International Conference in Tubular Structures-1996

ìl

Get the design fundamentals, straightfon¡vard concepts and key specificationsnecessary to take full advantage of manufactured steel tubes' mechanicalpropefties, light weight and aesthetic appeal in steel construction.

Intsnnalional Gonfsrenceûn Tuhulan Slnuctur'GsMay 9-10, 1996 . Vancouver, British Columbia

#H

This is a not-to-be-missed oppoftunity for structural engineers, fabricators, and

architects to be briefed on static design. fatigue design, seismic design, bridge design,

concrete-filling, innovative joining methods. and computer-based tools by some of the

world's leading experls on Hollow Structural Sections (HSS).

American Welding Society

Endorsing 1rganizaîions . American Society of Civil Engineers . American lnstitute of Steel Construction. Steel Tube lnstitute of North America . University of Toronto

sponsors

6 welding lnstitute of Canada

\-rl

:f

I

I

1

I

II

Thble of Contents

Keynote Presentation:Limit States Design, Hollow Structural Sections, and Welds 1

,:D. J. L. Kennedy, Univenity of Alberta i .

Desigu Rules Key to CompeÌitive Tirbular Structu¡es . . .. : . : .... 19

R. M. Bent, Welding Instituæ of Canada

Resistance Ïhbtes for Welded Hollow Structurat Section Tbuss Connections 32

J. A. Packer, University of Toronto; G. S. Frater, Hatch Associates;and S. Kitipornchai, University of Queensland i

Welded Circular lfollow Section Tbuss Connections 48

P. W: Marshall, MHP Systems Engineering

Simpte Beam Connection to Ilollow Structural Section Columns 55

D. R. Sherman, University of Wisconsin

Fatigue of Hollow Structural Section Welded Connections 64

A. M. van Wingerde and J. A. Packer, University of Toronto

Earthquake-Resistant Design Provisions for Tl¡bular Structurcs 74

Y. Kurobane and K. Ogawa, Kumamoto University

Fire Performance of Concrete-FilledTirbular Columns. ...... 86

V. K. R. Kodur and T. T. Lie, National Fire Laboratory

Tubular Offshore Structures 97

P. W. Marshall, MIIP Systems Engineering

Design of Hollow Struqtural Section Columns and Beam-Ç61¡¡mns . . 110

D. R. Sherman, University of Wisconsin

Guide to the Ilollow Structural Section Guides and Codes .. . . 118

J. A. Packer, University of Toronto; a¡rd S. Kitipornchai, University of Queensland

Concrete-Filled Hollow Steel Sections. .. .. 126

H. G. L. Prion, University of British Columbia

Fundamental Criteria for Welding Thrbular Steel . 137

R. M. Bent, Welding Insútute of Canada

ill

Bending, Bolting and Nailing of lfoilow Structural Sections. . 150

J. E. Henderson, Henderson Engineering Services

Fabrication and Tnspection Practices for l{elded Ïtrbutar Connections . . 162J. 'W. Post, J. rñ/. Post Associates,Inc.

i : -i

Design of llalf-Through or'?ony" Thuss Bridges Using Squarg orRectangular llollow Structural Sections. . . 179

S. J. Herth, Continental Bridge ,:

Case Studies of Recent Ti¡bular Stnrctures .... . 189

C. M. Allen, Adjeleian Allen Rubeli, LTD' ,:':

lVelding of Structural Alrrminum Ïbbing.R. Bonneaû, Canadian Welding Bureau

The Challenge of Knowledge-Based Expert Systemsin the Future of the Design of Ttrbular Structures . . . . . 216G. Davies, W. Tizani, and K. Yusuf, University of Nottingham

lv

LIMIT STATES DESIGN, HOLLOW STRUCTURAL SECTIONS, AND \ilELDS

D. J. L. Kennedy*

ABSTRACT

The rationale of limit states design with its inherent advantages over working stress design isdiscussed. Among other advantages, because, for the ultimate limit states, LSD focuses on thepossible modes of failure, it fosters an examination of the true behaviour and the writing ofstrength or resistance formulations that reflect this behaviour. Within this conceptual basis, thedevelopment of some of the provisions of design standards for hot and cold formed hollowstructural sections, concrete-filled hollow structural sections, partial penetration g¡oove weldsand fillet welds at varying orientations is presented. The resistance formulations includeresistance factors that account not only for the variation in material and geometric properties butalso for the statistical fit of the formulation to the test results, i.e., the bias coeffrcient and theco effi ci ent of va¡iation of the test-to-predi cted ratio.

KEYWORDS

Fillet welds, hollow structural sections, Iimit states design, partial penetration groove welds,resistance formulations, resistance factors, statistical evaluation, test-to-predicted ratios.

LIMIT STATES DESIGN

General

Limit States Design, the only design methodology sanctioned for steel stn¡ctures in the NationalBuilding Code of Canada since 1990, is rapidly gaining world-wide acceptance. In the UnitedStates of America, when applied to steel structures, it is called Load and Resistance FactorDesign, while for concrete structures, the term Ultimate Strengfh Design is used. Thedesignation as used here is more universal in use and encompasses all the classes of limit states

and not just those related to ultimate or failure conditions.

Limit States and its classifications

Limit states are those limiting states or conditions of a structure at u'hich it ceases to fulfill someintended function. Therefore the probability of exceeding any limit state is kept to an acceptablelow level. Limit states design is that design philosophy in which the designer, recognizing thevarious limit states, proportions the structure such that these probabilities are attained. Currently

Professor Emeritus, Dept. of Civil Engineering, University of Alberta, Edmonton, AB. Canada, T6G 2G7

limit søtes a¡e classif¡ed as seviceability, fatigue and ultimate limit states.

Seviceability limit states are those associated with the provision of proper acceptable service

conditions such as the limitation of deflections, vibrations, permanent deflections, cracking, and

foundation settlements. The seviceability limit states are to be satisfied during the life of the

stn¡ch¡re at levels of load that are likely to occur with reasonable frequency. These are the so-

called working loads of working stress design and are now called the specified loads. In the

National Buitding Code of Canada (Ref. l), for example, the specified wind load is that of the Iin l0 yearwind.

The fatigue limit state is that associated wittr crack growth under the stress raûge spectn¡m

occtrrring under service conditions. Miner's rule may be used for combining stress range levels.

As well we may need a method for counting the cycles of stress ranges such as the reservoir

method and a method for assessing the remaining fatigue life.

The ultimate limit states are those associated with collapse of all or part of the stn¡cture and

include, rupture or fracture, crushing, buckling, local buckling attainment of the critical, yield

or fully plastic mometrt, mechanism formation, overturning, sliding or foundation failure. The

ultimate limit states must be satisfied during constn¡ction and during the life of the stn¡cture at

levels of load that occur very infrequently, i.e., that have a small probability of being exceeded.

From this we see that Limit States Design (LSD) provides a unified approach in that the designer

explicitly recognÞes the various limit states, i.e., the failure modes and designs against them, all

the while taking into account the statistical variation of both the loads and the resistances.

Formulation of Limit States Desien

Fig. I depicts schematically the probability density functions for the effect of a load, S, and the

resistance, R, of some structural component

:-QR= aSRI

Frequency

Densrty

Magnitude

Fig. L Frequency distribution functions for the effect of a load and a resistance

The nominal values are indicated by S and R while mean values are indicated by S and R. Inworking stress design (WSD), to attempt to keep Ç¡, gretter than S-"*, the nominal values S

and R also shown in Fig.l, ate separated by a global factor of safety, G, thus:

In Limit States Design (LSD), recognizing that both the loads and resistmces vary and that theirprobability density functions will differ from load to load and from resistance to resistance, twofactors, a resistance factor and a load factor a¡e used thus:

G=R/S

R:GS

$R>øS

n>9s0

V. = os/S

P.:S/S

(1a)

(lb)

(2a)

(2b)or

as illustrated in Fig. l, where the LSD inequa,ity is just satisfied.

Comparing eqùations (1b) with (2b) we see t rat the global factor of safety, G, is replaced withthe combination, c/S, but now these two àctors are determined based on their statistical

variations. For more than ¡ryo loads the LSD *xpression becomes:

$R: Ðcr;S¡ (2a)

Currently, in LSD, the two measures of the probabiliry density functions used are the mean

value, e.g., S and R-, and the dispersion about the mean as measured by the standard deviation,

o. The coefficient of variation, V, equal to tlle standard deviation divided by the mean value is

more often used. As the reference or nomir,al value used is unlikely to be the mean value, as

shown in Fig. l, the bias coefficient, p, equa: to the ratio of the mean to nominal value, and its

mean value are also required. Thus we have, ior example, for the effect of loads:

and

(3a)

(3b)

(4a)

(4b)

The probability of failure can be expressed in va¡ious forms such as:

P¡:P(R-S>0)

P¡: P(R/S >1.0)

3

or

P¡= P(ln R/S >0) (4c)

We let X = ln R/S and plot iæ probability density function as shown in Fig. 2.

X = ln (R/S)

Fig.2 Probability Density Function of X

From Fig.2, because the total area under the curve is 1.0, then the area to the left of the originrepresenting values of X less than zero, is the probability of failure. By making the value of Rlarger we shift the curve to the right - as far as we can afford. We position t}re curve such that

the distance from the mean value, i, to the origin is a number, p, times the standard deviation,

o*, of X. The reliability index, p, is selected by calibrating against current good practice. Aftersome mathematical manipulation, lrye obtain, for log-normal distributions and a number of loads:

Ec,s, (5)

where the symbols have their previous definitions and the mean values of the bias coefficienß

are used. The load factors and the resistance factors are linked by this equation and therefore

they are not independent. Furthermore, for both the loads and the resistances, the bias

coeffrcient and coeffrcient of va¡iation, e.8., pn and V¡, are needed. Putting aside how the data

for loads are developed and ho'r load combi¡rations are handled what information is needed to

develop these two measures for the resistances?

The resistance of any structural component depends on the variability of three different

quantities. These are the variability of a material property such as the yield strength, Fy, the

variability of a geometric property such as the plastic section modulus, Z, and the variability ofthe predictive capacity of the design equation such as Mp = ZFy, as determined from

comparisons of test results with that predicted by the simple equation. This laner variability

arises from the fact that all design equations, in the interests of simplicity, contain some

=fts*ololu ]

I.t

4

approximations. In the present case, the formulation is based on fully plastic stress blockswithout strain hardening. The first of these is not attainable and therefore the prediction is toohigh while the second is likely to be present and therefore the prediction is too low. As well,moment gradients have not been considered. Thus there is a va¡iability around the mean for allthree quantities.

Because the three variables are independent the mean value and the bias coefficient of theresistance are given simply as the product of the respective values while the coefficient ofvariation is obtained as the squa¡e root of the sum of the squares of the three quantities thus:

PR = Pc tPu'Pp = PztPry.Py (6a)

(6b)VR=

These equations are used subsequently.

Advantages of Limit States Design

Some argue that LSD only complicates design and increases design time without any realadvantages. This is not factual. Once the initial learning curve is mastered, designs are as e¿rsyor easier to carry out, increased understanding of the design process results and advantagesaccrue as follows.

I.0 Resistance formulations are written transparently as member strengths

The output of structural analyses is the stress resultants acting on the members such as ærialforces, bending moments and shears. That being the case, why not write member resistances ina parallel manner? The resistance formulations are based on the actual behaviour of thecomponent, member or structure. Thus the designer is made aware of the possible failure modesand can then design against them rationally. Inelastic member behaviour is accommodatedautomatically in LSD. For example, the nominal moment resistance of a compact section asformulated in LSD is:

M=Mp =ZFy

However, in WSD this must be expressed in terms of stresses; frequently the extreme frbre stressof stress blocks that vary linearly across the cross-section. Thus, introducing a global factor ofsafety, G, and dividing by the elastic section modulus, S, gives:

(7a)

vfr+vfr+vf vi +vf +vf

o* = M/GS - ZFvlGS = l.l0Fy / G =l.lOFy /1.67 =0.66ry (7b)

This formulation obscures the actual behaviour and appears to suggest that compact beams can

have higher allowable stresses. Moreover the ratio of ZIS varies considerably from the value ofl.l0 used here. Thus, in LSD, the designer is made aware of the behaviour and resort need notbe made to fictitious allowable süesses.. The same condition applies in composite constructionwhere, in LSD, fully plastic stress blocks are incorporated, when appropriate, for both the steel

and concrete. Working stress design does not give a consistent rational method of assessing theflexural resistance.

2.O .Non-linear geometric effects'. '

Progressive standards now reçire that second order geometric effects be considered in the

analysis. In LSD these are evaluated at the factored or collapse load level and therefore are

propedy established as they contain the product of the factored loads acting on the factoreddeflections. Second order amplification at the working load level underestimates these effecg as

indicated in Fig. 3. Ar¡ analysis at the working load level cannot include the second-order non-linear effects due to the change in geometry at the ultimate load.

Load

Deformatíon

Fig. 3 NonJinear geometric effects

3.0 Separate Load and Resistance Factors Determined Statistically

These give rise to reliability levels that are much more consistent and at the same time lead to

better safety and economy. Both these facts are illustrated in Fig. 4 based on Allen (Ref. 2) forthree different design standards. The broad line represents usual load combinations and the fine,

all combinations. Ideally there would be no variation in the reliability index, p, but this would

make the load cõmbinations too complex. The range of I is the least for the LSD standard. It isby far the most consistent. By eliminating low values of p the safety or reliabiliry is improved

and by eliminating the high values economy is achieved.

4Â.

6

6

5

4

1It r¡sr¡at

casqs

all

combinationsF32

I

0

sl6, wsD AIsc, wsD st6.l, LsD

Fig. 4 Range of the reliability index, p, for three standards (Ref. 2)

The Ferry Bridge Cooling Tower collapse, as reported by Allen (Ref. 3), was precipitated byfailure of the tensile reinforcement. This also illustrates the superiority of limit states designarising from the use of both load and resistance factors. The reinforcement was designed, usingWSD procedures, to withstand the difference between the uplift due to wind and the dead loadeffects, which were 0.85 of the uplift, using an allowable stress of 0.50 Fr. Thus thereinforcement a¡ea is found from:

0.50 orA"*, = \l¡,r - D : W - 0.85W : 0. I5W

or Ar,,:0.30Wo,

where the subscript "w" stands for WSD. The wind force to cause yielding, Wu*, is:

0.85o, Ar : 1.50W - 0.85D : 1.50W - 0.85 x 0.85W : 0.778W

or A¡ = 0.915 Wo,

where the subscript "L" stands for LSD. The wind force to cause yielding would be:

Wv¡ - D = Wyl - 0.85W : orA, = 0.915W

(8a)

(8b)

Wy* - D = W, - 0.85W = orAr.,, : 0.30\il (8c)

or Wy*: l.lsW (8d)

that is, only lilToabove the specified wind load. Had LSD been used, witl load factors of 1.50

on wind and 0.85 on the dead load when it is counteractive, and with a resistance factor of 0.85

on yield, the design equation would have been:

(ea)

(eb)

7

(ec)

or WyL= 1.76W (9d)

The increased reinforcement as required by LSD would have prevented collapse at little cost.

This illustrates that a single factor of safety, as used in WSD, simply does not work.

4.0 Tailored Load and Resistance Factors

The use of statistical analyses also paves the way for the rational development of load and

resistance factors tailored to the specific site conditio$¡ as may be desirable for major

engineering structures. Such was the case for the Northumberland StrÀit Fixed Crossing. Load

and resistance factors were de¡¡eloped by MKM Engineering (Ref. 4) taking into account the

particular environmental conditions such as wind and ice loadings, values of p of 4 or more as

iequired by the ou¡ner, and recoeûizing the tight contol on the manufactr¡ring of the structural

components.

5.0 Changes in Reliability Levels

As was established, the load and resistance factors are directly related to the value of the

reliability index, p. Thus, by varying the value of p, values of the load and resistance factors can

be determined to take into account such factors as the consequences of failure, the behaviour ofthe component, and the like. Table l, paralleling the work of Allen (Ref. 5) gives values for the

change in p, i.e., Âp, proposed by Kennedy (Ref. 6). Other values of Âp could be considered.

