Journal of Constructional Steel Research 88 (2013) 206–219

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Journal of Constructional Steel Research

Experimental study on seismic behavior of high-strength concrete filleddouble-steel-plate composite walls

Jian-Guo Nie a, Hong-Song Hu a,⁎, Jian-Sheng Fan a, Mu-Xuan Tao a, Sheng-Yong Li b, Fu-Jun Liu b

a Key Laboratory of Civil Engineering Safety and Durability of China Education Ministry, Dept. of Civil Engineering, Tsinghua University, Beijing 100084, Chinab RBS Architecture Engineer Design Associates, Guangzhou 510170, China

⁎ Corresponding author. Tel.: +86 10 13426244289.E-mail address: hhsong05@gmail.com (H.-S. Hu).

0143-974X/$ – see front matter © 2013 Elsevier Ltd. Alhttp://dx.doi.org/10.1016/j.jcsr.2013.05.001

a b s t r a c t

a r t i c l e i n f o

Article history:Received 17 August 2012Accepted 3 May 2013Available online 15 June 2013

Keywords:Concrete filled double-steel-plate compositewallsHigh-strength concreteExperimental studySeismic behaviorSuper high-rise buildings

In order to improve the ductility of the core wall in super high-rise buildings subjected to high axial compres-sive force and seismic effect, a new detailed concrete filled double-steel-plate (CFDSP) composite wall usinghigh-strength concrete is proposed. This CFDSP composite wall is composed of concrete filled steel tubularcolumns at the two boundaries and concrete filled double-steel-plate wall body which is divided into severalcompartments by vertical stiffeners transversely connected by distributed batten plates. In order to inten-sively investigate the structural mechanism of this new type of CFDSP composite walls, twelve specimensare tested under large axial compressive force and reversed cyclic lateral load. No evident buckling of surfacesteel plates can be observed due to reasonable width-to-thickness ratios of steel plates and properly arrangedbatten plates, so that the surface steel plates and infill high-strength concrete can work compatibly in thewhole loading process. All the specimens exhibited good energy dissipation ability and deformation capacitywith full hysteretic curves and large ultimate drift ratios, thereby indicating that high-strength concrete canbe used in seismic-resistant structures when the proposed new detailed walls are adopted. Based on the testresults, the stiffness and strength degradations are analyzed, and the deformation characteristics of all thespecimens are discussed in detail. Finally, a strength prediction approach based on the section analysis meth-od is presented, and some detailing requirements for routine design practice are recommended.

© 2013 Elsevier Ltd. All rights reserved.

1. Introduction

In recent years, several super high-rise buildings have been built inChina. The two tallest buildings in China—PINGAN IFC and ShanghaiTower both exceed 600 m. In a modern super high-rise building, aframe (or mega frame)-core tube hybrid structural system is usuallyused for its superior seismic behavior, large lateral stiffness, low costand rapid construction, etc. Since the core tube sustains the majorityof the seismic action and plays a significant role in the energydissipation, the structural wall of the core tube is one of themost criticalelements for the seismic design of the whole structural system.

To ensure adequate deformability of the core walls under lateralloads, the axial load ratio should not exceed an upper limit value.With the increase of building height, the axial compressive force atthe base of core wall grows quickly. As a result, for the traditionalRC structural walls, the only effective way to limit the axial loadratio is to increase the core wall thickness when the maximum avail-able concrete strength is restricted for the consideration of ductility.The excessively thick wall would increase the difficulty of construc-tion and greatly reduce the usable floor areas. In addition, the crosssections of the frame beams and columns should also be designed

l rights reserved.

very large to ensure that the frames can resist a certain part of lateralforces so that a dual system can be realized. The increase of wallthickness may also result in the increase of the structure self-weight, and correspondingly the increase of seismic effect. Therefore,it is hard to obtain an economical and rational design, or even impos-sible to accomplish the design, when the building reaches a certainheight and the traditional RC structural wall is adopted.

One effective way to solve this problem is to employ new forms ofstructural walls that have high bearing and deformation capacitiesunder large axial compressive force and cyclic lateral load with ac-ceptable wall thickness. Steel plate–concrete composite walls are re-cently developed structural walls and have been used in the actualpractice [1–3]. These steel plate–concrete composite walls can beclassified as steel-plate reinforced concrete (SPRC) composite wallsand concrete filled double-steel-plate (CFDSP) composite wallsaccording to different relative positions of steel plates and concreteas shown in Fig. 1.

Several experimental programs have been conducted on the seis-mic behavior of SPRC walls. Lv et al. [4] tested 16 SPRC walls with dif-ferent parameters and 3 RC walls for comparison. The measuredultimate top displacements of the SPRC walls were much largerthan those of the control RC walls. In the tests by Chen et al. [5],SPRC walls also exhibited much greater deformation and energydissipation capacities than RC walls. The compressive strengths of

Concrete reinforcing steel plateMesh reinforcement

Concrete Surface steel platesMesh reinforcement

a) SPRC composite wall

b) CFDSP composite wall

Fig. 1. General details for SPRC and CFDSP composite walls.

Boundary element

Concrete infill

StiffenerBatten plate

Stud

Surface steel plates

Fig. 3. A new detailed CFDSP wall.

207J.-G. Nie et al. / Journal of Constructional Steel Research 88 (2013) 206–219

concrete used in these two experimental programs ranged from16 N/mm2 to 36 N/mm2, and tests on steel-plate reinforced high-strength concrete walls are still not available in the literature.

