• caisson foundation stiffness reductions; • pore pressure induced total and differential

settlements for offshore wind turbine structures; • analysis of the effect of scour on pore pressure

gradients. The model can be improved to reflect more realistic soil behaviour. As some zones underneath the suction caisson liquefy, the load is transferred to other parts of the foundation. This leads to secondary pore pressure increases which are not yet considered in the presented model.

If sufficient soil data are available, the cyclic shear strength curves could include dependency on the relative density and initial shear stresses in the soil.

Finally, a large part of the vertical load on suction caissons is taken by friction between the caisson skirts and the soil. A systematic study of the influence on the liquefaction potential would be interesting.

7 ACKNOWLEDGEMENTS

The work described in this paper was performed as a part of the author’s master thesis (Versteele 2012), supervised by professor Charlier (Université de Liège), whose guidance is gratefully acknowledged. Development of the model and calculations were performed at, and with support of Cathie Associates SA/NV.

8 REFERENCES

Andersen K.H. and Berre T. 1999. Behaviour of a dense sand under monotonic and cyclic loading. Proceedings of the 12th ECSMGE, Vol 2, Geotechnical Engineering for Transportation Infrastructure, 667-676

Byrne B.W. and Houlsby G.T. 2003. Foundations for offshore wind turbines. Phil. Trans. R. Soc. Lond., Vol 361, 2909-2930

DNV 1992. Classification notes No 30.4 – foundations. Det Norske Veritas, Norway

DNV 2011. Design of offshore wind turbine structures, Offshore Standard DNV-OS-J101, Det Norske Veritas, Norway

EWEA 2011. Wind in our sails – The coming of Europe’s offshore wind energy industry. http://www.ewea.org

Houlsby G.T., Ibsen L.B. and Byrne B.W. 2005. Suction caissons for wind turbines. Proc. International Symposium on Frontiers in Offshore Geotechnics (ISFOG), Taylor & Francis Group, Perth, Australia

Lee K.L. and Focht J.A. 1975. Liquefaction potential at Ekofisk tank in North Sea. Journal of Geotechnical Engineering Division, ASCE,101(GT1), 1-18

Rahman M.S, Seed H.B. and Booker J.R. 1977. Pore pressure development under offshore gravity structures. Journal of Geotechnical Engineering Division, ASCE, 103(GT12), 1419-1436

Seed H.B. and Idriss I.M. 1980. On the importance of dissipation effects in evaluating pore pressure changes due to cyclic loading. International Symposium on Soils under Cyclic and Transient Loading, Swansea, 569-570

Senders M. 2008. Suction caissons in sand as tripod foundations for offshore wind turbines. Ph.D, The University of Western Australia, Australia

Taiebat H.A. 1999. Three dimensional liquefaction analysis of offshore foundations. Ph.D. Thesis, The University of Sydney, Australia

Verruijt A. and Song E.X. 1991. Finite element analysis of pore pressure build-up due to cyclic loading. Deformation of soils and displacement of structures, Proc. 10th European Conference on Soil Mechanics and Foundation Engineering, 277-280

Versteele H. 2012. Cyclic loading of suction caisson foundations for offshore wind turbines. M.Sc. Thesis, Université de Liège, Belgium

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