The target value of the reliability index may be found as:

9r:3.50+EÂF, >2.0

Table l: Adjustment factors, Åp, to the reliability index, p

(10)

Life Safety Factor Description ^PConsequences of failure

Component behaviour

System behaviour

Number of persons at risk

essential for post-disaster services

normalsmall probability of loss of life or economic loss

sudden brittle failurelimited ductilitygradual ductile failurecomponent failure leads to total collapse

component failure leads to contained collapse

component failure leads to local failurelarge loss of lifemoderate loss of lifeminimum loss of life

+0.300

-0.300

-0.35-0.70

0-0.35-0.70+0.30

0-0.30

8

7.0 The Fostering of Research

Limit States Design fosters research. It soon becomes evident in examining the Limit States

Design equation that much research needs to be done to define better the loads and load effects.The first of these deals with the assessment of loads acting on structures; whether they areenvironmental, use and occupancy, vehicular, dynamic, the weight of the structure itself orwhatever. The determination of the effect of loads is the analysis of the structure under theaction of the loads. Computer analyses that take more and more factors.into account and reduceor eliminate drudgery represent significant advances.

On the other side of the equation is the assessment of the resistance of the particular structnralcomponent. While quality control has reduced both the variability of the geometry and of thecharacteristic strength of components, the structural engineering researcher continues to look forbetter models of the behaviour of members, components and structures. The goal is to developmodels for which the bias coefficient is close to 1.0 and the coefficient of variation is low.Reducing the lauer, in particular, is likely to enhance the resistance factor for a given reliabiliry.This is not the problem that is faced by a design engineer where it is perfectly acceptable tomake simplifying assumptions provided only that they are conservative. The question beingaddressed by the researcher is what model predicts the behaviour closely and consistently. It is

this latter aspect that is examined in the next two sections for some aspects of hollow structuralsections, HSSs, and welds.

HOLLOW STRUCTURAL SECTIONS

Class H Hollow Structural Section Columns

Kennedy and Gad Aly (Ref. 7) proposed, in I980, thatColumn Curve of the Stn¡ctural StabilityResearch Council could be used for Class H hollow structural sections, produced in Canada inaccordance with CSA Standard G40.20-1976, with a resistance factor generally greater than0.90. Subsequently the Sl6 Committee on Structural Steel for Buildings incorporated this intothe 516.1 Limit States Design Søndard (Ref. 8) with a resistance factor of 0.90. Thisrepresented a considerable increase in the factored compressive resistance as compared to thelower SSRC Curve 2. It is important to note, as the Standa¡d continues to state, that this higherstrength is for Class H sections and not for Class C sections which have considerably differentproperties. Furthermore the sections must be produced to CSA Standard G40.20 the current

edition of which is that of 1992 (Ref. 9). Hollow structural sections manufactured to ASTMStandard 4500-93 (Ref. l0) do not qualify as the tolerances on wall thicknesses are considerably

less stringent in it. In S 16. I -94, SSRC Curve I is expressed in double exponential form as:

oAFv ( * *"lX

9

Cr= (l l)

in which the resistance factor, Q, is 0.90 and the exPonen! n, is 2.24. The data given in Table 2

are based on the original analysis in which the equation of Galambos and Ravindra (Ref. I l) for

the resistance factor, incorporating a separation factor, d,¡, of 0.55, was used- This is:

0 = pn exp(-ÞcrnVn)

A reliability inder; p, of 3. 0, consistent with the NBCC, was used-

Table 2. Statistical data for HSS Columns

(r2)

Variable vStatic yield strength, FyCross-sectional area, ATest-to-predicted ratio, P

Unit resistance, Ffor l, = 0.00

î,:0.40I = 0.80?,.= 1.20

I = 1.60l. = 2.00ì'= 2.40l,:2.80

1.2400.9850.965

t.2401.229t.174r.0251.040

t.021t.0251.035

0.0920.0340.040

0.0920.0870.0640.0330.0290.0330.0330.031

1.179Ll68l.l l60.9740.9860.9700.9740.984

0.1060.1020.0830.0620.0600.0620.0620.061

0.9900.9870.9730.8800.8920.8750.8800.890

The test-to-predicted ratio is based on tests of Birkemoe (Ref. 12) and the entire procedure was

confirmed by exagining the results of 158 tests reported by Sherman (Ref. 13). In table 2, the

bias coefficients and the coefTicients of variation given for the unit resistance for different values

of the slenderness parameter take into account the bias coeffÏcient and the coefficient ofvariation of the yielå strength, the radius of gyration and the modulus of elasticity and the fact

that as the slendemess incieases the influence of the yield strength decreases and that of the

modulus of elasticity increases. For any value of ln the bias coefficient, Pc,, iS the product of

those for the cross-sectional are4 the test-to predicted ratio and the unit resistance while the

corresponding value of Vç, is found as the square root of the sum of the squares- Thus, for

example, for I:0.80, equation (12) gives:

0=0.985x0.965xl.l74exp(-3.0x055@)=0.973(l3)

We note that the bias coefficient is reasonably close to 1.0 and that the coefficient of variation is

not too high for the range of slenderness ratios. In CSA Standard 516.l a single value of the

resistance factor of 0.90 is used.

10

'I

I

Concrete-Filled Hollow Structural Sections in Flexure

Based on 12 flexural tests on concrete-filled hollow strucA¡ral sections and control tests on the

five different hollow structural sections used, Lu and Kennedy @ef. 14) developed n¡¡o models

to predict the strength of concrete-filled hollow stn¡ctural sections; a "research model" and a"design model". Classifications of the sections, based on measured dimensions and properties,

ranged from Class I to Class 4. By using rectangular sections with the long dimension oriented

both horizontally and vertically and by using sections with a considerable variation in wallthickness, ratios of the concrete and steel areas in compression of 3.1 to 5.6 were tested. Aswell, shear span to depth ratios of I.0 to 5.0 were investigated. Neither of these factors had any

effect on the test-to-predicted moment ratios and therefore the models developed a¡e considered

to be independent of these facûors.

The moment curvah¡re relationship is initially elastic followed by increasing inelastic softening

culminating with a very long plateau of slightly increasing slope until failure occurs. Failurewas precipitated by an upward buckle of the steel top flange. The concrete in the tension zone

was heavily cracked a¡d in the compression zone was crushed where the steel had buckled. On

the average, steel strains reached 14 000 ¡re in compression and23 000 pe in tension.

The concrete prevented inward movement of the steel webs and therefore provided rotational

restraint to the edges of the top flange which could only buckle upwards. Thus the compressive

strains in the steel at failure were very large. Observations indicated that there was no loss ofcomposite action between the steel and the concrete due to lack of shear transfer by friction orbond. Confinement of the concrete by the steel increased its load carrying capacity such that the

ratio of the maximum concrete stress in flexure to the cylinder strength should be taken as 1.00

and not just 0.85 as is the case in reinforced concrete design. The effective rectangular stress

block in the concrete should be taken to extend to 0.85 of the depth to the neutral axis.

The "research" model to predict the ultimate moment resistance is suitable for use when the

strengths of the steel and concrete are known. The concrete compressive resistance is taken as

the concrete strength multiplied by 0.85 of the area of the concrete in compression i.e., the

rectangular stress block extends to 0.85 of the depth to the neutral axis. The steel stress is taken,

both in tension and compression, as the average of that at 14 000 and 23 000 pe. This is valid

for Grade 350 steel with b/t ratios as high as 36.0. The position of the neutral axis is determined

to satisfy horizontal equilibrium. For design, because the strengths of the steel and concrete are

not known a priori, the model is based on the specified minimum yield strength and the 28-day

concrete strength with the neutral axis position again established to satisfy equilibrium. Table 3

shows the test-to-predicted ratios for the two models where, for the design model, the measured

steel yield strength and the measured cylinder strength have been used in the prediction

calculation.

The coefficient of variation for both models is very low indicative of a narrow distribution about

the mean. For the research model the mean value of the test-to-predicted ratio at 1.016 is very

close to l. Thus the research model predicts the strength exceedingly well. The design model

under-predicts the moment by about l9o/o on the average. This is due to the under-assessment

11

Table 3 Test-to-predicted ratios for two models

Test Test momentkì.Iom

Predicted moment, klrlom Test-to-predi cted ratio

Research Design Research Design

cB13cBl5cB22cB31cB33cB35cB4lCB4r'.cBs2cBs3cB55

72.0t39.72t2.4ztt.7211.3

275.2274.7t42.5t4t.4141.4

62.9t23.1t76.2r75.6t75.3248.7248.0tr7.lI16.5t16.4

l.Ml0.991

1.0680.9920.9950.983

1.031

1.0271.015

1.038l.0l I

1.190l.l34l.1901.196

1.200

1.184

1.141

1.138t.236t.260t.227

75.t71.3

t46.5210.72t0.7207.6283.8282.2t4.7t46.7142.9

72.2 63.1

p l'016 1' 188

- Y - - ,- - o'o2s- 0:034

of the steel contribution because the yield strength for the cold rolled HSSs, obtained by the

O-2vo offset method, is considerably less than the stress levels obtained at the large strains the

steel was able to undergo before failure. This high mean value would not be disadvantageous

for design because a resistance factor derived using this test-to-predicted ratio together with the

bias coeffrcients for the yield strength and the cross-sectional properties and with the respective

coefficients of variation would give the desired reliability levels automatically.

WELDS

Partial Penetration Groove,lVelds

partial penetration groove welds do exist. Gagnon and Kennedy (Ref. 15) tested 75 such welds

made with matching electrodes in grade 300W and grade 3504 steel plates, to determine the

overall behaviour -¿ ttt" effects on the strength of percent penetration, plate strength, and the

eccentricity of the load arising from the fact that the welds are not aligned with the æris of the

plate. Nominal penetrations ranging from 20 to 100% were used. The plates were tested singly

*¿ in pairs to establish any differãnces between eccentrically loaded welds and the concentric

loading of a pair of specimens.

The inherent ductility of the welds allowed lateral deflections and straining to take place so that

the eccentrically loaded welds were as strong as concentrically loaded welds' The strengh of

12

the welds is greater than the strength of the plate multiplied by the percent penetration andincreases with increasing lateral restraint that occurs with decreasing penetration as shown inFig.5.

¡I

I

¡¡.l¡l¡¡

!!l¡

60 100

Percent penetration

Fig. 5 Ultimate stress versus percent penetration

This increased strength was attributed to the fact that the weld, heavily strained in tension,attempts to contract laterally but is restrained from so doing by the adjoining less heavily loadedplate material. A biaxial or even triaxial stress state is set up which increases the failure stress.

Extending the von Mises-Hencky yield criterion to the ultimate, for the case when the out ofplane stress is zero, and for the case when the strains in the two orthogonal directions are zero,gives l.l5 and 1.75 times the ultimate tensile stess respectively,for v = 0.3. Furthermore, foreccentrically loaded welds, the moments developed in the plates tend to cause the plates to self-align under the tensile force and the moments are reduced. (This cannot occur for the plates

tested in pairs as they keep each other in the original alignment.) However, in both cases, whenall the weld cross-section is yielded in tension there can be no moment on the weld. A veryreliable model is, therefore, to take the tensile resistance of the partial penetration groove weld,made with matching electrodes, as:

T, =0* pAp Fo

o"

t!¡

804020

(14)

where p is the decimal fraction of the penetration, An and Fu are the area and tensile strength ofthe plate and the resistance factor is to be determined from equation (12) in a slightly modifiedform. Because load and resistance factors have to determined consistently, if equation (12) isused to determine resistance factors with a reliability index of, say 4.0, the correspondingequation for load factors should also be based on an index of 4.0. However, because it is

convenient to use one set of load factors in design based on the general index for members ¿Ìs a

whole, say 3.0, for example, an adjustment factor must be applied to equation (12). Based on

13

Fisher et al. (Ref.16), this is, for our case taking the two indices as 3.0 and 4.0 resPectively,

about 0.93. Gagnon and Kennedy (Ref. 15) give the following bias coefficients: plate strength,

1.090; weld are4 0.978; plate areq 1.015; and the test-ùo predicted ratio, l.l52for the data in

Fig. 5 resulting in pn obtained as the product, equal to 1.246. Combining the corresponding

coãflicients of variation of 0.1013; O.147;0.013; and 0.l72to give V¡ equal to 0-248 results in:

0o, = 0.93 p¡ exp(- Þanvn)

0w =0.93x 1246 exp(-+.oxossx0.2a8) = 0.67

(l sa)

(lsb)

or

.*,

which is that used in CSA Standard Sl6.l. Because the partial penetration groove welds fracture

with little deformation, even though the welds are ductile, to get overall ductile behaviour the

plate must yield before the weld fractures, hence:

pApFutAo& (l6a)

(l6b)

Fillet Welds

Although it has been known for years that the strength of trarsversely loaded frllet welds is

gr""trr-th* that of longitudinally loaded fillet welds and that welds loaded at intermediate

Lgtes have intermediaæ strengths, as reported by Spraragen and Claussen F"f. ll) and bV

Frãeman (Ref. lB) respectively, u-oog others, it is only relatively recently that this has been

recognizeà in design standa¡ds. The more recent work of Butler and Kulak @ef. 19) formed the

basis-for the design tables for eccentrically loaded weld groups in the 7977 Edition of the Limit

States Design fvfanuaf of the Canadian Instin¡te of Steel Constnrction (Ref. 20). Only in the

lgg4 edition of cSA Standard sl6.l is the variation of the factored shear strength with the

direction of loading recognized in the equation:

Y, =0-67 0* A* Xu(1.00+0'50sinls 0)

RrD>¿^Fo

(17)

where S* is the resistance factor for welds, A* is the throat area, X,, is the electrode classification

and 0 is the angle between the æris of the weld and the line of action of the force. This is based

on the work of Miazgaand Kennedy (Ref. 2 I ) and of Lesik and Kennedy (Ref' 22) ' Miazga and

Kennedy gave two ,-."ron, for the increased shear strength as the loading changed from the

tongituiinãl direction, i.ê., parallel to the weld arcis, to the transverse direction' i'e''

oemendicular to the weld a,xis. First, the angle of the failure plane changes continuously from

ffi;¡ ;r ,¡. rongitudinal weld to about 140 for the transverse weld as shown in Fig' 6,

where are plotted their test results as well as the failure angle predicted using the mæ<imum

shear súess theory. Second, as was the case for partial Penetration groove welds' the lateral

14

shear stess theory

Fractr¡re

angle

45

Angle between longinrdinal and load a¡ces

Fig. 6 Variation of fracture angle with the angle between the longitudinal and load axes

restraint provided by the less heavily loaded adjacent plates increases the strength. Forlongitudinal fìllet welds in shea¡ this influence is zero but it increases to a maximum for thetransverse welds when there is a considerable normal force component acting on the rveld inaddition to the shear force. Equation 18, from Lesik and Kennedy (Ref.22), a simplificarion ofthe more complex equations -eiven inMiazga and Kennedy (Ref. 2l), is plotted in Fig. 7 aeainstthe test results reported in the latter. The shear stress is computed for convenience as if it acted

on the throat although this is the failure plane only for the longitudinal welds. This accounts, inconsiderable measure, for the increased strength of transverse welds.

The data in Fig. 7 give a mean bias coefficient for the test-to-predicted ratio of 1.010 w'ith therelatively low coeffrcient of variation of 0.089. With these, the statistical values of orherparameters as reported in Lesik and Kennedy are incorporated as follows. The mean value ofthe effective throat area to the nominal value, p6 is 1.034 with a coefficient of variation, \rç, of0.026. There are two material factors to be considered; the ratio of the tensile strength ro theelectrode classifrcation and the ratio of the shear strength to the tensile strength. This latter ratiois taken as 0.67 in the resistance equation. The mean value of the ultimate tensile strength ofelectrodes divided by the electrode classification, pÀ{r, is 1.123 u'ith Vâ, equal to 0.077. Themean value of the shear suength to the ultimate tensile strength is 0.749. Thus p¡12 is 0.74910.67= l.ll8 and the corresponding value of V¡a2 is found to be 0.12I. Multiplying the biascoefficients together gives a value of p¡ of l.3l and for the coefficient of variation the squareroot of the sum of the squares gives a value of 0.170. As before for partial penetration groovewelds, using a p of 4.0 with an adjustment factor of 0.93 so that the resistance factor can be used

with load factors determined for a 3.0, results in:

90075603015

0r" = 0.93p¡ exp(-Þonvn)

0,,.. =0.93 x 1.3 I I exp (-+.ox 0.55x 0.1 70) = 0.86

(1sa)

(1sc)

i5

1.0 + 0.50 sint'0

U6

Í*

45 60 75

Angle betwecnlongitt¡dinal and load a:ces

Fig.7 Variation of normalized shear strength with angle between the longinrdinal and load æres

This exceeds the value of 0.67 given h 516.l considerably. Lesik and Kennedy give a lower

value, but still greater than 0.67, when test results of others are incorporated.