Studies on CFDSP walls have been carried out by several re-searchers in the past years, and several different configurations havebeen proposed. Wright et al. [6] first conceived a composite wallformed from two skins of profiled steel sheeting filled with concreteas shown in Fig. 2a, which was to be applied in common commercial

a) Proposed by Wright et al. [6]

Concrete infill

Profiled steelsheetingSteel

frame

Bulkheads

Verticdiaphrag

C

Concreteinfill

Vertical diaphragm

Shear bar

Stud

Embedded steelPenetrated sleeve

Surface steel plates

b) Tested

d) Developed in Japan for nuclear power plants [14

Fig. 2. Previous develo

buildings. An integrated research program was carried out to investi-gate the performance of this kind of composite walls under differentloading conditions and finally to establish design formulas for actual

Surface steel platesalm

oncrete infill

Concreteinfill

Verticaldiaphragm

Horizontaldiaphragm

Surface steelplates

Concreteinfill

Transverse frictionwelded bars

Surface steel plates

by Link et al. [12] c) Proposed by Emori et al. [13]

] e) Bi-Steel [19]

ped CFDSP walls.

Table 1Properties of specimens.

Specimen Cross section(mm × mm)

Shearspanratio

Cubic (150 mm × 150 mm × 150 mm)compressive strength of concrete (N/mm2)

Meshreinforcement

Steel plate thickness ofboundary columns (mm)

Steel plate thicknessof wall body (mm)

Steelcontentratio

Axialcompressiveforce (kN)

CFSCW-1 1284 × 214 2.0 87.5 0 5 5 7.1% 7375CFSCW-2 1284 × 214 2.0 86.1 0 5 5 7.1% 7319CFSCW-3 1284 × 214 2.0 86.1 0 5 5 7.1% 7319CFSCW-4 1284 × 214 2.0 89.8 0 4 4 5.8% 7535CFSCW-5 1284 × 214 2.0 88.1 0 3 3 4.6% 7404CFSCW-6 1284 × 214 2.0 65.0 0 5 5 7.1% 5863CFSCW-7 1284 × 214 2.0 102.6 0 5 5 7.1% 7900CFSCW-8 1284 × 214 2.0 88.4 0 6 4 7.1% 7807CFSCW-9 1284 × 214 2.0 83.3 ∅ 8@130 mm 5 5 7.1% 7375CFSCW-10 750 × 125 2.0 83.7 0 3 3 7.1% 2756CFSCW-11 750 × 125 1.5 80.7 0 3 3 7.1% 2718CFSCW-12 750 × 125 1.0 88.0 0 3 3 7.1% 2816

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practice [7–11]. Link et al. [12] studied the static behavior of a CFDSPwall with vertical diaphragms as shown in Fig. 2b, which wassubjected to transverse and longitudinal forces for the considerationof the offshore loading conditions. Emori [13] developed a similarcomposite wall with both horizontal and vertical diaphragms asshown in Fig. 2c. In the mid-1990s, a CFDSP wall for nuclear powerplants as shown in Fig. 2d was proposed in Japan to promote rational-ization of field construction techniques [14], and several experimen-tal and theoretical studies were carried out by some researchers[15–18]. The Corus Company developed a CFDSP construction called

a) Front view

2568

1284

214 214 214 214214 2141284

214

30 60

140

2433

145×

16=2

320

6845214

25

1 1

2

2

c) 2-2 Section

End plate

Wall

Load beam

b) 1-1 Section

Fig. 4. Details of CFSCW-1.

Bi-Steel™ [19] with two surface steel plates connected by an arrayof transverse friction welded shear connectors (Fig. 2e), and a seriesof researches have been conducted on Bi-Steel [20–22]. In addition,Eom et al. [23] tested 5 isolated and 2 coupled CFDSP walls with rect-angular and T-shaped cross sections under in-plane cyclic loading.The two surface steel plates of their test specimens were connectedby tie bars, which was similar to Bi-Steel. In their tests, the axial com-pressive load was not considered, and the wall specimens failedmainly due to the tensile fracture of the welded joints at the wallbase and coupling beams, or the local buckling of the steel plates. Inspite of extensive studies of CFDSP walls in the literature, the applica-tion of CFDSP walls in super high-rise buildings using high-strengthconcrete has not been fully addressed.

High strength materials are preferred in the design of superhigh-rise buildings to meet the high strength demand of the structur-al elements. However, for the brittle post-peak behavior of high-strength concrete, the required compressive strength of concrete in

750

1500

125 10 20

99

125 125 125 125 125 125750

15

125

85×

15=

1275

1365

3060

1 1

2

2

End plate

Wall

Load beam

c) 2-2 Section

b) 1-1 Section

a) Front view

Fig. 5. Details of CFSCW-10.

Table 2Measured material properties.

Type Component Elastic modulus(N/mm2)

Yield stress(N/mm2)

Ultimate stress(N/mm2)

Steel plate 3 mm 190,047 442.8 552.84 mm 186,452 351.4 517.45 mm 205,438 305.6 444.86 mm 212,921 363.0 512.310 mm 210,333 431.7 588.0

Reinforcingbar

HPB235(∅8)

195,838 327.4 484.1

-1000

-500

0

500

1000

Later force (kN)

Load step0

1/200

Drift ratio

1/100

3/200

1/50

-1/200

-1/100

-3/200

-1/50

Fig. 7. Loading procedure for CFSCW-1–9.

209J.-G. Nie et al. / Journal of Constructional Steel Research 88 (2013) 206–219

super high-rise buildings is usually below 50 N/mm2. Enlightened bythe ductility improvement of concrete confined by the steel tube, anew detailed CFDSP wall applicable in super high-rise buildings isconceived as shown in Fig. 3. This CFDSP wall is composed of concretefilled steel tubular (CFST) columns at the two boundaries and con-crete filled double-steel-plate wall body which is divided into severalcompartments by vertical stiffeners transversely connected by dis-tributed batten plates. The high-strength concrete in each compart-ment of the wall body is confined by surrounding steel plates toenhance the concrete ductility, and the steel consumption can be

2

57 6

1

4

3

8

9

10

2North South

1—Girder of load frame 2—Column of load frame 3—Specimen 4—Hydraulic jack 5—Load head 6—Extending tube 7—Actuator

8—Steel pedestal 9—Lead screw 10—Handle jack

a) Schematic diagram

b) Photograph

Fig. 6. Test setup.

greatly saved compared to some previous CFDSP walls using verticaldiaphragms.

In order to intensively investigate the seismic behavior of this newdetailed high-strength concrete filled double-steel-plate compositewall, twelve wall specimens were tested under high axial compres-sive forces and reversed cyclic lateral loads. Failure mechanism,hysteric behavior, strength and deformation capacities, etc. arediscussed in detail, and the design recommendations for the strengthprediction and construction requirements are also proposed.

2. Experimental program

2.1. Test specimens

Twelve specimens with different parameters were designed andtested. The primary parameters were the steel content ratio, concretestrength, steel plate thickness of boundary columns and wall body,mesh reinforcement and shear span ratio. As presented in Table 1,CFSCW-1–3 were three fundamental specimens and had the sameparameters. They were 1/7 scale models of the prototype structure,which yielded the cross-sectional size of 1284 mm × 214 mm.CFSCW-4–9 were designed based upon the three fundamentalspecimens. Thinner steel plates were used for CFSCW-4 andCFSCW-5, resulting in smaller steel content ratios. Compared to thefundamental specimens, the cubic (150 mm × 150 mm) compressivestrengths of concrete for CFSCW-6 and CFSCW-7 were 20 N/mm2

lower and higher, respectively. For CFSCW-8, steel plates with differ-ent thicknesses were used for the boundary columns and wall body,but the steel content ratio was unchanged. Unlike other specimens,mesh reinforcement (∅ 8@130 mm) was placed in the wall body ofCFSCW-9. CFSCW-10–12 were designed to have different shear spanratios, which were 2.0, 1.5 and 1.0, respectively. Because of thelimitation of the loading capacity of the test setup, CFSCW-10–12were 1/12 scale models of the prototype structure, which yieldedthe cross-sectional size of 750 mm × 125 mm.

Details of CFSCW-1–9 are given in Fig. 4. The specimens hadsquare CFST columns at the boundaries, and the wall body was divid-ed into 4 compartments by the vertical stiffeners transverselyconnected by batten plates. The layout of the batten plates is shownin Fig. 4(c). Shear studs (∅ 8 × 75 mm) were welded on the innersides of surface steel plates, with horizontal spacing of 107 mm andvertical spacing of 100 mm.

Details of CFSCW-10 are given in Fig. 5, which were similar tothose of CFSCW-11 and CFSCW-12 except the wall height. Shearstuds (∅ 6 × 45 mm) were welded on the inner sides of surfacesteel plates, with horizontal spacing of 62.5 mm and vertical spacingof 100 mm.

The steel plates of specimens were welded to a 50 mm thick endplate, and the end plate was bolted to the steel pedestal.

300

500

500

500

500

268

220

20 145 263 428 2014526312

0012

00

1

2

1

1—Displacement meter 2—LVDT

500

500

500

400

150

200200 200 200 200 200 4242

3—Steel strain gauge4—Steel strain gauge rosette

3

4

a) Displacement meters and LVDTs b) Strain guage arrangement

Fig. 8. Instrumentation for CFSCW-1–9.

210 J.-G. Nie et al. / Journal of Constructional Steel Research 88 (2013) 206–219

The cubic compressive strengths of concrete for each specimen aregiven in Table 1. The measured properties of steel plates andreinforcing bars are given in Table 2.

2.2. Test setup and loading procedure

The specimens were tested by the 20,000 kN multi-functionalloading device in Tsinghua University. The vertical loading capacityof the device was 20,000 kN, and the horizontal was 3500 kN. Thetest setup is illustrated in Fig. 6.

Fracture

Run

Vertical fracture

a) b)

e)d)

Fig. 9. Failure modes of

The axial force was first applied to the specimen by a vertical hy-draulic jack. The axial compressive force N was calculated fromEq. (1) when nd was selected as 0.5. The applied axial compressiveforce for each specimen is given in Table 1.

nd ¼ 1:25Nf0cAc=1:4þ f yAs=1:11

ð1Þ

where nd is the axial load ratio for design; fc' is the compressive strengthof concrete, which can be calculated from the cubic compressive

Local buckling through

c)

f)

specimen CFSCW-3.

a) CFSCW-5 b) CFSCW-8

e) CFSCW-12

d) CFSCW-11c) CFSCW-10

Fig. 10. Final failure modes of typical specimens.

211J.-G. Nie et al. / Journal of Constructional Steel Research 88 (2013) 206–219

strength using Eq. (2) [24]; fy is the yield strength of steel plates; Ac andAs are the cross-sectional areas of concrete and steel plates, respective-ly; 1.4 and 1.11 are partial safety factors for the concrete and steel ma-terials, respectively.

f0

c ¼ 0:66þ 0:002f 150cu

� �f 150cu ≥0:76f 150cu ð2Þ

where fcu150 is the cubic compressive strength tested from the concrete

coupons with the size of 150 mm × 150 mm × 150 mm.The axial load ratio is usually defined as Eqs. (3) and (4) for RC

structures and steel reinforced concrete (SRC) structures, respective-ly. The calculated values of n1 and n2 using the actually applied axialcompressive forces given in Table 1 range from 0.32 to 0.44 and0.28 to 0.31, respectively, which are much higher than the previoustested RC structural walls [25].

n1 ¼ N

f0cAg

ð3Þ

n2 ¼ N

f0cAc þ f yAs

ð4Þ

where n1 and n2 are the axial load ratios for RC and SRC structures,respectively; Ag is the cross-sectional area of the specimen.

The lateral force was first controlled by load and then by top dis-placement. The loading procedure for CFSCW-1–9 is given in Fig. 7.In the load control stage, lateral load was applied in three levels andthe corresponding lateral forces were 500 kN, 750 kN and 1000 kN,respectively. When the drift ratio reached 1/400, it was changed to

displacement control. Two cycles were repeated for every drift level,and the displacement increment was 6.42 mm, corresponding to thedrift ratio of 1/400. When the lateral force dropped below 85% ofthe maximum strength or the axial force could not be sustained, thetest was stopped. The loading procedure for CFSCW-10–12 wassimilar to CFSCW-1–9.