Equation (17) can be used to develop the inelastic strengths of eccentrically loaded weld groups

of arbitrary configurations, using the method of instantaneous centres, when it is expanded to

include a tefm that accounts for the load deformation respoff¡e of the weld. Thus, writing in

normalized form by dividing by the longinrdinal weld strength and without resistance factors the

resistance of a unit area of weld is:

30l5 g0o

(l 8)

where Í (p) is given by Lesik and Kennedy (Ref. 22) in polynomial form as obtained by

conelating with the load-deformation response for all 42 tests of Miazga and Kennedy' A

polynomiãl is used in order that the descending branch of the curves, after reaching the

mocimum load, can be modelled. It is further necessary to ensure that the ma<imum

deformation that the weld can undergo at the particular angle between the weld æris and the load

is not exceeded. In using equation (18) the test-to-predicted ratio is no longer determined for a

single weld but by comparisons of the load carried by different weld configurations to that

preãicted by equaiion (18). Such work,including the determination of resistance factors, was

caniedout-byiesikandKennedy. Itwasfoundthatthe516.l resistancefactorof 0-67 wasat

leas¡ 6Yo cons ervative.

liang (Ref. 23) advanced this procedure another step by developing and verifying a computer

progiu* for the analysis and ãetermination of the factored resistance of eccentrically loaded

i.tã groups of any arbitrary configuration of line segments when loaded in plane by a load

acting-at any orientation. Input data are: the line of action of the in-plane force, the weld

r"o = [r.oo * ,¡nr-50 e]r(o)

Ír¿ L

16

configuration, the weld size, and the electrode strength. The solution is iterative and begins withan ¿$sumption of the location of the instantaneous centre of rotation. The program then carriesout the iterations necessary to arrive at the correct location of the instantaneous centre and theultimate load that the weld group can carry. One analytical experiment showed that an arbitraryone-third reduction in the deformation that the weld could undergo did not decrease the strengthof the welded connection.

SUMI}ÍARY AND CONCLUSIONS

In addition to the fundamental advantage of Limit States Design over Working Stress Design inproviding much greater consistency in the reliability of structures and in providing economy atthe same time, it has been shown, by particular application to hollow strucû¡ral sections, partialpenetration groove welds and fillet welds, that LSD allows the rational development ofresistance formulations that take into account the inelastic action that occurs ineviøbly inattaining the ultimate load that stn¡ctr¡ral components can carry. The obvious extension is thesecond-order inelastic analysis of structures under factored load combinations accounting forboth material and geometric nonJinearities.

ACKNOWLEDGMENTS

The support of the Natural Sciences and Engineering Resea¡ch Council throughout the course ofthe projects cited here in which the author played a role is gratefully acknowledged.

REFERENCES

l. NBCC 1995. National Buildine Code of Canada. Associate Committee on the NationalBuilding Code, National Research Council of Canada: Ottawa ON

2. Allen, D.E. 1975. Limit States Design - A Probabilistic Snrdy. Canadian Journal of CivilEnsineerine 2 (1) 36-49

3. Allen, D.E. 1969. Safety Factors for Stress Reversal. International Association forBridee and Structural Eneineering Publication 29-II 19-27

4. MKM Engineering Consultants 1993. Ultimate Limit States Load Combinations. LoadFactors and Resistance Factors for the Desien of the Northumberland Strait FixedCrossine Report to SCI Ltd.

5. Allen, D.E. 1992. Canadian Highway Bridge Evaluation: Reliability Index. Canadian

Journal of Civil Eneineering l9 (6) 987-991

6: Kennedy, D.J.L. l99l . Tareet Values of the Reliability Index Report to ISO TechnicalCommittee 167 SCl, Document N 259E

7 . Kennedy, D.J.L., and Gad Aly, M. 1980. Limit states design of steel structures -

performance factors. Canadian Journal of Civil Engineerins 7 (l) 45-77

8. CSA 1994. CSA Standard S 16. I Limit States Desien of Steel Structures. Canadian

Standards Association, Rexdale ON

17

9. csA lgg2. CSA Standard G40.20 General Requirements for Rolled or welded Structr:ral

oualitv steel canadian standards Association, Rexdale oNASTM 1993

Materials, PhiladelPhia PA

Galambos, T.V., and Ravindrq M.K. 1973. Tentative load and resistance factor desigrr

criteria foi steel buildines. Research Report No. l8 S nuctural Division, Civil and

ffiering Deparrnent, washington university, st. Louis, Mo

Birkemoe, P.C.1976.publication No. 76-09 Departrrent of Civil Engineering University of Toronto

sherman, D.R. 1974. Tentative çriteria for strucnral applicati

gipe. American Iron and Steel Institute, Washington DC

tJ.e., and Kennedy, D.J.L. 1994. The flen¡ral behaviour of concrete-filled hollow

stn¡ctural sections. Canadian Journal of Civil Eneineerine 2l (l) I I l-130

Gagnon, D.p., and I<.oo.¿y, O.J.L. 1989. Behaviour and ultimate tensile strength of

p"rriul joint penetration gróóve welds. Canadian Journal of Civil Engineerine 16 (3) 384-

399Fisher, J.W., Galambos, T.V., Kulalq G.L. and Ravindr4 M. 1978. Load and resistance

factor design for conneótions- ASCE Journal of the Structr¡ral Division 104 (sT9) 1427'

t441Spraragen, w., and claussen, G.E. 1942. Static tests of fîllet and plug welds - a review of

the liteiature from l93Zto January 1, 1940. Welding Journal 2l (4) l6ls -197s

Freeman, F.R. 1932. Strength of arc-welded joints. Weldine Jourqal I I (6) 16'24

Butler, L.J., and Kula¡, C.L. ßlt Strength of fillet weld as a funcúon of direction of

loading. Weldins Journal 50 (5) 23ls'234sCISC,-tg . Canadian Institute of Steel Consiruction,

Willowdale ONMiazg4G.S., and Kennedy, D.J.L. 1989. Behaviour of fillet welds as a function of the

ande of loading. Canadian Journal of Civil Eneineerine l6 (4) 583 - 599

Lesik, D.F., anã fenneJy, D.J.L. 1990. Ultimate strength of frlled welded connections

Ioaded in plane. canadian Journal of civil Eneineerine 17 (l) 55'67Jiang, Y. 1995. M. Eng. Thesis,

Oepã4ment of Ciuit "nd

Environmental Engineering, Carleton University. Ottawa ON

10.

ll.

t2.

13.

t4.

15.

16.

t7.

21.

23.

18.

19.

20.

22.

American SocietY for Testing and

18

DESIGN RTJLES KEY TO COMPETITTVE TIJBI,JLAR STRUCTURES

R. M. Bent*

ABSTRACT

Despite a host of superior properties, many structural designers and.fabricators shunned thegeneral use of Hollow Structural Sections (HSS) in the arly 1970's. Although the fundamentalengineering guidelines were relatively straight forwa¡d, and have remained so for over 25 years,HSS designs were often uneconomical when compared to conventional structures. Moreover,frnished products were not always pteasing to the eye ... some were aestheticly zgly since theconnections were particularly bad.

This early disillusionment left HSS with a stigma. Accordingly, architects became the primeusers of HSS, teki¡g advantage of the fine aesthetic qualities. Major fabrications, m¿rny waryof past experiences, continued to use traditional shapes unless othenryise instructed by the client.A further impediment in Canada - still not quite fully remedied - was the lack of a single, all-inciusive, universally accessible HSS Design Standard. Much useful information was scatteredthroughout various Standa¡ds or squirrelled away in obscure technical libra¡ies.

Stelco Inc., an active member of CIDECT, published perhaps the most useful andcomprehensive set of HSS design guidelines in Canada until 1982. However, having a limiteddistribution precluded these f,rne manuels from having a major impact. As Engineers andFabricators gained experience with HSS, competitive tubula¡ stn¡ctures soon became a realityin the construction markeþlace. Designers had to reverse their traditional mindset of mínímumweight. . . HSS demanded a much tougher target.

INTRODUCTION

GeneralNotwithsta¡ding the early problems, appropriately designcd HSS structures can, and have,proven to be competitive in the markeçlace. Not surprisingly, the role of the structuralengineer has proven to be the deciding factor; his choice of member sizes and joint orientationpredetermines both the quality and economy of the final weldment. Designers also mustappreciate that (1), not all structures lend themselves to HSS, and (2), simple substitution ofequivalent HSS in lieu of existing shapes seldom succeeds.

Of the numerous factors that the structural designer needs to be cognizant, the following areespecially signifi cant:

* Senior welding Engineer, welding Institute of canada, oakville, ontario

19

J

(1)

Q)(3)(4)(5)

Inherent advantages of using HSSDesign guidelines for both members and jointsDesign pitfallsCompetitive truss designsDesign references

(l)

Q)

INHERENT ADVATYTAGES OF USING HSS

Structural designers must take advantage of the inherent properties of HSS.

Strength Sections made to CSA G40.zl have a yield strength of 50 Ksi (350 Mpa).Thus, for satically designed structures, members can be designed for an allowableworking súess of 30 Ksi; for aSTM 436, the equivalent allowable is 22 Ksi.

Torsional Resistance Being closed sections, HSS offer excellent resistance to torsionalforces. Similarly, sections have favouable H/r slenderness ratios, making excellentcompression members, especially bracing members. Also, a significantly longerunsupported length can be used for beams. These same properties give added stiffrressto fabricated units, facilitating field erection. For example, it is not unusual to see 50ft. pedestrian wallovay trusses being brought to the construction site in one piece.

Reduced Slendemess Ratio Research has shown that the calculated values for kl/r maybe further reduced in truss chord and web compression members. When combined withitems (l) A (2), the load+o-weight ratio can be exceptional.

Corrosion Resistance Being hollow, corrosion takes place only on the outside surface.Likewise, only the exterior surface need be painted. The rounded edges promote a clea¡rsurface.

Fewer Gussets For many welded connections, gusset plates are not required.

Aesthetic Oualities Given the smooth lines of HSS, and the elimination of most gussetplates, aesthetically pleasing designs can be produced for a growing number of industrial,commercial, ild domestic uses. Combined with high strength, HSS is particularlyfavoured by architects.

(3)

(4)

(5)

(6)

DESIGN GI,'IDELIIYES

Philosophy of HSS Connec{ionsThe concept for obtaining an optimum economic design for HSS fabrications is zof based onminimum weight, the benchmark used so effectively for fabrications from conventional sections(Tees, Wide Flanges, Plate, etc.). With HSS the objectives are (l), to simplify the jointconfiguration, and (2), to maximize the joint strength . . . minimum weight is not the primegoal.

20

The strength of a welded connection benveenunreinforced HSS members is often a

function of geometric parameters of thesections being joined (the relative dimensionsand wall thicknesses). The profile of theintersection between a branch member and amain member passes along a path of varyinglocal stiffness in the main member. Simplystated, one must ñot forget that thesemembers are hollow, and thus the percentageof the branch load that is ransferred throughthe chord member depends on the degree oflocal joint deformation. Stiffness variationsproduce wide ranges in weld loading. Itfollows that, when "portions" of the weld

transfer little or no load, the strength of the connection is generally less than that of the member,regardless of weld size.

Connection capacity expressed as "connectionresistance" effectively defines the capacity ofHSS members that have unreinforcedconnections. Therefore, designers are advisedto consider the available connection resistancewhen member sizes are being determined.Members selected solely on the basis ofminimum mass may require expensivereinforced connections in order the loads.

The load carrying capacity of an HSS joint isdirectly dependent upon the geometry andconfiguration of the members framing into theconnection. Thus, the designer's choice fora truss diagonals (branch member) must beable to effectively transfer axial loads throughthe chosen chord member. The performanceof the resulting joint is intimately linked toboth members. Unlike many other structures,the fabricator may have little or no

opportunity for substitution when dealing with HSS. Simply using an alternate HSS memberwith similar load carrying capacity does not ensure the integrity of the joint strength.t

The load carrying capacities for various combinations of chord and wallgeometries, as tabulated in the Stelco Inc. Design Manual of 1982, should notbe used today, except for estimating initial sections. Safety may be at risk.

Figure 1: Reduced neffectiven length for comprrssionchord and web members.

KIH = 0.751¡f

Klp = 0.91p

l^ |

K JOINTS

Fignrre 23 Typical load failure, web to chordface.

21

Simply stated, the web force can cause a localized failure in chord walls, particularly in theupper face, somewhat akin to the "high heel phenomena". The flexibility of the tubular wallsgive rise to such failure mechanisms ÍN excessive deformation, punching shear, plasticity, andbuckling (Figure 1). Not unexpectantly, the degree of load transfer across the joint is criticallyimpaired. In other words, the strength of the joint will be less than the súength of the webmember.An equally important observation, the welding may not be a factor in overall performance.Many HSS design principles of the late 1960's for attaining optimum connections still apply,particularly the simple rules relating to member geometry and configuration. Resea¡ch a¡oundthe world, much of it under the auqpices of the International Institute of lVelding (Iltil), has

better deñned the va¡iability of stress transfer benveen web and chord members. In particular,Professor J.Packer at the University of gauge tests on fulI scale models for à variety of differentjoint configuration, i.e., "Nn and "Kn.

One tangible result has been significantly increased joint resisances, thereby allowing greaterload transfers from branch members. Although the formulae and graphical design charts a¡emore complex, current design calculiations also have greater reliability over the qpectrum ofinfinite joint conñgurations.

FTJNDAI\{EIYTAL DF^SIGN RTJLES

MembersThe design of individual HSS members differs little from conventional practice. One still usesthe appropriate CSA 516.l criteria for tension, compression, bending, and allowable stress.However, the effective design length for HSS truss members in compression can be reduced,thereby increasing the allowable compressive load. For continuous chords, use 0.9 kllr; forwebs, use a 0.75 factor (Figure 2).

.IointsThe following guidelines a¡e consistent with CSA 516.1 criteria. Choosing web and chordmembers having compatible geometries will result in joints having:

High joint efficiency (they will carry larger loads)Simple preparations and fit-up (no gussets or stiffener plates)Accessible fillet welds

The net result should be a high-quality, economical design that is competitive with traditionalfabrications. The designer, however, will ultimately determine the outcome. If the workreaches the shop floor with overly complicated joints, its too late for the welders and fitters torectify the situation.

However, the old "load tables" shown here in Appendix A do illustrate thedramatic effect of geometry on HSS joint resistance.

(l)(2)(3)

22

Q)

Some Basic Rules

(l) Connection capacity increases as thewidth of the joining HSS members(web and chord) approach the samevalues. Unfortunately, costs increasewhen welds are placed on the cornerradius of the chords and the finalquality may well depend on the skillof the individual welder (Figure 3).A poor fit-up may necessitate a

backing ba¡ inside the tube, aparticularly diff,rcult task at largeradius corners.

Choose a web that is narrower thanthe chord by at least 5 times the chordwall thickness. This small adjustmentwill provide enough space to use asimple fillet weld around the fullcircumference.

Gap connections are preferable tooverlap connections because themembers are easier to prepare, fit,and weld. A gap joint facilitates theuse of a simple fìllet weld a¡ound theHSS periphery, provided that there issufficient clearance between adjacentmembers. The recommended clea¡distance between "toes" is four timesthe average web wall thickness, butnot Iess than 16mm.

Gap joints usually result ineccentricity and secondary bending.However, these effects can be

Wllh rmrll r¡cllurcom.rt. rulttDlad.t¡ll torlongtludlnrlgroova wald

Figure 3: tfarinun ef f icieacy isexpensive aud difficult weld.

Figure 4i Gap joints are usuallythe most economical.

Chord members with thick walls offer greirter joint efficiency. Efficiency is furtherincreased when a thin walled web member is used. Thus, the designer should maximizethe ratio of:

Chord wall thickness / lVeb wall thickness

Also, thin web walls require smaller fillet welds for a full strength joint, another tangiblesaving.

(3)

(4)

g = 16 minimum

r00x100x8

23

dismissed in joint design if the intersection of the centre lines of the web members lies

within the following range measured from the centre line of the chord: 25Vo of the chord

depth towa¡ds the outside of the truss, and 55To of chord depth towards the inside of the

truss (Figure 4).

-0.55 < e/h" <0.25

(5) If a given lap joint does not provide .

adequaæ efficiency, then either (l)change member parameters to achieve

a stronger gap joint, or Q) change theconnection to a lap joint with at least

25Vo overlap (Figure 5).

With a lap joint the forces aretransferred directly between the webmembers, thereby eliminating localchord wall failures. Consequently,lap joints have both a higher static and

fatigue life than gap joints. However,lap joints require two preparation cutsand a tighter fit-up, both cost-adding features. To simplify fit-up, place the narrower

tension member onto the wider compression member. AIso, the bottom inside a¡ea need

not be welded (Figure 6).

100x100x8

Figfure 5: Use ]$$rmirnitnrmoverlap joints if agapjoint will not wor*.

Given the almost infinite number ofcombinations of member size,periphery, wall thickness, and

orientation, alternative gap jointdesigns with equal or higherefficiencies are readily attained at thedesign stage.

(6) In web members that are inclined tothe horizontal by 60 degrees or more,the welds can be classified as fillets(Figure 7).

It should be readily apparent that Figrre 63T0€ of overlapped member is not

designing in HSS is very much atríal- welded.

and-error process. However, themethodology is straight forward and designers will readily discover the great versatility

of tubular sections. The same unit weight can be attained by a multitude of available

sections.

24

(l)

POTEI'ITIAL PITFALLS

The number of potential pitfails can be infinite; however, there are a

discussion.

few that merit sPecial

Eigrure 7: The angle of HSS mernber is

ir nportant design feature.

A "f"rk" roof truss - not well suited for

er exists, provide a "drain"

HSS Redesign First, do not redesign

¿rn existing structure bY merely

substituting HSS members of equal

load carrying capacity. The results

are not likely to win many accolades,

as the fabrication costs may set new

records. For examPle, when a "Fink"tnrss (Figure 8) was redesigned in

HSS some Years ago @Y the author),

the number of different sizes, lengths,

peripherals, web orientations, laP

joints, etc. was indeed a Pooradvertisement for mY comPanY's

product. The lesson learned, ofcourse, was that HSS structures must

be designed from "scratch" to take

full adva¡tage of the inherent

qualities.

Ice Damaee While studies have

shown that there is minimal chance

for corrosion on the interior surface oftubula¡ structures, it is usually prudent

to seal or cap all open connections. Ifwater gets inside a tube (during the

erection perid or if exPosed to

elements while in service), the

damage wrought bY the freezing ofeven a small amount of water can be

quite depressing (Figure 9).Whereuer the potential for such a disasl

0r", = 120o

(2)

tigure E:

flss.

(3) Minimum Weigùt Competitive HSS constn¡ction is driven by optimizing the geometry

of the .onnõñ* and úy simplifying the fabrication process. Having satisf,red this

criteria, there is little to Ue gaiìø by attempting to shave a few pounds off the total

weight. Use as few sections as possibie - this will standardize production. For example,

for a group of web members, use the same section whenever possible; to procure a

virtual kaleidoscope of members having a different width, depth, or thickness solely for

the purpose of reducing wight would negate purpose of using HSS' The extra handling

and tracking problems would definitely increase costs.

25

(4) Gussets & Stiffeners With a little manipulation of HSS sizes, the designer should beable to eliminate gusset plates and stiffeners (Figure 10). These items add extra materialand cost. However, such chord member reinforcement provides excellent results whenfabricated and welded according to the empirical methods developed by Korol et al(1982).

(5) lVeld Efficiency The angle of the web to the chord directly affects the efficiency of theweld. For angles where 120" > 0 > 60o, use simple f,rllet welds all around the outside.For angles where 6ü > 0 < 3V, the weld on the heel must be considered a PJFG weld.For angles less than 30 degrees, the heel weld is not considered to be effective inresisting the applied member force.

COMPETITTVE IR,USS DESIGN

The "Wa¡ren" tn¡ss (Figure 1l) is particularly wellsuited to take full advantage of HSS. Such designs,¿rs outlined below, have consistently provedcompetitive in the market place; where tenders forboth an HSS and an equivalent, traditional design(Tees & angles) have been called, the HSS has beenthe clear winner.

IISS lVarren TrussThe following criteria result in high quality,economically competitive HSS truss designs. Thesame criteria will also result in higher jointresistances and load carrying capacities.

o web members having the same single cut endpreparation (say 60)

o continuous, parallel chordso gap jointso fillet weldso design based on F, : 50 Ksio reduced "kl/rn ratio for compression members

(0.9 for chords, 0.75 for webs)

Figure 9: FrozED satercracks aud ðeforus EBgcolr¡¡¡.

o high ratio of web-to-chord width (no weld on corner radius)o thick chord wallso high ratio of chord wall thickness-to-web wall thicknesso member sizes kept to a minimum

26

h6d",l"f"rma-tion for the design of HSS

HSS DESIGN REFEREI{CES

Figure 1o:chorô vorlrs

\. ./^

t*.*ttt is still difficult to find: references

i" scattered among different codes'

It o¿.tát, countries, organizadons' and

ä"t"J publications' There is no single'

authoritative source of useful data' This

information vacuum represents a serious

hurdle to designers, and may partially explain

their reluctance to use HSS' There is no

frovision for fatigue design in the current

Canadian Standards.

Reinforced Bgawel1.

truss

(2)

(1) CIDECT . ^ ---.:a^^ r^- +ha cnrrr., en¡l f)eveloDment of TubularCIDECT(fhelnternationalCommitteefortheStudyan.dDweJorStructures), a major sponsor for research, was

"tpontiutt for much of the early design

material. Howevef, its work is not readily ut"t"ibl" to most Canadian engineers'

Stelco Inc.ñtit -tggZ, Stetco assimilated the

CIDECT research and disseminated

the knowledge in several company

Design Manuals. At the time, these

tanutl, were a Primary source of

HSS design information in Canada'

UnfortunitelY, being a commercial

oublication, these excellent manuals

ïere not widely distributed and many

designers were unaware of their

existence.

IIlVftt. UW (International Institute offorefront on HSS reseatch, with

recommendations. CoPies of IIWInstitute of Canada.

figure 11: A nWarrenn trr¡ss is well suited

HSS construction.

(3)Welding) Subcommiftee on HSS is now at the

an increasing number of Codes based on its

Oo.ut"n,t ãray Ue obtained from the Welding

(4)ffio"ralsteelweldingcodeissimila¡toCSAw59,theCanadianweldingdesigncode. Section 10 of nWSbt .çg4 and earlier is devoted solely to HSS' Two groups

of shaPes are covered:

o squares & rectangles (conform to IIW)

. circula¡ shapes (apply only to offshore)

27

(s) CISC Handbook of Steel Construction. FÏfth EditionSection 3.0 þage 3-83) covers HSS connections, having numerous diagrams and anexcellent summary of design parameters (based on IIW"). Tpical welding details are alsoprovided. The required length of weld for different wall thickness and periphery at givenangles is provided in tabular form. However, the Handbook is not a Code, and rnanydesigners are unaware of the information on tubular steels.

CSA Standard tV59As with CSA Standard S16.1M for stn¡ctural design, the tJ/59 welding code has nosection qpecifica[y devoted to tubular steel. However, the next edition, CSA W59-1966(this year) will for the fi¡st time have a separate section on statically loaded HSS.

CISC Pr¡blicationJeff Packer and Ted Henderson's Design Guídc for Hollow $nrcnral SectionComccrtons provides engineers with a practical and comprehensive 'state-of-the-art" text.The design examples conform to CAN/CSA 516.1-M89. This book covers the commonand the noþsæommon. It is consistent with IIW, but writæn to suit the Canadiancontext.

CI,OSING REMARKS

HSS offers the designer an alternative that, when appropriaæly designed and applied, produceswelded structures with high strength, quality, aesthetics, economy, and proven in-serviceperformance for many applications. Although major producers such as Stelco Inc. havesuccessfi¡lly used this product in a wide variety of applications for many years (roof tnrsses andbracing; pedestrian walkrvays and handrail; bridges and towers; conveyor supports in corrosiveenvironment; lighting standards), HSS has not yet received the same acceptance throughoutCanada.

Some early designs proved to be costly, and the anticipated aesthetics did not matchexpectations. One can list numerous factors for these shortcomings, but I believe that thefollowing were the most significant:

(l) Although guidelines were available, they were not published in a single CSA approveddocument. Therefore, they have not been readily accessible to all designers.

(2) HSS structures that are not built to appropriate guidelines can quickly alter a good designconcept into a fabricator's nightmare.

(3) The fabricating industry had liule practic¿l experience with HSS, often learning by trialand error.

(6)

(7)

28

Hopefully this paper h¿s been able ro porrray the design of HSS in a more favourable light.

ettî,oogt the design tools may still be Jomewhat widely dispersered: op"914l' in Canada' the

,o4or rif"r"n"r, liave been cláarly identifred and can be readily obt¿ined. HSS aesthetics, high

strength-to-weight ratio, and vast range of geometries have Proven to be competive over a wide

range of service applications. Along with the growing emphasis on co-ordinaæd global research,

the fuure bodes well for tubular stn¡cn¡res'

The designer is the key player in successful HSS constn¡ction. To emPhasil thit point one last

,i-.-the-designer's iniùar-decisions can make or break a project. Thus, the need for proper

raining and education is underscored'

As a f,rnal reminder to designers when using HSS:

1.

,,Sel¿ct nemben and evøhutc joínt effæicncy sinalt¿neously."

REFERENCES

cran J. A.; Gibson E.B.; Stadnyckyl s. 198l,Znded. Hollow Stn¡ctural sections, Design

Manual for Connecdons; Sælco Inc'packer, J.A.; Wardenier, J; Kurobane, Y; Duna, D.; Yeomans, N. 1992. Desien Guide For

4.

5.

CIDECT, Germany. ISBN 3-8249-0089-0

J. Fraær,G.S.; Packer, J-4. 1990. Desisn of Fillet W

;ä;; -ðn;tcf REpoRr ffio¡727;s70-2. universitv or

Toronto.packer, J.A.; Henderson, J.E. Igg2. Desien Guidg-for HouoY SjrycJural Section

ðã*.",ionr..cISc. ISBN 0-ggg11476-6. universal offset Limited, Martham' onta¡io.

Koral, R.M.; Mirri, H.; Mtrzu F.A. 1982. Plaæ Reinforced square HgUo* Section T-

ioino of Une4ual Width, Canadian Journal of Civil Engineering, Vol.9, No.2, pp' 143-

148.

Inærnational Insúrute of rwelding subcommission XV-E, Design Recommendadons for

Hollow stn¡ctral Joints - nedominantly statically Læaded,2nd ed., Irw Doc' xv-701-

89.Cmn, J.A.; 1982 WusineRe"tanzult-õhoi-ilM"mbels=rf!it4^P:tl:2^2: jæt3o]n¡'

(Final Draft) Part D' PP.22'29-òlSC Handbook of Sreel Constn¡cúon, Fifth Edition, 1993.

Aws D1.1-1994 Stn¡ctural welding code - sæel, section 10

CSA Standard S16.l-92 Limit Staæs Design of Sæel Structures

6.

7.

8.

9.i0.11.

29

APPE¡TDX A

Gap Joints - Maximum Allowable Venical Component of Force IV, in a Web Member (krps)

Rectangular Chords and round or box webs.

These joint efficiency Tables were based on early resea¡ch by W. Eastwood and A.A. Wood in1970, University of Sheffield, England.

30

Table 4.3-1 gives theralues of working

toã¿ (wu) foi tne maioritY of cæes

;är;;; in desisn. BY enterine.the

ä* *'ra;e chorã width and wall

ìü"*"ttì rJt.nd T)' and the averaç web

.. ¡6,+d3),member wrdm '--

the allowable vertical component of force

ì*iltl ," theweb svstem is obtained' lf the

ìt l"",u.r vertical component of f orce tn

" ;ö;;.;;mber is sreater than

-tn1;"j;;,; *"io s"P is not acceptable' since

"r"ä*"'o.tolrmation would occur in the

"ittJi..o"r face at ultimate load' An

ä"ãtoo ìãt", (section 4'3'2) st¡ould then

be considered'

Fy = 5o ks¡

TaHe 4.3.1

Gao Joints - maximum allowable venical component of force (Wv) in a web member (kips)

äãi;;ì;t chords and round web members

3.003.003.003.00

0.15000.18750.25000.3125

0.15000.18750.2500o.3125

0.18750.25000.31250.37500.4375

3.503.503.503.50

4.004.004.004.004.00

5.005.005.005.005.005.00

28.2935.3647.1458.93

28.1337.5046.8856.2565.63

18.05

36.5648.7560.9473.1387.75

0.18800.25000.31200.37500.43750.4500

i 24.00i zg.gsI so.ooI ¿z.oo

I as.zo

I

I 11.28

24.8233.0041.1849.5057.7559.40

37.6050.0062.5075.0087.50g0,oo

6.O06.006.006.006.006.006.00

0.18800.25000.31200.37500.43750.45000.5000

11.2815.0018.7222.5026.2527.OO

30.00

11.2815.0018.7222.5026.2527.OO

30.00

11.2815.0018.7222.5026.2527.OO

30.00

i 15.00I re.zzI zz.soI za-zsI zz.æì so.oo

16.9222.5028.0833.7539.3840.5045.00

28.2037.5046.8056.2565.6367.5075.00

33.8445.0056.1667.5078.7581.0090.00

I""r.n. Diameter of web ttt6sr (in')

31.5842.æ52.4263.0073.5075.60

22.5630.0037.445.0052.5054.0060.00

RESISTANCE TABLFÆ FOR WELDED HOLLOTV STRUCTURAL SECTIONTRUSS CONNECTIONS

J.A. Packer', G.S. Frater* and S. Kitipornchait

ABSTRACT

In recent years recommendations for the design of planar, welded, Hollow Stn¡ctural Section

(HSS), truss-q¡pe connections have appeared in a number of 'structural steelwork

specifications or design guides around the world. These recommendations have been in the

form of extensive sets of formulae for each connection shape, with limits of validiry attached,

and occasionally with graphs showing the influences of some principal paramaen. Toengineers unfamiliar with the jargon, failure modes and nomenclature, designing with HSS

often has the appearance of being formidable and the potential for error. This paper aims to

ameliorate those concerns by tabulating the limit states (LRFD) resistances of several

connection shapes in many popular member sizes. Designers will be able to gain confidence

by checking their calculations, perform approximate interpolations for other member

combinations, and accelerate the selection of members.