2.3. Instrumentation

Loads, displacements and strains at critical locations were mea-sured for CFSCW-1–9, as shown in Fig. 8. Six displacement meterswere mounted to a rigid steel reference frame to measure lateraldisplacements, with the top one at the same height of the loadingpoint. Two pairs of diagonally arranged displacement meters wereused to measure the shear deformations of the specimens. Six linearvariable differential transducers (LVDTs) were mounted verticallyalong the length of each wall so that the deformation of the wall basecould be determined. Steel strain gauges were arranged at differentlocations as shown in Fig. 8(b) to measure the strain distribution ofthe specimens. The instrumentation for CFSCW-10–12 was similar toCFSCW-1–9, and is not illustrated here.

3. Experimental results

3.1. Failure modes

Specimens CFSCW-1–11 all failed in a progression of fractureinitiation, local buckling, and fracture propagation. Specimen CFSCB-3 is taken for an example to illustrate the failure progression ofthese specimens. A list of observed behavior of CFSCB-3 is providedas follows:

c) CFSCW-3

-3000

-2000

-1000

0

1000

2000

3000

-3000

-2000

-1000

0

1000

2000

3000

-3000

-2000

-1000

0

1000

2000

3000

-3000

-2000

-1000

0

1000

2000

3000

-3000

-2000

-1000

0

1000

2000

3000

-3000

-2000

-1000

0

1000

2000

3000

-3000

-2000

-1000

0

1000

2000

3000

-3000

-2000

-1000

0

1000

2000

3000

-3000

-2000

-1000

0

1000

2000

3000

-1500

-1000

-500

0

500

1000

1500

-1800

-1200

-600

0

600

1200

1800

-60 -40 -20 0 20 40 60 80

-30 -20 -10 0 10 20 30 -20 -10 0 10 20 -20 -10 0 10 20-2400

-1600

-800

0

800

1600

2400

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Top displacement (mm) Top displacement (mm) Top displacement (mm)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Top displacement (mm)-60 -40 -20 0 20 40 60

Top displacement (mm)

-60 -40 -20 0 20 40 60

Top displacement (mm)

-60 -40 -20 0 20 40 60Top displacement (mm)

-60 -40 -20 0 20 40 60Top displacement (mm)

-60 -40 -20 0 20 40 60Top displacement (mm)

-60 -40 -20 0 20 40 60

Top displacement (mm)-60 -40 -20 0 20 40 60

Top displacement (mm)

-60 -40 -20 0 20 40 60 80Top displacement (mm)

l) CFSCW-12k) CFSCW-11j) CFSCW-10

g) CFSCW-7 h) CFSCW-8 i) CFSCW-9

f) CFSCW-6e) CFSCW-5d) CFSCW-4

a) CFSCW-1 b) CFSCW-2

Fig. 11. Lateral force versus top displacement curves for all the specimens.

212 J.-G. Nie et al. / Journal of Constructional Steel Research 88 (2013) 206–219

(1) Cycle at the drift ratio of 1/131, the lateral force was 0.81 Pm(where Pm is the maximum strength), 0.3 mm wide crackswere observed at the boundary corners of the wall base.

(2) Cycle at the drift ratio of 1/102, the lateral force was 0.92 Pm,the steel plates at the boundary corners fractured, and themaximum fracture width was approximate 3 mm, as shownin Fig. 9(a).

(3) Cycle at the drift ratio of 1/79, the lateral force was Pm, theplate fractures ran through along the wall thickness at the

boundaries, as shown in Fig. 9(b); local buckling occurredat the flange plates of the boundary columns, and the buckleamplitude was approximate 4 mm, as shown in Fig. 9(c).

(4) Cycle at the drift ratio of 1/66, P decreased to 0.94 Pm, the platefractures extended along the wall length, and the largest fracturelength was approximate 215 mm, as shown in Fig. 9(d); Thevertical welds at the boundary corners fractured, as shown inFig. 9(e); local buckling extended to the web plates of the bound-ary columns, and the buckle amplitude was approximate 10 mm.

-60 -40 0-3000

-2000

-1000

0

1000

2000

3000

CFSCW-1CFSCW-2CFSCW-3

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

Lat

eral

for

ce (

kN)

-3000

-2000

-1000

0

1000

2000

3000

CFSCW-3CFSCW-4CFSCW-5

a) b)

-3000

-2000

-1000

0

1000

2000

3000

CFSCW-3CFSCW-6CFSCW-7

-3000

-2000

-1000

0

1000

2000

3000

CFSCW-3CFSCW-8CFSCW-9

c) d)

-4000

-3000

-2000

-1000

0

1000

2000

3000

4000

CFSCW-3CFSCW-10

(sca led)

-30 -20 -10 0 10 20 30-2400

1600

-800

0

800

1600

2400

CFSCW-10CFSCW-11CFSCW-12

e) f)

Top displacement (mm)-20 20 40 60 -60 -40 0

Top displacement (mm)-20 20 40 60

-60 -40 0

Top displacement (mm)-20 20 40 60

-60 -40 0

Top displacement (mm)

-20 20 40 60

-60 -40 0

Top displacement (mm)

Top displacement (mm)

-20 20 40 60

Fig. 12. Envelops of lateral force versus top displacement curves.

213J.-G. Nie et al. / Journal of Constructional Steel Research 88 (2013) 206–219

(5) Cycle at the drift ratio of 1/55, P decreased to 0.77 Pm, the fracturecontinued to propagate, and the largest horizontal fracture lengthwas approximate 430 mm, as shown in Fig. 9(f).