KEYWORDS

Hollow Structural Sections, structural steel, tubes, connections, joints, trusses, design aids,

resistance tables, LRFD, limit states design

INTRODUCTION

One of the most popular applications for Hollow Structural Sections (HSS) is in truss

construction. Unlike structural design with open sections where it is easy to provide

stiffeners at critical points to strengthen connections, the closed section of an HSS is best leftunstiffened - whenever possible - at a connection. This produces a very clean and

aesthetically-pleasing appearance as well as a low-cost connection too. However, this entails

proper selection of the HSS members and performing the connection design at the member

selection stage. Thus, for HSS construction connection design should be performed by the

structural "ngin""r

rather rhan the fabricator. This is not a difficult task, as very detailed

design guidance is now available from the Canadían Institute of Steel Construction (CISC)

tnef. f i and elsewhere (Ref. 2). This CISC Guide has been used to generate the connection

resistance tables presented herein, which are directly applicable to either the Canadian limitstares steel design specification (Ref. 3) or the American LRFD steel specification (Ref. a)-

-D"p*.""*f Ct"tl Engineering, University ot Toronto. 35 St. George Sr, Toronto, Onta¡io M5S lA4, Canada

+Hatch Associares Ltd., 2800 Speakman Drive, Sheridan Science and Technology Park, Mississauga. On¡ario

L5K 2R7, Canada#Depanment of Civil Engineering, University of Queensland, Brisbane, Queensland 4072, Australia

32

HSS TRUSS CONNECTIONS

Some general tips that designers should- bear in mind in order to maximize the strength of an

HSS to HSS welded connection ¿re as follows:

.chords (or ,'through members") should generally have thick walls rather than thin walls

.web members (or ,,branch members") should háve thin walls rather than thick walls

.web members should be as wide as possible relative to the chord member' However' this is

offset b1, the fact that HSS web members should not be the same width as rsctangular HSS

chord members, (except in Vierendeel trusses), as this presents an awkward fla¡e-bevel weld

situation (possibly wiih backing bars) for tne joint at the corner of the chord section' A

preferred afrangement is just su-fficiently n*o,i", than the chord to permit the web member

and some of the frllet wðl¿ to sit on the "flat" of the rectangular HS-S- chord. member' The

outside corner radius of a North American cold-formed rectangular HSS member is generally

taken as rç,o úmes the wall thickness (r), alrhough the csA siandard (Ref. 5) allows over 3r

for some thicknesses.

The factored resistances of some popular, standard, welded truss connections are given in

Tables I to 12. Th¡ee connection shapes a¡e covered: 90o T connections' K gap connections

and K l0o7c overtuf "onn..tions,

wiìh the members subject to predominantly axial loads'

These three conneclion configurations have resistances tabulated for popular HSS

combinations. for square-to-square members and round-to-round members' The tables are

arso produced both in m.rric änd imperiar versions to facilitate design with either system of

units. The steel grade assumed in these tables has a guaranteed minimum yield strength of

350MPaor50ksi,andcanbeeithercold-formedorcold-formed'stress-relieved'Thesection sizes shown ^r.

no, an exhaustive list of atl available' but merely represent the ones

more coÍrmonly used. Further sizes available in canada are given in Refs' I and 6'

As noted previously, Tables I t9 12 are for use directly in conjunction with either the

canadia¡ limir states desi-sn specification (Ref. 3) or the American Lnrp specification (Ref'

4). No additional resistan-ce (ô) factors n.à¿ b. added. If Allowable Stress Design (ASD) is

used, a connection allowable load can be obtained by dividing the connection facto¡ed

resistance by 1.5. The K connections are for a specific web member angle (45") and a

particulü gap size tgi o, amount of overlap (O"), whereas in practice a huge number of

possible parameter .àäUinut¡ons is possible- ïn t. tables, however' will enable the designer:

(i) ro get a veñ' quick estimate of a connection factored resistance' even for a

slightiy differeni connection' and

(ii)toconltrmthatmanual'orcomputer-codedcalculationsforconnectionresistance formulae are being performed correctly'

For truss-type connections a structurur O.Jign.r can use these- tables to select HSS members

astutell,andtherebvavoidanysubsequentneedfor.connectionreinforcement.

Blank sPaces in these tables indicate that either:

(i) a particular combination of members is outside the range of validity of the

design formulae available' or

(ii,¡theconnectionisimpracrical(forexamplewebwidthgreaterthanchord

33

width), or(iii) the connection is not recommended (for example web member widths equal to

the chord member width, for square HSS connections).

Where such bfank spaces arise the combination of members may still be possible, and

recourse to the CISC Guide (Ref. l) is recommended for more detailed and definitiveguidance. In some tables, for example those for K gap connections, one should realize thæ

the specification of a particular parameter size (such as I = 30 mm) has severely restricæd

the number of possible connection combinations.

In Tables I to 12 most symbols are defined in the accompanying connection illustrations.

The subscript 0 refers to the chord (or "through") member, the subscript I refers to the web

member in a T connection or the compression web member in a K connection, and the

subscript 2 refers to the tension web member in a K connection. In overlapped connections

the subscript i is used to denote the overlepglng web member (usually the smaller or thinner

web member) and the subscript j is used to denote the web member which is overlgppgg[.

The factored connection resistances tabulated usually need to be reduced by a conection

factor,fln) or fln'), if the chord member is loaded in compression. where:

For Round HSS: f(n') =

For Square HSS: f(n) =

l+0.3n'-0.3n'2,and

1.3 + [O. bo lb,ln , but not greater than 1.0.

For axial comprcssion load in the chord, n and n' will be negative numbers. n is the axialforce in the chord (the larger for either side of the connection) divided by the chord member

squash load (area times yield strength). n' is the additional axial force in a truss chord at apanel point, other than that required to maintain equilibrium with web member forces (or the

"prestress force"), divided by the chord member squash load.

REFERENCES

l. Packer, J.A.; and Henderson, J.E. 1992. Desisn euide for hollow structural section

connections. lst. ed., Canadian Institute of Steel Construction, Willowdale, Ontario,

Canada.

2. Packer, J.A.; and Kitipornchai, S. 1996. Guide to the hollow structural section guides

and codes. Proc. International Conference on Tubular Structures. Vancouver, 8.C.,Canada.

Canadian Standards Association. 1994. Limit states design of steel structures.

CAN/CSA-S 16. l-94, CSA, Rexdale, Ontario, Canada.

American Institute of Steel Construction. 1993. Load and resistance factor desisn

specification for structural steel buildines. AISC, Chicago, Illinois, U.S.A.

Canadian Standards Association. 1992. General requirements for rolled or welded

structural qualitv steel. CAN/CSA-G40.20-M92, CSA, Rexdale, Ontario, Canada.

Canadian Standards Association. 1992. Metric dimensions of structural steel shaoes

and hollow structural sections. CAN/CSA-G312.3-M92, CSA, Rexdale, Ontario,

Canada.

3.

4.

5.

6.

34

T.CanadianStanda¡dsAssociation.|gg2.Structuraloualitvsteels.cAN/cSA-G4o.2I-M92, CSA, Rexdale, Ontario' Canada'

ACKNOWLEDGEMENTS

Financial support for the development of the "pre-engineered" connections presented herein

has been provided by lpsco Inc., of_Regin", s"rlocnJwan' Canada, the Natural Sciences and

Engineering Resea¡ch ^Council of Canada (NSERC)' and the Australian Institute of Steel

Construction-

'i,i

I

35

Table 1: T Connections Between Clrcular HSS Memäg,rc

steet Gnde: 35Ow (Accor(ting to cAll/csA G¿'O'n/402''Mg2)

Factor€d Connsslion Ræislancos (Nr') in kN forWeb wnlth (dt in mm) ol:Chord

do (mm) ro(mn) 60 89 t1¡t t68 219 2'r3 3,21 ¿t06 508 610

60 3.2 f¡.

60 3.8 t31

,r{

4i

t€

60 rl-B r&t

60 8.4 2ßæ 3.8 78 t4l

æ 4.8 117 212

89 6..1 t9t 331 r

89 8.0 291 ¡149

114 ¡1,8 89 r50 z3

114 .6.4 1¡l8 250 3'n I

'll¡l 8.0 22. 375 558

r68 4.9 66 96 132 241

r68 6.¡l 110 r60 21 402

168 8.0 t64 2& gl1 603

r68 9.5 27 3I3 4!i9 8:X¡

219 4.8 58 77 flg 167 zil

219 6..1 97 128 f66 278 121

219 8.0 145 192 2ß 417 dts

2r9 9.5 201 266 u 578 881

2r9 tt 2U 350 ¡153 760 1f60

273t 6-¡l 9t 112 138 2r3 311 43

2î3 8.0 136 168 206 319 ¿166 664

27.3 9.5 1Ãt æ3 286 43 6¡16 9ã)

27¡3 fi 219 97 376 58¡t 850 1210

273 r3 317 391 179 742 1080 t34{'

91 8.0 156 r84 268 375 521 6t8l

æ1 9.5 217 É5 371 5æ 72 952

91 11 285 336 ¡188 68tt 950 1zfi

p1 r3 3dt 128 62 8'n 1210 tÊ00

4{¡6 9.5 æ1 m 307 ¡106 5¡lO 69¡l 996

¿l0o tt 268 302 4Gt 5Ít5 711 913 r3to

¿t06 t3 g2 i 385 514 681 906 r160 1670

5{)8 It 2U 35t ¡139 557 69Íl 957 r370

5()8 13 361 47 s59 709 881 1irlo I r75o

6r0 r3 ¿113 4f¡¡l ñ2 726 969 r350 r8t0

CORRECÎON FACTORS: 1. ll lh€io ¡s corflptæsivo þ¡d in ctÛd' munity Þy teduclion lacþ' lln'l2.1'thewoÞm€mb€'É¡ncoíp'æsþn'th€ñarifîump€fmilledconß€clþn'as¡sbnce..vr..islimttedlo:

Fú d,l,tof, 30 35 40 ¿¡5 50

¡v,'-"i.i. trr.¡1, o.sas¡, o'3ogAr o'æ8ilr 0'273ù 0'266Ár

ïñete Ár ¡s lh€ web mgmbgr cross'ssclional arsa in rxn2'

36

!

Ii

ìI

I ',l Hiåru;.is in corn'ress'on'

Ëoiäii,"t' *ii'J,tii""äüi ", x,lrii :,'": ;^','i* ^'

rt0

connecüon":,?Y::#::tï'Y;'J:åtr"

*:#;'1:** äl;"' ""1'åo'øl a'"a in 'n'

nc€s n , < -^^ | 2/t.oo

Fadore'd Conne-ìe.ezs ! t't1 10.75

\

12.7516.0(

Chord

--{lr-so | 1'5ol I

xô (in.) ro (in.) 2.37s

t!.t'to

/h

tr{N

.12321.2

2.374

2.371

2.37

.r50 29.1

.r88 l

.250

.150

.188

t4

¡¡1.1

sa.7I z.sts

3.50

, 350

17.1

ú.2

31.7

47.6 f þ#j13.779.6 I

3.65.5

19.9

101r3 -f

50. ----r-i

-1¿FI3.5{t I

----'a'4.50 ï

3ß.õ

56.133.3

49.8¿.æ I zsc

¡so I 31:8¡1.0

21.5

125

29.8s4.1

90.4

1366.625

r88 1419.7

24.6t3õIls:

I

6.625-/50

I .313

I .375L--I 1n8

71.5

57.1

I e5.3

36.9 :-)-.^ I 1o3188

-

i

-L---6.625,o 37.5

,s.o\ ¡z\ =r 62.76.r

8.

8.

21.7 2E'u i

t6.0 sg'91rzs i .250

625 i '313

.625' '37'

.oes i 'ß

--a3,2.6 130

198

45.1 ?62

70.0103

17299.6

'14959.6

25.231.0

48.0

20.5 I r05

lo.?5 I30'7

I 37.8

I R23

t45 207

273io.us \ -191964.3

39.6

ñIs ì gzsE

I 6s.2 85.01v167

t'- 947

10-75.438 .--.1-

'r.s i-----j-

ß7.81og q r17 154

r0.75

12.75

.500 41.5 60.ó |

117162

215

ztt4I

.313 I ==-Lg-l 1, 51 5 83.5 283

.375 110 !

r40

68.9

91.1

r16

35912.75 ão . zo'o iq6 272

22412.75

.138 ãr.z i --sj r21

160

203

ì206

296Iz:s \ 'æo ¿5.8 51.6

12',1261

r5ô

376-tîiì lä 60.6 68.3 (11 216 309

39q76.9 86.7 .tâ

600 .500 64.r

81.3

?9.3 99.u +-IFI)-

bY ledt

59 - ræ \ ?15- ï- gog 407

lim¡led to:

.438_Wrdion lactor rj:],^"

resiíañce, N,'. i

20.00

æ.qt93.0.500

rioN2a.00

.5,o0loed in ch

coñProssivo marrißumPermrßi'-"''- SO

CORREGTION FACTORS:

37

38.58 Ar

TabteS:KoverlapConnectionsBetweenCtrcularHsSllemberclO = lAÙloand0, =0o=45o)

steet Grade: ssòvi ¡nccoøing to cewës¡ G4o2o/40'21'M92)

coFREcTlon¡FACToRS:t.llth€'åiscompfÊssiveload¡ncho'd.muniplybyredudionlaclo'/(,J2.Forlhecompfessimlveb'"-o",,,n",',",imump€nnnedconnedionfesßtance'N,..rsl¡mitedlo:

Fot dt!\ ol: 30 35 ¡10 45 50

¡¡i:-"1"i.ì'.Hl' o.34ltAt o'3o8ár 0'298Â! 0'273A1 0-266Át

whele A! is lhe web memÞet cross-sectional area rn mm2'

NOTE: The th¡Ckness ratio between weÞ members ¡s limitod as lolloìys: ¡rl, s t.o. where 7 telefs to lhe oveflapped membet

Chold Factorod connêcüon Ræisances (lV¡' or ¡t/21 in hN lor wob Wdltl (dr ¡n rm) ol:

60 t14 r6a 219 273 æ1 ¡106 508 6rods (mm) ,o (mm) 8f)

60 3.2 124

50 3.8 r68

,260 1.8 239

60 6.4 378

89 3.8 l¡t9 206

8!t 4.E æ8 2æ

89 6.¡l 3ã) ¡l¡11

89 8.0 ¿158 dÐ

114 ¿1.8 196 266 æ7

111 6.4 æ7 ¡l{x¡ ¡196

8.0 ¡tf 9 5€8 699114

r68 ¡1.8 1E2 252 306 &1I

168 6.¿l 2æ 369 u7 615

615 8,15168 8.0 385 fi7

168 9.5 503 663 g)4 1 t10

219 1.8 rqe 255 g)5 413 515

Sdt 19 s88 733219 281

8.0 379 4st 586 793 988219

219 9.5 ¡t89 6¡t1 755 1t20 1270

219 11 eog 761 912 f280

7't9 864273 6.¡l 2C2 36!) ¡tÍ¡6 581

95:¡ rl5{)273

273

8.0 3A7 ¿t8!) 578 771

9.5 491 621 735 980 1210 r46{¡

I 800n3 t1 606 767 908 r2to 15oo

1470 1810 21æ)74 t3 7g 9æ tl(x)

æ1 LO ¿t98

â9ß

767 941 tlg) 1300

7U lts5 r18{) 11æ 1630æ1 9.5

r4f{) 17gt 2000321 11 766 898 11 8{t

94 13 921 1080 t4A0 '1710

1180

2080 2410

r39() r6{X) 1gþ406 9.5 6¡19 751 f¡69

7AE 908 r170 1420 1690 19¿10 7W¡106 11

t080 t400 r700 2010

168()

2æO 27æ¡loo t3 93¡l

1920 2m 2no508 11

13

ft45 rãx) 1430

ro r690 198() 2t) 2710 3270508

172î 20æ n70 itæ I æao i 37æ610 r3 1450

38

i , Eet Grade: sivÜ 4ccord¡ng to cerur

----M,' or N2') in kiPs lot 'WebWrdth (dr in inchBs) or

=z-

-, ôore'd

connecton Resisþnc

8.1

ts ;l--,* I *-i25

1-I to.zs f tt.tt 16

I cnoto I

) (¡n.)

¿.375

z-375

2.375

ro (in.) 2.373 3.50

.125 29.7

I-j-o,, ',Ð*.150 37.7

I2n

53.7 T-85.0

íz.s7s

h

,n

33.6

#I .1 50

188i15.6 ú.2

3.5972.o aq-1

) 3.50

i s.o

2n

73s

112

313 103 142

14.1 59.7¿1.50 .188

.2æ 66.8 90.54.50 1¿8

!19.188

94.168.8

101

9/r.6

6.62543.0

62.9

86.5

fi1138

s2.9

6.625 .2W 138

6.625

6.625

8.625

8.625

.313 181

68.6

97-8

132

249

.375113 149

92.9 116

.188

.2æ

ÁÀ.1 5:1.232 165

63.3 81.5178 ?2

85.3 110 2æ

359

162

8.625 170 2æ110 112

â25 .37s177 ry

98.2

2ú194

8.625 .438 137 'r318¿.9 257

10.75t6ß 65.6 tæl 173 211

86.9 110272 327

10.75 .3131¡lo

æ2

'rt0.37s æ1

21

13

, T-- 273,:,l 33o

337 405

10.75

10.75

I.500

137 173 ¡¡Og 490

165 209;T 212 2ß

1'12 2æ 318 æ7

12.75

12.75

12.75

.313 451

511

35S

3751 ¿t1

zez | 327 391

.438173 ã

z¿s II

reg I

3æ

,18

392469

434207 2ú 313

1215 5æ¿¡fl6

3æ

3æ 6,2516.æ 2e5 518

16.00 .¡¡38210 213 314

379 43:l 519

16.00 .5æ 213 270 323508

iô 735

7æ379849

.438

.500æ.æmm

?51 317 5tl 608

3ú 387 45()

24.æ .@ñutripry byr€-dudion 1"":::#"'"

,esisrance. Nr" is limiled to:

load in dìord'

rabte4:Kovertap'i;:îËr:ìfiÏ^ËÅFt¿:,-:j.:::

CORRECTION FAGTORS: I iiï: :"#'"':ili"r'*''-# ""'lä' î*fr' "'i"i#ir;;" Ë,À.i::å,ii;ïå',"1i"1 ::":['J:*, sa' -"li:å' "ïi;: :" I ::"il;*"ïï,1ÏË'-"'i".i' ";1;;-';i;:1i::ïi:; , re,ers ro rhe overrapp'ed member)

The thickness ratio b€lwe€n weo memUers rs timit€'d as follows:

50

NOTE

39

Tabte 5: K Gap Connections Between Clrcular HSS llemberc(g = 30 mm and9, =02= *5o)

steet Grade: 350W (According to cAî't/csA G402U40-2','M92)

CoRRECÎION FACTOFS: r. ll lhcrr iS Corfþf€ssiv€ load in cfiord, munÞ¡y by tedt clk)n facrot /(¿'). 2. For tho co|rÞræsion uæò ñËmb€f. lhe mâximum perffúned conneclion f€sjsta¡c6. ¡vr" is limiled to:

Fot dtl,1 oa: 30 3:t Q ¿¡5 l)IVr'max. (kN): 0.3¡13Ár 0.308Ár 0.298/tr 0.273At 0.266Ár

whcrc A, is lho wcb tñambst ct6s-s€ctilnal a¡aa in mm2.

40

Chont Factorod Connec{¡on Ræbtancos (JVr' or JVr') h ld{ for Web WttÙt (dr ¡n mm) oft

do(rm) ,o (mm) 80 8Í) l1¡¡ t68 2t9 273 æ1 ¡f{r6 508 6lo

e0 3.2

60 3.8

,r'fi¿60 a.8 ]1à-r60 6.4 -Y,r

0l r)dt 3.8

89 ¡1.8

æ 6.¿l

4¡ 8.0

rlr a.8 142

114 6,4 2a2

tt¿l 8.0 365

168 ¡1.8 121 r60 193

r68 6.¡l 208 271 3íMl

r68 8.0 314 174 fi2r68 9.5 1g s72 693

219 4.8 112 111 t72 2æ

219 6-¡l 'tfx 250 299 ,t{xi

219 E.O 293 378 152 612

2r9 9.5 & 521 624 w219 1f s26 679 812 tlcx)

273 6.4 2æ 279 372 ¡¡60

273 8.O 358 4z,3 56¡l 698

273 9.5 ¿193 5€Ét TN 96t

273 tf 640 757 1010 125o

273 13 804 951 1270 1570

321 8.0 348 ,lO8 sgt 659

321 9.5 479 562 739 900

321 fl 621 728 95€ 1180

321 13 7fa 912 1200 1170

¡t06 9.5 97 706 656 rû20 1170

¿106 11 706 9fl 1110 r3f0 r510

¡106 13 881 tl¿l{l r380 t6¿to 1880

508 t1 890 1070 125() 1¡lilo 1710

50€ 't3 tt10 13ã) r5€o lTfo 2130

610 13 r3t0 r520 17æ 2050 21æ

Taþle 6: K Gap Connections BeWeen Cirlltar HSS lllembers

"^"" Ï-,gi l ^!l:nl

: ¿;:'za\ o o' o' o o "'

n

"

Iti';-:li';*,, 13.10^, T.uro, lä 'u^,. 3e'5eÁr 38'584r

i;.ii: lläi *"0 "-*t cross'secrio¡r¿¡r area in in ''

.is

in kiDs for web Width (dr in ¡nches) ol:

-- Faclored Connection Resislances (Nt'

IChordrl

.,z.ts | '16.00 zo.æ | z4.oo_,l¿.50 ' oszs I 8.6¿:

r. u", I 2's75 3.50do (¡n.) t_2.375 .125

.150 -_|2-375

2-373 .t88

476 .250

.r503.50

g3.50

.188

r.3t3

It-

¿¡.50 r88 31.9

54.¿t4.50 .250

.313 9r-927.2

¡¡-5{¡35.8 439

71.9--râ)E .188

6.625 .250 46.8 61.7 I

I

zo.e i 93.1 1r3

l!1.lÂ 7

6.625 .313

6.625 .375 v:25.1 32.3

52

JI

8.625 18867.3 sl_1

138

I

ii

I

i

II

II

l-250 43.6 56.2

6sJ90.8

99!117

102

1408.625 .313 tr9l

24s I8.625 .375

II

1531f983.8 r03

10.75 .250 53.1 62.9

80.5a5¿ 1?7 157

t---r-----l10.75 .313t31 175 216

37511r irIt

ii10.75

1/¡4 tzr ! 228'Ix¿ | 285-lge.o i rzt-^- I tA.ß

282

*zs I .os8 ssz !

rtozs I .500181

t¡18

12.7s i .grg !::108

2UIt2.75 .375

ros I 216 265

12.73 .438140

33t175 261

12.75 500reg I 159 r92 22L

r6.00 .375 296 3:19

-438160

422r 6.00 310 36

t-16.00 .500?o1 241 2A3 322 387

a7820.00 438 298 : 350 3S9

162 55320.00 500 293 v2 388

.500 i_21.00Þy reducrron laclot /(n') ^.--^- À, . is limited to

41

Table 7: T Connections Between Square HSS Membe¡s

steel Gnde: 350w (Aærding to cAìacsA æ0.20/4021'M92)

Chord Fado¡ed Connedbn ReÉsnncæ (IVt') in trN br Web Wül (ör h mm) ot:

óo (mm) to (mm) 5l 6¿ 76 89 1ù2 127 152 20Ít 29 305

5r 3.2

5l 3.E I5t 4.8

64 3.2 60

to

h6¡l 3.8 86

64 t.8 136

æ 6.4 239

76 32 39 70 *'-'---'-'-t a fl76 3.8 56 101 l-å.J76 4.8 87 r58

76 6.4 t54 2T'

dt 32 31 u89 3.8 ¡15 d¡

89 ¡1.8 70 t(x,

89 6.4 124 176

1V2 3.2 27 35 49

1ú¿ 3.8 39 50 70

1V2 ¡¡-8 61 78 111

1Û2 6.4 r08 t38 r96

1t2 8.0 169 217 {710i2 9.5 213 312 411

127 ¡l-8 52 61 75 96 137

127 6.¿t 92 r08 132 r69 212

127 8.0 114 169 206 265 380

27 9.5 æ7 213 296 3EO 5.15

52 ¿t.6 47 sl 6t 72 08 t60

s2 6.4 ü¡ 9¡¡ 108 127 156 243

52 8.0 131 148 170 N 245 443

52 9.5 r86 212 24 287 3!i1 636

s2 t3 gì3 sn ¡133 5to €'¿4 fl30

æ3 6.4 75 81 88 97 10f, r39 197

M 8.0 117 127 139 152 170 219 æ8

zxt 9.5 r68 182 rgft 2r9 24 314 113

2qt r3 æ8 æ,1 354 389 1g 5s8 747

231 8.0 117 125 134 14a i 169 æ6 371

2g 9.5 r68 179 r92 zot | 213 æ5 s37

2g 13 298 318 3¡lt 368 432 525 9Í¡

305 9.5 r68 lT7 t88 I 212 213 3¡15 628

305 r3 æ8 3r5 3g 376 (11 6r5 1120

coRRECTtON FACTOR: il thefe is compressiv€ toad ¡n chofd, mullÞly by roduclron ledof /(n)

NOTE: The widlh to th¡ckness ralio lot wab members mul b€ 5 35

42

Table 8: T Connections Between Square HSS Memberssteel Grade: 50w (According to cAN/csA G4o.2o/40.21-92)

ì

I

Chord Facto¡ed Connection Res¡stances 1Nr') in tips for Web Width (ö, in inches) of:

òo {in ,o (in. 2.æ 2.50 I 3.OO 3.50 4.00 5.(þ 6.00 i a.oo I ro.oo r2.00

2.æ 125

2.æ .150

2.æ r88

2.9 .125 13.4 I

2.fi 150 19.4

2.9 r88 æ.4 ¡

2.æ .29 5:t.8

3.00 125 8-7 15.7

3.O0 r50 12.5 22.6

3.00

3.OO

.188

.2n19.6 35.5 Lu.7 62.8

3.50 125 7.O 9.9

3-50 .'r50 r0.0 14.3

3.50 .188 15.7 i 22.4

3.50 .2æ zz.a i 39.6

¿1.00 125 6.1 7.8 11 .1

4.00 .1 50 8.7 11.3 16.0

4.æ 188 13.7 17.7 zs.t i

4.00 2æ 24.3 31.3 14.4 I

4.00 .313 38. r 49.1 69.6 ¡

4.00 .375 I U.7 70.4 ooo ¡

5.00 .r88 11.7 13.7 16.7 21.5 30.4

5.fi) .2æ 20.6 24.3 29.6 38.0 53.8

5.(x) .313 J<.J 38.1 46.¿ 59.5 84.3

5.00 .375 46.4' il.7 'i 66.6 85.4 121

6.æ r88 10.6 12.O | 13.7 16.1 rs.e | 3s.5

6.00 .2æ 18.7 21.1 21.3 28.5 34.7 62.8

6.00 .313 29.3 33.f 38.1 44.7 54.3 98.5

6.00 .375 42-1 47.6 54.7 64.2 ta.o I 141 I

6.00 .500 71.9 84.6 97.2 114 139 251

8.00 .2fi 16.8 18.2 19.9 21.9 za.s I 3r.3 4/ta I

8.00 .313 26.3 28.5 31.1 3¡1.3 38. r 19.1 eg.e I

8.00 .375 37.7 40.9 41.7 49-2 54.7 70.4 99.9

8.00 500 67.1 | 72.8 79.4 87.4 97.2 125 178

10.00 i za.s 28.0 30.0 32.3 38. t ¿a.q t 84.3

10-00 .375 37.7 40.2 43.1 cø.q I s.7 66.6 : 121

ro.æ I .500 67.1 71.5 76.6 82.5 i 97.2 118 215re.æ I 375 37.7 39.8 42.1 47.6 il.7 I zs.o 141

rz.æ |

.500 67.1 70.8 74.9 84.6 97.2 r39 251

coRREcrloN FAGToR: ll there is compressive load rn chord. muttiply by reducrron lactot t lnlNOTE: The wrdth to lh¡ckness rat¡o lor web memòers must be s 35

43

Wehs Fadorsd Conn€clk¡n Resistancs(.¡Vr'orwz1 ¡n kN

ô,, ô, (mm) tr, å (rrn)

5t 3.2 r9t

5t 3.8 2U

5l 4.8 g)3

64 32 233

6¡l 3.8 245

a¡ ¿1.8 367

64 6.4 508

76 3.2 276

76 3.8 335

76 1.8 ¡l30

76 6.¡l 593

89 32 310

89 3.8 386

89 4.8 4f¡¿l

89 6.¡l 677

102 3.2 362

102 3.8 439

102 4.8 560

t02 6.4 765

l02 8.0 984

102 9.5 121l)

127 4.8 685

127 6.¡l 93r

t27 8.0 ilfx)127 9.5 t480

r52 4.8 8rlts2 6.¡l 1tfi)

152 8.0 r4m

152 9.5 r710

152 13 2370

2o3 6.4 t¿t4O

203 8.0 r83{)

203 9.5 2220

203 13 æ50

251 8.0 2250

254 9.5 2730

254 13 3730

305 9.5 3240

æ5 r3 4410

Table 9: K Overlap Connections Between Square HSS ltrembers(O, = 10O7", 0, = 02 = 45o and æme web members)

steel Grade: 350w (Aæo¡ding to cÂl,t/csA G/n.20/40.21-M92)

to

ffil--u"J

I{OTES: (t) Thc witû to thaclorrss ró torircò mcnùenmust b€ 3 3aL Also. ça,lp .cÉrs¡on $rrùs rfitÉlbe CSA-S16.1 C¡ass t.(Êælb de*¡n) s€dþns.

(2) Tho w¡dû tô lhiJgleõs rat¡o tor tha chordmeßrba? must bc f 40,

(3) Thê w*tth ratio bttrvr.n wcb m€nrb€rs an lchord rrü¡st be ¡O25.

M

Webs Fadored Conneciþn Resislanc€(IV'' or lfr') in kips

ö,,, ô, (in.) úr, t2 (¡n.)

2.æ .125 42.8

2.OO .150 52'5

2.OO .188 68.O

2.50 .125 52.3

2.æ .150 6¿t.0

2.9 .188 42.3

2.æ .250 114

3.00 .125 61.9

3.00 150 75.4

3.00 188 96.7

3.00 .2æ r3:]

3.5() .125 71.1

3.50 150 86.8

3.50 r88 11r

3.50 2fi 152

,1.00 .125 80.9

4.00 .r50 98.2

¡1.0O r88 125

4.OO 2æ 171

4.00 .3r3 2æ

,1.00 .375 271

5.(x) 188 15¡l

5.00 .2æ 20s

5.O0 .313 268

5.æ .375 328

6.00 .188 1fft

6.OO .zfi 217

6-(x) .313 316

6.00 .375 385

6.(x) .500 53i¡

8.OO .2fi 321

8.00 313 ¡ll 1

8.OO 375 500

8.00 .500 685

ro.(xl .313 506

ro.oo .375 61¡l

10.00 .500 838

12.00 .375 728

12.æ .500 9{X)

Tabte 10: K Overtap Connections Between square HSS Members(O" = 10iOy", 0t = 0z = 45o and same web memberc)

steel Grade: 50w (According to àAN/CSA G40-20/40'21-92)

NOTES:

to

ffil-'J

(1) The wicnh lo lhid('less ral¡o tor web mottlb€ts

mu$ be 3 35. Also, compr€ssion w€bs mulbo CSA-S16.1 Class 1 (plasüc d€sign) s€clions.

(2) Thc width ro thi{:l(noss ratio fottà€ cùord

mombef musl b€ 3 ¡Í1.

(3) Thê widlh tafto b€twa€n w€b msmbers and

chord must b€ ¿ 025.