The failure processes of specimens CFSCW-1, -2, -6, -7, -9 weresimilar to CFSCW-3, for same steel plates were used for these speci-mens. Thinner plates were used for specimens CFSCW-4 andCFSCW-5 as shown in Table 1, so local buckling occurred earlier forthese two specimens compared to the aforementioned six specimens.For CFSCW-4, local buckling occurred at the drift ratio of 1/100, whilefor CFSCW-5, local buckling occurred at the drift ratio of 1/132. Fur-thermore, the buckling region of CFSCW-5 was larger than those ofother specimens of the same size, as shown in Fig. 10(a).

The steel plates of specimens CFSCW-5 and CFSCW-8 didn't frac-ture at the wall base, but fractured at the locations that approximate170 mm from the wall base, and the fracture didn't extend beyondthe regions of boundary columns during the loading process, asshown in Fig. 10(a) and (b).

For specimens CFSCW-10 and CFSCW-11, the steel plates frac-tured at the locations that approximate 120 mm from the wall baseat the drift ratio around 1/133, and the fracture didn't extend beyondthe regions of boundary columns during the loading process; localbuckling also occurred at the drift ratio around 1/133, and graduallyextended to the middle of the wall specimen. At the end of testing,local buckling occurred along the whole length of the wall base asshown in Fig. 10(c) and (d).

Fig. 13. Determination of the yield and ultimate points.

-60 00

400

Kj (

kN/m

m)

CFSCW-3CFSCW-4CFSCW-5CFSCW-6

CFSCW-8CFSCW-9

CFSCW-7300

200

100

-40 -20 20 40 60

Top displacment (mm)

Fig. 14. Cyclic stiffness versus top displacement curves for CFSCW-3–9.

214 J.-G. Nie et al. / Journal of Constructional Steel Research 88 (2013) 206–219

The typical failure modes of CFSCW-12 were local buckling of steelplates and obvious vertical weld fracture with slight horizontal frac-ture at the boundary columns as shown in Fig. 10(e).

For specimens CFSCW-1, -2, -3, -4, -6, -7, -9, the steel plates frac-tured at the weld regions at the wall bases. In the initiation of frac-ture, the lateral force was already close to the maximum strength,and at the maximum strength, the fractures hadn't propagatedalong the wall length. Therefore, the weld fractures might not signif-icantly affect the loading-carry capacities of the specimens, for theloading-carry capacities of the specimens had nearly fully developedbefore the initiation of fracture. However, the fracture propagationcaused the final loss of loading-carry capacities, which might affectthe deformation capacities of the specimens. In the actual practice,the structural walls are anchored in the RC foundations, so this failuremode will not occur.

3.2. Lateral force versus top displacement curves

Lateral force versus top displacement curves for all the specimens areshown in Fig. 11. When the drift ratio of CFSCW-1 reached −1/100,the maximum top displacement in the north direction was keptunchanged in the subsequent loading process. The displacement meter

Table 3Measured characteristic strengths and displacements.

Specimen Loading direction Py (kN) δtop,y (mm) δtop,y/hwa

CFSCW-1 (+) 2320 18.9 1/136(−) 2220 17.9 1/143

CFSCW-2 (+) 2235 18.7 1/137(−) 2070 13.6 1/189

CFSCW-3 (+) 2386 22.9 1/112(−) 2333 20.7 1/124

CFSCW-4 (+) 1919 15.5 1/166(−) 1908 15.3 1/168

CFSCW-5 (+) 1860 16.2 1/159(−) 1794 15.9 1/162

CFSCW-6 (+) 2088 19.2 1/134(−) 1880 16.7 1/154

CFSCW-7 (+) 2437 20.4 1/126(−) 2175 18.0 1/143

CFSCW-8 (+) 2152 17.5 1/147(−) 2051 17.2 1/149

CFSCW-9 (+) 2327 18.6 1/138(−) 2127 17.6 1/146

CFSCW-10 (+) 1013 10.0 1/150(−) 895 8.0 1/188

CFSCW-11 (+) 1151 7.8 1/144(−) 1162 7.0 1/161

CFSCW-12 (+) 1820 5.1 1/147(−) 1676 4.8 1/156

a hw is the wall height.

that controlled the loading process of specimen CFSCW-2 was brokenwhen the top displacement was around 20 mm. This was found whenthe actual top displacement had already exceeded 30 mm, so fewer cy-cles were performed in the testing of CFSCW-2 compared to specimensCFSCW-1 and CFSCW-3. These factors led to the hysteresis loops ofspecimens CFSCW-1–3 varying from each other. However, the envelopcurves of these three fundamental specimens were close to each other,as shown in Fig. 12(a). This indicates that the test results are reliableand consistent, and CFSCW-3 can be taken to represent the fundamentalspecimens in the subsequent analysis.

Pinching phenomena can be seen in the hysteresis loops ofspecimens CFSCW-1–4, CFSCW-6, CFSCW-7, and CFSCW-9, whichare caused by the fracture propagation and closure at the weld re-gions at the wall bases. The hysteresis loops of specimens CFSCW-5and CFSCW-8 are plumper, as the steel plates fractured at thelocations approximate 170 mm above the weld regions at the wallbases and the fracture didn't significantly propagate. Therefore,CFSCW-5 and CFSCW-8 had higher energy dissipation capacitiesthan other specimens of the same scale. The hysteresis loops ofspecimens CFSCW-10–12 are even plumper as shown in Fig. 11.The reason is also that the steel plates of these three specimensfractured more slightly. Based on these results, it is clear that the

Pm (kN) δtop,m (mm) δtop,m/hw δtop,u (mm) δtop,u/hw

2691 34.7 1/74 44.7 1/572602 29.5 1/87 – –

2655 28.3 1/91 40.6 1/632422 31.6 1/81 41.6 1/622726 32.4 1/79 39.3 1/652667 32.1 1/80 39.3 1/652181 25.1 1/102 42.0 1/612215 25.2 1/101 39.9 1/642165 31.4 1/82 46.6 1/552074 30.5 1/84 39.2 1/662472 32.2 1/80 38.6 1/672242 26.4 1/98 35.4 1/732808 32.0 1/80 38.5 1/672524 25.3 1/101 37.9 1/682511 32.1 1/80 46.9 1/552365 32.1 1/80 43.1 1/602717 31.6 1/81 37.0 1/692497 25.6 1/101 34.5 1/741194 14.9 1/101 23.3 1/641040 13.9 1/108 23.1 1/641330 14.3 1/79 16.5 1/681400 12.0 1/94 14.8 1/762064 11.5 1/65 13.9 1/541971 11.9 1/63 13.7 1/55

Kj/K

0

CFSCW-3CFSCW-10CFSCW-11CFSCW-12

1.0

0.020.010.00-0.01-0.020.0

0.2

0.4

0.6

0.8

Drift ratio

Fig. 15. Kj/K0 at different ratios for CFSCW-3 and CFSCW-10–12.

215J.-G. Nie et al. / Journal of Constructional Steel Research 88 (2013) 206–219

fractures at the weld regions at the wall base can have a profoundeffect on the energy dissipation capacity of the wall specimen. Theperformances of the wall specimens can be improved if the wallsare embedded into RC foundation girders so that the fractures atthe wall bases can be prevented.

Specimens CFSCW-3–5 were designed to vary the steel platethickness and steel content ratio as shown in Table 1, but the valuesof fyAs are close for these three specimens, and the ratio of thevalues of fyAs is 1.06:1.00:1.00. This is because the yield strengthincreases when the steel plate thickness decreases as shown inTable 2, which violates the original intention of the test. Theenvelop curves of specimens CFSCW-4, -5 are very close as shownin Fig. 12(b). The peak force of CFSCW-3 is about 22% larger thanthose of CFSCW-4 and CFSCW-5.

Specimens CFSCW-3, -6 and -7 were designed to vary the concretestrength. The concrete cubic compressive strengths of these threespecimens are 86.1 N/mm2, 65.0 N/mm2, and 102.6 N/mm2, respec-tively as shown in Table 1. The positive and negative peak forces ofCFSCW-6 are 89.5% and 84.1% of those of CFSCW-3, respectively,while the peak forces of CFSCW-3 and CFSCW-7 are close. CFSCW-7is expected to have higher load-carrying capacity than CFSCW-3,but due to severe fractures in the test, the load-carrying capacity ofCFSCW-7 didn't fully develop.

It is theoretically expected that the peak force of CFSCW-8 is largerthan that of CFSCW-3, but the test result is opposite as shown inFig. 12(d). The reason might be the premature fracture of the steel

0

CFSCW-3

CFSCW-4CFSCW-5CFSCW-6

CFSCW-8CFSCW-9

CFSCW-7

1.1

1.0

0.9

0.8

0.7-60 -40 -20 20 40 60

Top displacment (mm)

a) CFSCW-1~9

Stre

ngth

red

uctio

n fa

ctor

jη

Fig. 16. ηj at differen

plates at the locations above the wall base. The lateral force ofCFSCW-8 degraded more slowly after the peak force was attained.

The envelop curve of CFSCW-9 is close to that of CFSCW-3 as shownin Fig. 12(d), which indicates that the mesh reinforcement hardlyaffects the mechanical behavior of the common proportioned CFDSPwalls. However, it is necessary in the actual super high-rise buildingsto restrain the shrinkage and creep of the infill concrete so that theconcrete separation from surface steel plates can be prevented.

The design parameters of specimens CFSCW-3 and CFSCW-10 areidentical except the specimen scale. The envelop curve of CFSCW-10is scaled to that of the same model size as CFSCW-3 as shown inFig. 12(e). The scaled CFSCW-10 has larger initial stiffness and peakforce than CFSCW-3, while the deformation capacities of them areclose. The ratio of the peak forces of them is 1.29:1, which is partlycaused by the higher strength of steel plates used in CFSCW-10. Thesmaller size model may exhibit higher load-carrying capacity, butthe size effect may not have significant effect on the behavior of thewall specimens. The behavior of the actual size wall can be predictedfrom the test results of the wall specimens.

Local buckling occurred along the whole length of CFSCW-10 andCFSCW-11, while the steel plate buckling of CFSCW-12 mainly oc-curred at the boundary columns. This indicates that compression–flexure behavior occupied more primarily as the shear span ratio in-creased. The peak force increases as the shear span ratio decreasesas shown in Fig. 12(f), but the maximum base moments of thesethree specimens are close to each other. This is because the shearstrengths are all higher than the flexural strengths, and the flexuraldominated failure occurred for these three specimens.

3.3. Characteristic strengths and displacements

The yield and ultimate points are determined from the envelopcurve as shown in Fig. 13. δtop,u and δtop,y are the ultimate and yielddisplacements, respectively.

The measured characteristic strengths and displacements of allthe specimens are listed in Table 3. The ratios of the maximumstrength Pm in the positive direction to the negative direction foreach specimen are between 0.95 and 1.15. The yield strength Py isabout 85% of the maximum strength for each specimen. The driftratios corresponding to the yield strengths are between 1/189 and1/112, with an average value of 1/147. The ultimate drift ratios arebetween 1/76 and 1/54 with an average value of 1/64, which demon-strates the high deformation capacity of the specimens under highaxial compressive force.

CFSCW-10CFSCW-11CFSCW-12

1.0

0.8

0.6

-0.02 -0.01 0.00 0.01 0.02

Drift ratio

b) CFSCW-10~12

Stre

ngth

red

uctio

n fa

ctor

jη

t loading levels.

0

0

10

20

CFSCW-3CFSCW-4CFSCW-5

CFSCW-8CFSCW-9

CFSCW-6

CFSCW-7

Top

dis

plac

emen

t due

tosh

ear

defo

rmat

ion

(mm

)

Top

dis

plac

emen

t due

tosh

ear

defo

rmat

ion

(mm

)

0

0

5CFSCW-10CFSCW-11CFSCW-12

-10

-20-60 -40 -20

Top displacment (mm) Top displacment (mm)20 40 60

a) CFSCW-1~9

-30 -20 -10 10 20 30-10

-5

10

b) CFSCW-10~12

Fig. 17. Top displacements due to shear deformations.