45

Tabte 1l: K Gap Connectlons Between Squarc HSS Members(g = 30 mm,0t =02= 45o and egwl widû webs)

steel Gnde: 3501u (Aærding to cA¡'llcsA G¿t0.2u&.21'M92)

Chord . Faclo¡¡rt Comsctirn R€sbtattcæ (ivr' orJV21 in hN fotW€b W¡dû (ôt in rm) ot

ôo (rrn) to (rún) 5l {. 7E 80 1@, 1Zf t52 æ3 & 3CH'

51 326a 3.t¿

öx6¡l 38 .( b,

76 3.8 1G¡ trKer

7A 3.8 t3!i

76 4.8 t89

89 32 95 119

ë, 3.8 125 r56

æ ¡t.8 175 219

1ül 3.2 89 Í1 13(¡

1ú¿ 3.8 117 1,16 171L

1@, ¡t.8 16¡l æ5 215

1æ 6.¡t 251 313 375

ln 4.8 m 257 2!Xt

1tî 6.¡l 3ft6 3g¡l ¡149

1n E.O 471 551 6ãt

15¿ ¿1.8 268 335

r52 6.4 410 513

l5:t 8.0 575 719

15¿ 9.5 7g 944

2ût 6.4 sCt

2dt 8.O 717

zct 9.5 981

ãx, r3 1510

H 8.O 889

251 9.5 t170

29 t3 r80o

s5 9.5 r3Ílo

305 r3 æ50

coRRECnON FACTOR: lf üìcrc is comø¡ssivc bad in chotd, muniply by feducrioalacTo. l(n,NOTE: th. widül lo Üickmss ralio lot wrþ rîcnù€rs must bo 3 35

46

ffiorN2')inry(in.¡

2.Ð

Chord

2.00

2.æ

I 3.oo

3.00

3.50

3.50

I 3.1)

¡l.OO

4.00

5.OO

5.00

5.OO

6.(x)

6.OO

r 6.00

8.00

8.00

Ì 10.m

10.00

10.(þ

r 12.00

12.00 .500

COBRECTION FACTOB:

NOTE

3.00

ll there is compressive load in chord' mulliply by redudion lactot f(n)

#'î;h;t;.|(ness ratio lor web membeß must be 135

47

WELDED CIRCULAR HOLLOW SECTION TRUSS CONNECTIONS

by Peter W. Marshall *

ABSTRACT

This paper discusses the following elements of the subject:ArchitectureCharacteristics of Tubular ConnectionsNomenclatureFailure ModesReserve StrengrthEmpirical FormulationsDesign ChartsSummary and Conclusions

ARCHITECTURE

"Architecture' is defined as the art and science of designing and successfully executingstructures in accordance with aesthetic considerations and the laws of physics, as wellaspractical and material considerations. Where tubular structures are exposed for dramaticetfect, it is often disappointing to see grand concepts fail in execution due to problems inthe structural connections of tubes. Such "failures" range from awkward ugly detailing, tolearning curve problems during fabrication, to excessive deflections or even collapse.Such failures are unnecessary, as the art and science of welded tubular connections hasbeen codified in the AWS Structural Welding Code (Ref AWS D1.1-96).

A well engineered structure reguires that a number of factors be in reasonable balance.Factors to be considered in relation to economics and risk in the design of welded tubularstructures and their connections include: (1) static strength, (2) fatigue resistance, (3)fracture control, and (4) weldability. Static strengÊh considerations are so important thatthey often dictate the very architecture and layout of the structure; certa¡nly they dominatethe design process, and are the focus of this paper. Many of the other factors also requireearly attention in design, and arise again in setting up QC/QA programs duringconstruction; these are discussed further in sections of the Code dealing with materials,welding technique, qualification and inspection.

CHARACTERISTICS OF TUBULAR CONNECTIONS

Tubular members benefit from an efficient distribution of their material, particularly inregard to beam bending or column buckling about multiple axes. However, theirresistance to concentrated radial loads are more problematic. For architecturallyexposed applications, the clean lines of a closed section are esthetically pleasing, andminimize the amount of surface area for dirt, corrosion, or other fouling. Simple weldedtubular joints can extend these clean lines to include the structural connections.

@ystems Engineering, Kingwood, Texas

48

(713) 358 &+15

Althouoh manv different schemes for stitfening tubular connections have been devised?ffiääú. ïgdä1, inè simptest is to simply weldlhe branch member to the outside surfaceüìÏã';ä; mãlnü iór ðr¡ord). Wherbihe main member ig ¡et3!y9ly c_ompac-t (D/t less

ttä 16 õi âol, añã tt\ã orancr¡ member thickness is limited to 50o/" or 60% of the mainrãrUð,, tn¡cfñ'ess. ano a prequalified weld detail is used, the connection will develop theilií';t"¡";äp*¡tv'of the ri-renioers joined. Where the foregoing conditions are not met,

;.ä.-ùiù{ rãöe o¡ámeteitubês, a sn'ort length of heavier material (or joint can) is insertedi"ìË ti.rälnoio to ióòally reinforce the connõction area. Here, the design.problem reducesió'il;i;ãiãcting the'right combination of thickness,,yield. stLe!"tgtf', and notch toughne.ss

ior yrãJoinl tân." rnL ãeta¡ted considerations involúed in this design process are thesubject of this Paper.

NOMENCLATURE

Non-dimensional parameters for describing the geometry of..a tubular connection areoiven in the folroñ¡ng'iisr get", èta, thetá andzeta déscribe the surface top.ology-

öñ;ä¿¡ã iãü;;e ñró "et imþortánt thicknes.s parameters. Alpha (not shown) ¡-s-el

ovalizino Darametéi, àependíng ón bad pattem (it was formerly used for span length in

beams lóáOeO via tee connections).

P is branch diameter/main diameter

4 is branch footprint lengrth/main diameter

0 is angle between branch and main member a¡<es

Ç is gap/main diameter (between batancing branches of a K-connection)

Y is main radius/thickness ratio

T is branch thicknesd main thickness

ln AWS Dl. 1, the term "T-, y-, and K-connection" is used geneTcally to.describe structural

connections or nooËr, ar'oprjol"ã to co-a¡<ial butt and laþ joints., â l"ttgt gf tle alphabet

0-, V, K;ijlJus"Jto buóräå þicture of what the node subássemblage looks like.

FAILURE MODES

A number of unique failure modes are possible in tubular connections. ln addition to the

usual checks on rãlå-rä"és, prou¡oéd ior in rnost d.esign codes, the designer must check

f- tË'iãiô*¡ng ät-r;¿'mõä-e;, ¡ìsæo rogether with ihe relevant AWsDl. 1-96 code

sections:

2.40.1.12.40.1.22.40.1.32.42, C4.12.4.4, and 2.1.32.36.6

Local failure (Punching shear)General collaPseÚnzipping (prbgressive weld failure)fr¡áiäriadp rob I éms (f ractu re and de lam i n ation)Fatigue

49

Local failure. AWS design criteria for this failure mode have traditionalfy be.e¡formulated in terms of puncihinq shear. The main member acts as a rylindrical shell inresisting the concentraied radiã line loags (l)l/mm¡ delivered to it at the branch memberfootprin-t. Although the resulting localized stresses in the main member are quite-coniplex, a simpli-fied but still qúite useful representation can be given in terms ofpunching shear stress, vp:

acting vp =f6 r sin 0

where f¡ is the nominal stress at the end of the branch member, elthe¡ a"xid of bending,which aib treated separE¡tely. The allowable punching shear stress is given in the code asa function of main membdr yield strengrth and gamma ratio, as well as Qq, reflecting.connection type, geometry, ãnd load pattgm.. lnteractions between branch a,xial andbending loadéi aó úeil as bianch and ch'ord loads, are also covered.

Since 1gg2, the AWS code also íncludes tubular qonnection design criteria in total loadultimate strôngth format, com.pat¡ble with an LRFD design code.formulation. This wasderived from, ãnd intended to be comparable to, the earlier punching shear criteria.

General collapse. ln addition to local failure of the main member in the vicinity of Febranch membdr, a more widespread mode of collapse may_occur, ê.g. general ovalizingplastic failure in'the cylindrical'st¡el¡ of the main member. To a.large extent, this is nowbovered by strength criteria which are specialzedby connection type and load pattern.

For design purposes, tubular connections are classified according to their c.onfigurationff, Y, K,X, ätc.). For these "alphabet" connections, different design streng(t formulae areappfêO lo each'different type.

'Until recently, the research, testing, and.analysis leading to

tÉése criteria dealt only viith connections tiaving their members in a single plane, as in aroof truss or girder.

Many tubular space frames have bracing in multiple planes. For some loading conditions,thesä ditferent planes interact. When they do, crite¡g for the "alpåabet' joints are.nolonger satisfactóry. ln AWS, an "ovalizing párameter" (alpha, Appendix L) may beused toestímate the beneficial or deleterious'effect of various branch member loadingcombinations on main member ovalizing. This reproduces the trend of increasinglysevere ovalizing in going trom K to Tl/ toX-connections, and has been shown to provideuseful guidancé in-a númber of mop .adyers.e planar.{e.9.. doublecross,-Marshall &tuytiesigS2) and multi planar (e.g. hub, Paul ,1988) situations. However, for similarlyloáOeO members in adjäcent þlañes, e.g. paired KK connections in delta trusses,Jâpanese data indicate ihat no'increase -iñ.

iapacrty oygr_lhe coresponding uniplanarcoirnections should be taken (Makino 1984, Kurobane 1995).

The effeA of a short ioint can (less than 2.5 diameters) in reducing the ovalizing orðrustinõ caóaðity of cross conndctions is addressed in AWS section Z.qO¡.2(2\. Sinceóvãlønó ¡s ieês éevere in K-connections, the rule of thumb_is !ha! the. joint can need.onlyextend õ.ZS to 0.4 diameters beyond the branch member footprints to avoid a short-canpenalty. lntermediate behaviorwould apply to Tl/ connections.

A more exhaustive discussion would also consider the following modes of generalôollápsã, in aããNòn to ovalizing: beam bending of the c..frord {in T+oñnection.tests),. bealn;h¿äji; inã gáp of K-conneclions),.transverée. crippling of the main member sidewall,and loial UucfÍinþ due to uneven load transfer (either brace or chord).

50

Unzipping or progressive failure. The initial elastic distribution of load transfer âcros:tne we'lO ¡ñ a trinutãr connection is highly non-uniform, with the peak line load often bein¡a factor of two higher than that indicated on the basis of nominal sections, 9.pomq!ry, antstatics. Some loïal yielding is required for tubular connections to redistribute this antreãch their design caþacity.-lf the weld is a weak link in.the system, il T.y."ul?ip" befonthis redistributioln can hafpen. Criteria given in the AWS code are intended to preventhis unzipping, taking advâhtage of the higher reserue strength in weld allowable stresse:than is the nõrm elsõwhere. Fbr mild steeltubes and overmatched E70 weld metal, welceffective throats as small as70o/o of the branch member thickness are permitted.

Materials problems. Most fracture.control problems in tubular structures occur in thewelded tudular connections, or nodes. These require plastic deformation in order tc

reâch their design capacity. Fatigue and fracture problems for many different nodegeometries are b-roughi into a common.focus. by use oJ the "hot spol" stress, as would_beñreasured by a straiñ guage, adjacent to and_perpendicular to the toe of the weld joininç

branch to máin membðr, ¡ñ tne worst region of localized plastic deformation.

Charpy impact testing is a method for qualitative assessment of materid toúghness. Themethbi1 häs been, añd continues to bé, a reasonable measure of fracture ,safety, wheremployed with a definitive program of nondestructive testing toeliminate weld area flawsTne AWS recommendations Ïor material selection (C2.42.2.2) and weld metal impacltesting (C4. 12.4.4) are based on practices which have provided satisfactory fractureexperi"eÀce in offshóre structures loóated in moderate temperature environments, i.e-.40'Oet-f (+SC) water and 1 -deg-F (-10C) air exposure. For environments which eithelmore dr lesó hostile, impact teðtingìempêratureé should be reconsidered, based on LASI(lowest anticipated service temperature).

ln addition to weld metal toughness, consideration should be given to. controlling. theproperties of the heat affected ãone (HM). Although the heat cycle of welding s-ometimesimjroves hot rolled base metals of low toughness., this.regiqn will more often havedeþraded toughness properties. A.number of éarly failures in welded tubular connecti'onsinv"olved fraciures wn¡in either initiated in or

-propagated through the HAZ, oftenobscuring the identification of other design deficiencies, e-.9. inadequate static strength.

Undemeath the branch member footprint, the main member is subjected to stresses in

the thru-thickness or short transverse direction. Where these stresses are tensile, due toweld shrinkage or applied loading, delaminatiqn. ma.y occur -- either. by opening. up.pre-existing lamin"ations,'dr by laminaitearing in which miôroscop¡.. itl"_lr"-!-ons link up.to give a

fracturé having a woody appearance,-.uêually r¡. .or_l-ear the HAZ. Th"qg problems areaddressed ¡n Ãpl ioint cãn'sieel specification-s 2H, 2W, and 2Y. Preexisting laminationsàre detected with'ultrasonic testirig. Microscopic inclusíons are prevented by restrictingsulfur to very low levels (<60 ppm)ãnd by inclusion shape control metallurgy. in the steelmaking ladlé. As a practìcd rñättér, weldinents which sÛruive the weld shrinkage phaseusudlf perform satiðfactorily in ordinary seruice

Joint can steel specifications also seek to enhance weldability with limitations on carbonãná oinàr alloying elements, as expressed.by.carbon equivalent or Pcm formulae. Suchcontrols are increäsingly important'as residuâl elements accumulate in steel made fromscrap. ln AWS Appeñdix Xl, the preheat ¡equired to avoid HAZ cracking is related tocarb'on equívaleni,'base metal thi'ckness, hydrogen level (from welding consumables),and degree of restraint.

Fatigue. This subject is discussed in the companion paper on tubular offshore structures(Marshall 1996).

51

RESERVE STHENGTH

While the elastic behavior of tubutar joints is well predicted by shell theory and finiteelement analysis, there is considerable reserve strength beyond theoretical yielding, dueto triaxiality, plasticity, large deflection effects, and load redistribution. Practical designcriteria make use of this reserve strength, placing considerable demands upon the notchtoughness of joint-can materials. Through joint classification (APl) or an ovalizingparãmeter (AWS), they incorporate elements of general collapse as well as localfailure.The resulting criteria may be compared agains-t the supporting data base of test resufts tofenet out biãs and uncertainty as measures of structural reliability. Data for K, Tl/, and Xjoints in compression show a bias on the safe side of 1.35 beyond the nominal safetyfactor. Tension joints appearto show a larger bias of 2.85; however, this reduces to 2.05for joints over O.12-in, and 1.22 over 0.5-in, suggesting a possible size effect for testswhich end in fracture.

For overload analysis or tubular space frame structures, we need not only the ultimatestrengrth, but also fhe load-deflection behavior. Early tests showed ultimate deflections ofO.ffi lo 0.07 chord diameters, giving a typical ductility of 0.10 diameters foi a brace withweak joints at both ends. As more different types of jointg were tested, a wider variety ofload-defl ection behaviors emerged, making such generalizations tenuous.

Cyclic behavior raises additional considerations. One issue is whether the joint wille*perience a ratcheting or progressive collapse failure, or will achieve stable behaviorwith plasticity contain-ed at local hotshots, a process called "shakedown]' (ag ¡rlshakedown ciuise). While tubular connections have withstood 60 to several hundredrepetitions of load in excess of their nominal capacity, a conservative analytical treatmentis to consider that the cumulative plastic deformation or energy absorption to failureremains constant.

When tubular joints and members are incorporated into a space frame, the questionarises as to whether computed bending moments are primary (i.e. necessary forstructural stability, as in a sidesway portal situation, and must be designed for). or_

secondary (i.e. air unwanted side effect of deflection which may be safely ignored ofreduced). When proportional loading is imposed, with both axial load and bendingmoment being maintained regardless of deflection, the joint simply fails then it reaches itsfailure enveloþe. However, when moments are due to imposed lateral deflection, aldthen a,rial load ís imposed, the load path skirts along the failure envelope, shedding themoment and sustaining further increases in a,rial load.

Another area of interaction between joint behavior and frame action is the influence ofbrace bending/rotation on the strengrth of gap K-connections. lf rotation is prevented,bending moménts develop which permit the gap region to transfer additional load. lf theloads remain strictly axial, rotation occurs in the abèence of restraining moments, and alower joint capacity is found. These problems arise for circular tubes as well as boxconneôtions, ànd á recent trend has been to conduct joint-in-frame tests to achieve arealistic balance between the two limiting conditions. Loads which maintain their originaldirection (as in an inetastic finite element analysis), or worse yet follow the deflection (asin testing arrangements with a two-hinge jáck), result in a plastic instability of thecompresõion braõe stub which gr.ossly undêrstates the actualjoint strengrth. Existing databases may need to be screened forthis problem.