216 J.-G. Nie et al. / Journal of Constructional Steel Research 88 (2013) 206–219

3.4. Stiffness and strength degradation

The stiffness properties of the specimens under cyclic loading canbe evaluated using the cyclic stiffness, which can be calculated fromEq. (5) [26].

Kj ¼∑nj

i¼1Pij

∑nji¼1δ

ij

ð5Þ

where Kj is the cyclic stiffness; δji is the maximum top displacement atthe i cycle of the j loading level; Pji is the lateral force corresponding toδji; nj is the cyclic number of the j loading level.

The initial stiffness considering flexural and shear deformationscan be calculated using Eq. (6).

K0 ¼ 1h3w

3EIgþ þ khwGAg

ð6Þ

where K0 is the initial stiffness; hw is the wall height; EIg is the flexuralrigidity of the composite section, which is equal to EsIs + EcIc for thesymmetry of the cross section; GAg = GsAs + GcAc is the shear rigid-ity of the composite section; k is taken as 1.2 for the rectangularsection.

CFS

CW

-3

CFS

CW

-4

CFS

CW

-5

CFS

CW

-6

CFS

CW

-7

CFS

CW

-8

CFS

CW

-9

CFS

CW

-10

CFS

CW

-11

CFS

CW

-12

0.0

0.2

0.4

0.6

0.8

1.0

Shea

r de

form

atio

nco

ntri

butio

n

Fig. 18. Ratios of shear deformation contributions at the ultimate drift levels.

The cyclic stiffness of CFSCW-3–9 and Kj/K0 of CFSCW-3 andCFSCW-10–12 at different loading levels are presented in Figs. 14and 15, respectively. Considerable reduction in stiffness can be seenin each specimen as larger top displacement is imposed. At theultimate drift ratio, the ratios of the cyclic stiffness to the calculatedinitial stiffness for all the specimens range from 19% to 27%.CFSCW-3–9 have close cyclic stiffness at all the loading levels,which indicates that changes in the concrete compressive strengthand steel plate thickness don't significantly change the stiffnessproperties of the specimens. Furthermore, the reduced-scale andshear span ratio don't evidently affect the stiffness degradation pat-terns as shown in Fig. 15.

Because of the cumulative damages in the specimens under cyclicloading, the peak force attained at the later cycle of the same loadinglevel may be smaller. The strength reduction factor ηj is defined asshown in Eq. (7).

ηj ¼P2j

P1j

ð7Þ

where Pij is the measured peak force at the i cycle of the j loading

level.

Fig. 19. Stress–strain relation for steel plates and reinforcement.

Fig. 20. Stress–strain relation for concrete.

Table 4Calculated and measured maximum strengths.

Specimen Loading direction Pmt (kN) Pm

c (kN) Pmc /Pmt

CFSCW-3 (+) 2726 2191 0.80(−) 2667 0.82

CFSCW-4 (+) 2181 2112 0.97(−) 2215 0.95

CFSCW-5 (+) 2165 2059 0.95(−) 2074 0.99

CFSCW-6 (+) 2472 1916 0.78(−) 2242 0.85

CFSCW-7 (+) 2808 2374 0.85(−) 2524 0.94

CFSCW-8 (+) 2511 2367 0.94(−) 2365 1.00

CFSCW-9 (+) 2717 2203 0.81(−) 2497 0.88

217J.-G. Nie et al. / Journal of Constructional Steel Research 88 (2013) 206–219

Fig. 16 plots the values of ηj at different loading levels forCFSCW-3–12. Before the maximum strength is reached, ηj basicallyranges from 0.94 to 1.0, and occasionally exceeds 1.0 for thedeviation of measurement. After the maximum strength is reached,ηj decreases evidently. The values are mostly below 0.95 and someare even below 0.85. For CFSCW-12, no evident strength degradationis observed before the ultimate drift level, while great strength reduc-tion takes place at the second cycle when the drift ratio is 1/57, whichis quite different from other specimens.

3.5. Deformation analysis

Both flexural and shear deformations contribute to the top dis-placements of specimens. The top displacement due to shear defor-mation can be evaluated based on the measurement of diagonallyarranged displacement meters, although some deficiencies exist inthis approach [27]. As shown by the linear relationship in Fig. 17,the contributions of shear deformations to the top displacementsare almost unchanged in the whole testing, although slight increasesare observed as the imposed top displacements increase. Fig. 18 pre-sents the ratios of shear deformation contributions to the top dis-placements at the ultimate drift levels in the positive direction. Asexpected, the contribution of shear deformation to the top displace-ment increases as the shear span ratio decreases. For the specimenswith shear span ratio of 2.0 (CFSCW-3–10), the ratios of shear defor-mation contributions are all about 20%. For CFSCW-11 and CFSCW-12,the ratios of shear deformation contributions are 48.5% and 63.0%,respectively.

4. Design recommendations

4.1. Strength analysis

Flexural failure occurred for all the specimens, so section analysismethod can be used to calculate the maximum strengths of speci-mens. Several hypotheses are introduced. Plane section is assumedto remain plane after deformation. Buckling and fracture of the steelplates are not considered in the analysis. A tri-linear model [28] isused for the stress–strain relation of steel plates and reinforcementas shown in Fig. 19. Fig. 20 presents the stress–strain relation of con-crete. The model suggested by Belarbi A and Hsu T [29] is used for thetensile stress–strain relation of concrete. The concrete in the bound-ary columns and wall body is all restrained by surrounding steelplates, so the compressive stress–strain relation for concrete confined

by the steel tube [30] is used. The parameters in Fig. 20 are calculatedusing following equations:

f0

cc ¼ f0

c þ 4f le ð8Þ

f le ¼ −6:5Rf0

c

� �1:46f y

þ 0:12 f0

c

� �1:03 ð9Þ

R ¼ bt

ffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffiffi12 1−ν2

s� �4π2

s ffiffiffiffiffif yEs

sð10Þ

εcc ¼ εc0 1þ 5f0

cc

f0c−1

!" #ð11Þ

εc0 ¼0:002 f

0

cb40N=mm2� �

0:0028−0:000840f0c

f0

c≥40N=mm2� �

8><>: ð12Þ

r ¼ EcEc−f

0cc=εcc

ð13Þ

Ec ¼ 3320ffiffiffiffiffif0c

qþ 6900 N=mm2

� �ð14Þ

Z ¼0 R

f0

c

f y≤0:0039

!

23400Rf0

c

f y−91:26 R

f0

c

f y> 0:0039

!8>>>><>>>>:

ð15Þ

εcu ¼

0:04 Rf0

c

f y≤0:042

!

14:50 Rf0

c

f y

!2

−2:4Rf0

c

f yþ 0:116 0:042bR

f0

c

f yb0:073

!

0:018 Rf0

c

f y≥0:073

!

8>>>>>>>>>><>>>>>>>>>>:

ð16Þ

where b and t are steel plate width and thickness, respectively; vs isthe Poisson's ratio of steel; Es and Ec are the elastic modulus of steeland concrete, respectively. The stiffeners and batten plates areequivalent to a continuous steel plate with the same cross-sectional area.

The cross-sections of the specimens are subdivided into severalfibers including concrete fibers, steel plate fibers and reinforce-ment fibers. A MATLAB program is developed to calculate themoment–curvature curves of specimens based on the preceding

Steelflange

Stud

Secondary stiffener

Local bucklingregion

Secondarystiffener

Local bucklingregion

Stud

Primarystiffener

Steel flange

a) CFSCW-5 b) CFSCW-10 and CFSCW-11

Fig. 21. Local buckling regions.

218 J.-G. Nie et al. / Journal of Constructional Steel Research 88 (2013) 206–219

hypotheses. The maximum moment Mm can be obtained from themoment–curvature curve and the maximum strength is calculatedas

Pm ¼ Mm=hw: ð17Þ

The calculated maximum strengths Pmc and measured maximum

strengths Pmt of CFSCW-3–9 are compared in Table 4. The calculated

results correlate well with the test results for most of thespecimens, and the strengths are generally underestimated by thenumerical method.

4.2. Detailing requirements

Local buckling is not evident in larger-size specimens exceptCFSCW-5 which had 3 mm steel plates. For CFSCW-5, local bucklingoccurred in the region closed by adjacent stiffeners (or the steelflange of boundary column) and studs as shown in Fig. 21(a).

The secondary stiffeners of CFSCW-10 and CFSCW-11 did not playa role in preventing the surface steel plates from buckling for its tooweak stiffness. As a result, for these two specimens, the local bucklingregions were observed within adjacent primary stiffeners (or thesteel flange of the boundary column) and studs as shown inFig. 21(b).

The deformation of the surface plates of the boundary columnsand each compartment of the wall body is restrained by the infill con-crete and steel plates perpendicular to them, which is similar to thatof the tube plates of the CFST columns. Therefore, the upper limitvalue of the steel plate width-to-thickness ratios of CFST columnscan be referred to use for the boundary columns and wall body com-partments of the CFDSP walls. The upper limit value of width-to-thickness ratios for CFST columns recommended in ACI 318-99[31] and Eurocode 4 [32] is given in Eqs. (18) and (19), respectively.

bt≤

ffiffiffiffiffiffiffiffi3Esf y

sð18Þ

bt≤52

ffiffiffiffiffiffiffiffiffi235f y

s: ð19Þ

For Es = 2.06 × 105 N/mm2 and fy = 345 N/mm2, the upperlimit values of width-to-thickness ratios calculated using Eqs. (18)and (19) are 42.4 and 42.9, respectively. The width-to-thickness ra-tios of the larger-size specimens (the width was taken as the distancebetween two adjacent stiffeners) using 5 mm, 4 mm and 3 mm steel

plates and smaller-size specimens (the width was taken as the dis-tance between two adjacent primary stiffeners) are 21.4, 26.8, 35.6,and 41.7, respectively, which all satisfy Eqs. (18) and (19). However,more evident local buckling occurred as the width-to-thickness ratioincreased.

5. Conclusions

A new detailed high-strength concrete filled double-steel-platecomposite wall was conceived, and twelve specimens were testedunder high axial compressive forces and reversed cyclic lateralloads. The proposed CFDSP wall exhibited good seismic behavior inthe tests and could provide a favorable choice for the core wall insuper high-rise buildings. The main conclusions drawn in this re-search can be summarized as follows:

(1) No evident local buckling could be observed in the whole load-ing process as a result of reasonable width-to-thickness ratiosof steel plates and properly arranged batten plates, which indi-cates that the surface steel plates and infill high-strength con-crete can work compatibly in the whole loading process.

(2) The new detailed composite wall with high-strength concreteexhibited large lateral deformation capacity under high axialcompressive force since the measured ultimate drift ratios ofall the specimens ranged from 1/76 to 1/54 with an averagevalue of 1/64.

(3) The contribution of shear deformation to the total top displace-ment is almost unchanged in the whole loading process. Forthe specimens with shear span ratio of 2.0, 1.5 and 1.0, the ra-tios of shear deformation contributions were about 20%, 48.5%and 63.0%, respectively.

(4) The proposed strength calculation approach based on the sec-tion analysis method can give satisfactory predictions formost of the specimens when compared to the test results.

(5) The upper limit value of steel plate width-to-thickness ratios toprevent the steel plates from local buckling before yielding arediscussed for design purpose.

Acknowledgments

The work reported in this paper was supported by the NationalKey Technology R&D Program of China (2011BAJ09B01), NationalScience Fund of China (51178246) and Tsinghua University InitiativeScientific Research Program (2010Z03078). The support is gratefullyacknowledged.

219J.-G. Nie et al. / Journal of Constructional Steel Research 88 (2013) 206–219

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