52

EMPIRICAL FORMULATIONS

Becauseoftheforegoingrgseryg-s]rengithissues,AWgde¡io-1-9r!:llllÎ*beenderivedfrom a data base ðtïn¡ñate strengtñ'iiüËïilii¿:F co-mparison with the data base

iläüärr"r"tvîlää'äî'ã6õrqiäi*Ïï:i jj,?ï"ï"Fjj*:?',?Ulî'"""äi'"lLiËjËë

nã:gU¿l**g*t"'."[i3ifñ!!þ;äãmoãi.n";dä;'i;¡"tî"ñ,in"ðnã¡cècítsaretvindex is simirar ,Jinä ;.=.äi; äiã¡üläinéi structurai rãríroéi", áher than the higher

sarety marsins ,Jriär"ff;.äå#-h¡pi;Ëiú"Ë'ö;nõt¡on items iike werds or bo*s'

when the ultimate axial load are.used in the.context of Atsc-LRFD' with a resistance

factor of 0.8, nWS ,riimate strength-is"nffiil?ÙyËq..lüidiiopunchiirg shear altowable

stress desisn fnsbi, tór structurds H"ilî';o%'dã+g it;åää;ã boø livã load' LRFD falls

on the safe side óí'nso ror structuiéö Ë"u¡ng " ro*äiiôb;t'T-gldead load' Alsc

criteria for tension'and compress¡oî mem¡eré appeãi tã naue made the equivarency

trade-ofr at 25o/oär,äo niJä;'îh';ihä [ãËit;t¡t;ñä g¡"en ov nws would appear to be

conseryative for "iãrü pã-ri of the population of structures.

ln canada, using these resistance factors with slightly different load factors, a 4'2"/"

difference ¡n overärlsafety factor ,"Jritr*-l*itñ¡n ðãíiËia:i¡än ätîut"w (Packer et al 1984)'

DESIGN CHARTS

Research,testing,andappliqati-onghavepr.oglgssedtothepgintwheretubularconnections are áËbrt as råriabre":,i'hé õth"i stnictuääèróñir frn¡"n desiqners deal

with. one of theãrincipat bars to üåïã*ìËd'""d-üä;;;Jto oe unfamitiaritv' To

a'eviate this proolËå, i"ldühàñr nãu" been þresenæo in a Appendix to this papen

,,DesigningTubularConnec.tionswithAWsP-l.l,,byP.W.Marshall,originallypî6ìËliää1n,ìíi" i¡¿" Wrng Joumal, March 1989.

The capaciw of simpre direct werded tuburar connections is given in terms of punching

shear ehicieñcY, Ev, where

allowable Punching shear süess

main member allowable tension srtress

There is also a step-by-step procedure for applying the charts in practical truss design

situations.

53

SUMMARY AND CONCLUSIONS

This paper has served as a very brief introduction to the gubject of designing weldedtubulär ðonnections, for circular hollow sections. More detail on the backgróund- and useof AWS D1.1 in this area can be found in the autho/s book on the subject (Marshall,1ee2).

REFERENCES

AWS D1.1-96, StructuralWelding Code - Sfeef 1996 edition, American Welding Society,

-,Kurobane, Y. (1995) Comparison of AWS vs tntemational Criteria, ASCE StructuresCongress, Atlanta

Makino, Y. et a (1984) Ultimate capacity of tubular double-K joints, Proc 2nd 1 1W Conf onWelding of Tubular Structures, Boston

Marshall, P. W. (1986) Design of þtemally_stiffened tubular joints, Proc llWlAlJ lntlGonfon Safety Criteria in the Desþn of Tubular Structures, Tokyo

P. W. Marshall (1992) Design of Wetded Tubutar Connections: Basis and IJse of AWS D./. /, Elsevier Science Publishers, Amsterdam

Marshall, P.W. (1996) Otfshore Tubular Structures, Proc AWS lntl Conf on TubularStructures, Vancouver

Marshatl, P.W., and Luyties, W.H (1982), Allowable stresses for fatigue design, Proc lntlConf on Behaviour of Off-Shore Structures, BOSS-82 at MlT, McGraw Hill

Packer, J. A. et al, Canadian implementation of CIDECT Monograph 6, 11W Doc. XV-E-84.072

Paul, J.C. (1988) The static strengrth of tubular multi planar double T-joints, 11W Doc. XV-E-88-139

54

SIMPLE BEAM CONNECTION TO HSS COLUMNS

D. R. Shermant

ABSTRACT

Nine different types of simple connectiorìs rypically used for l-shaped beams are examined foruse with HSS columns' The only failure limii states identified wit¡ ttre Hss are punching shearwhen a thick shear øb was used with a thin walled HSS and shear adjacent to welds. The sheartab also produced the largest wall distortion. However, column tests show that this distortionis not detrimental to the column strength as long as trrå HSS is not classified as thin-walled.Therefore, the economical shear tab can confioently be used with HSS columns as long as asimple punching shear criteria is met, and all of tire other connections can be used withoutconcern for the HSS.

INTRODUCTION

In ¡ecent years, the use of square and rectangular hollow stn¡ctural sections (Hss) as columnsin building constnrction has become increasiigly popular For connecting wide-flange beams,desþers have adapted many of the standard ri-pl" ôonnections typicatty rîsed with wide-flangecolumns, even though liftle data is available r.g;ditg their use *itt, Hés colum¡rs. However,concerns are still raised regarding these connections. The concerns are whether there is a limitstate in the HSS that could govern the connection design or if local disrortion of the HSS wallcould reduce the column capacity.

This paper presents an overall discussion of nine different fypes of simple framingconnections used with HSS columns. These are listed below an¿ strown in Figure l.shear tabsthrough-platesdouble anglestees with vertical fillet weldstees with flare bevel groove weldssingle angles with L shaped fillet weldsingle angles with two vertical fillet weldsunstiffened seated connectionsshear end plates

In all but the shear end plate, the connecting elements are werded to the Hss column and bolredto the web of the wide-flange beam, with tñe exception or the seat angle where the flange bearson the outstanding leg' For the shear end plare, the plate-is welded to the beam web and bolted

l- university of wisconsin-Mir-waukee, Mirwaukee, wr 5320r., usA

55

SHEAR TAB

DOUBLEANGLE

THBOUGH.PI.ATE

SHEAB END PLATE

FIGUBE 1 . TYPES OF CONNECTIONS

SINGLE ANGLE

SEAT

56

to the HSS column using blind expansion bolts (Ref. 1) or a flow-drill process (Ref. 2, Ref- 3)

that produces a tapped hole which replaces a nut in blind connections'

There are two categories of weld positions on the HSS for the connections shown in Figure l.The shear rab, through-plate and single angle with vertical fitlet welds have welds at the center

of the HSS face, *hil. the others have welds near the edges. Center welds will tend to distort

the wall of the HSS more than edge welds, excepr for the through-plate which provides stiffening

of the wall.

The connections are classified as simple (negligible end moment in the beam-) Rotational

flexibility is provided by distortion of the connecting elements, particularly the column legs of

angles oi R"ng"r of teei. Mosr of rhe connecrions are standard shear connections described for

use with wide-flange columns in the AISC Manual of Steel Construction (Ref. 4)' Two

exceptions are the through-plate, which is unique to hollow members, and the single angle with

vertical fillet welds. \ilhen a single angle is welded to the flange of a wide-flange column, a

vertical weld at the heel would be in line with the web and rotational flexibility would be lost.

Therefore, the standard welding pattern is an L-shaped weld with a vertical segment at the toe

and horizontal segment across rhe bottom. This permits distortion of the column leg of the angle

so that the connection can be classified as simple. With an HSS column, however, flexibility

is provided by the HSS wall in a manner similar to the shear tab. Therefore, a single angle

connection with two vertical welds is considered-

The shear tab is a special connection, even with wide flange columns, due to restricted rotational

flexibility. Distortion musr come from local yielding of the tab combined with slippage and

bearing ãistortion of the bolts in their holes. Additional flexibility is provided when the tab is

used wittr an HSS column, but some designers fear excessive distortion of the HSS wall. Hence

through-plate are somerimes specified to reinforce the wall.

The paper begins with a discussion of the relative economics of the various types of connections'

Ho*ever, thã primary focus of the paper is a discussion of the limit states considered in the

design of the connections. These were studied in a series of test programs involving 24 tests

of sinple connecrion to HSS columns (Ref. 3). Potential limit states in the HSS are discussed

and eväluated. Strain measurements indicate the relative degree of distortion in the HSS wall

and data is presented to verify that the connection producing the highest strain levels in compact

HSS columns does not reduce the axial load capacity'

RELATTVE CONNECTION COSTS

In order to put the discussion in a good perspective, information on the relative costs of the

connections ls desirable. Since a number of connection types were being studied and tesæd at

the same time, an excellent opporn¡nify was presented to determine relative costs. Relative costs

for 3 bolt connections are liired in Table I based on the least expensive (single angle with L

shaped fillet weld) being given a value of unity. The costs are for the connecting material and

the labor to fabricate thi ionnecrion, including welding to the HSS or to the beam web in the

case of the end plate. The cost of the end plate is somewhat uncertain since blind bolting or

57

flowdrilling the holes are not routine operatioris at this time. The costs do not reflect shoppreparation of the beam or field erection.

The high cost of the Tee with the flare bevel weld is due to labor and consrmable electrodesrequired for the multipass welding. Vertical fillet welds are much more economical. For asimple shea¡ connection, there is no behavioral advantage for the flare bevel welds. In amoment connection where horizontal tees are used between beam flanges and the column, flarebevel welds provide a good transfer of the tension and compression forces ino the.side wallsof the HSS and, therefore, may be warranted.

It may also be noted in Table I that the through-plate connection is more than twice as expensiveas the shear tab. This is due to the labor involved in laying out and sloning the HSS to insertthe plate. In addition, there are interference problems if connections for perpendicular beamsare required. Consequently, considerable research has been conducæd to justify the use ofeconomical shear tabs.

CONIYECTION LIMIT STATES

The connection strength is governed by limit states associated with the bols ro the beam web,connector material, welds and the HSS. Possible limit sates are listed in Table 2 with anindication of which apply for various types of connection according the AISC Manual (Ref. a).After applying the appropriate resistance factor, the lowest value govems the strength of theconnection, or the criteria can be used to establish a size limit so that a particular limit srate willnot control. The eccentricities are the result of the small distance berween the bola and weldsand do not imply that a significant end moment exists in rhe beam. Since rhe criteria for variousconnections were developed from different research programs that may have been separated byseveral years or decades, there are inconsistencies in the present state-of-the-art. For example,

TABLE I - RELATIVE CONNECTION COSTS

SINGLE ANGLE, L-shaped Welds

SINGLE ANGLE, Vert. IVelds

END PLATE

58

weld eccentricities are evaluated by elastic vector analysis in some cases and by an inelasticultimate analysis in others.

Connection design is somewhat simplified since it is unlikely that beams would be coped at thetop flange. Therefore, the bolt edge distance limits in the connecting material can be met and

no bearing reductions are required for less than minimum edge distance.

TABLE 2- LTMIT STATES FOR THE CONNECTIONSCONNECTIONTYPE A B C D&E F G H I

BOLTSShear with no eccentricity X X X XShear by ultimate analysis X X X

CONNECTOR MATERIALBoltbearing,L.,>1.5d X X X X X X XGrossshearatyield X X X X X X X XNetsectionshearfracture X X X X X XFlexural yieldFlexural rupture XBlockshear X X X X X X

WELDSShear with no eccentricity XShear by vector analysis X XShear by ultimate analysis X X X X

TUBE WALLShearatweld X X X X X X XBolt bearingPunching Shear X X

A - shear øbsB - through-platesC - double anglesD - tee with vertical fillet weldsE - tee with flare bevel weldsF - single angle welded at toe and bonomG - single angle welded at toe and heelH - unstiffened seat

I - shear end plate

Table 2 indicates three limit states associated with the HSS column. Bolt bearing applies onlyfor the shear end plate which requires bolting to the HSS. When the connector is welded to the

HSS, shear in the wall adjacent to the weld may control the capacity of the weldment. One way

x

X

x

59

to consider this is to determine the maximum th¡oat dimeruion of the weld for which the weldmaterial will govern.

(1)

where F" is the ultimate strengltr of the materialFor flrllet welds where the throat is 0.707 of the weld size and the nvo resisance factors are thesame according the AISC Specification (Ref. 5), the maximum effective"weld size is

(throaÈ),**=a$ffi+*

{2 Futnsst -aeff - -i-

-ËsS' u(wE[,Dl

t.* . L.z?ßs' +ss'Y(cab''

When the acu¡al weld size is less than w.6¡, the weld dictaæs the capacity while for larger welds,the effective weld size controls.

The other limit state associated with the HSS in Table 2 is punching shear. This is a tearingthrough the thickness of the HSS wall adjacent to the weld. This cao occur in shear tab andsingle angle connections with vertical welds where tension in the material ¡ssulting fromeccentricity pulls directly at the upper part of the weld. It can be prevented by a simple criæriathat keeps the maximum pull as determined by the yield strength in a unit length of thecoonector material being less than the shear fracn¡re capacity througb the two secdons of theHSS wall on either side of the weld or pair of welds.

F"tr*t tr"ø ( 2 (0 .6 Fut "l

) Ë"""

(2)

(4)

(3)

or

Punching shear will not occur in through-plate connections where the HSS wall is reinforced orin other connections where the pull is transferred to a perpendicular element of the connector.

One limit state for the HSS that is not shown in Table 2 is that associated with a yield linemechanism. In all the tests that were conducted with the beam simply supported at both ends,there was never enough distortion of the face of the HSS to develop a yield line mechanism.Therefore, the limit states associaæd with the HSS can be prevented from controlling bydetermining a maximum effective weld size and by limiting the thickness of the projectingconnection material when it is directly welded to the HSS wall.

The experimental strengths reported in Ref. 3 generally match or exceed the strengths predictedby the limit states criteria. Distortion due to gross yielding was usually observed at loads less

than the corresponding limit state, but this did not represent a loss of load capaciry in theconnection. Actual failure modes do not always match the theoretical critical limit stare.

However, the designs were well balanced so that several limit states have nearly the same

60

.ì

I

I

1

iI

capacity, making it uncertain to clearly discern the failure mode in the tests. The conclusion is

tbat the AISC tables for connection strength (Ref. 4) can be conservatively used for HSS

columns provided that the weld does not exceed the effective weld size determined from the HSS

thickness and that the punching shear criteria is applied for shear tabs.

The economically attractive shear tab connection was tested to a greater extent than the others.

It was determined (Ref. 6) that the shear eccentricities were generally between the weld and boltline and less than those used in the AISC tables (Ref. 4), except for combinations of HSS withvery low width/thickness ratios and flexible beams. However, in the latter cases the

experimental eccentricities reasonably matched those used in the AISC Manual. Since a smaller

eccentricity leads to greater capacity in the bols and welds, it is conservative to use the AISCTables for shear tabs.

HSS WALL DISTORTION AND COLI.JMN STRENGTH

In order to determine the effect of the connection types on local distortion of the HSS columns

in the 24 comection tests, strain gages were mounted at the center of the wall one inch below

the connecting elemenr. The transverse strains measured or extrapolated at a common 50 kips

shear form the basis for comparison (Ref. 3).

Connecdons such as tabs and single angles that have load transfer through a weld at the center

of the HSS have the highest transverse strains. These will typically exceed yield even at service

Ioads. An exception to this is the through-plate that inherently reinforces the center of the walland the rransverse strains are negligible. Connections with welds near the sides of the HSS have

significanrly less transverse strain at the center of the wall. The end plate and seat angle

connections produce little transverse strain. Longer connections with five bolts produce less

transverse strain than 3 bolt connections and HSS with thinner walls or higher b/t tend to have

larger strains.

In order to address the question of whetherlocal distortion of the HSS has a detrimentaleffect on the column capaciry, a series of tests

were conducted to compare the influence ofshear tab and through-plate connections. These

rypes of connections represent the extremes ofinducing transverse strain into the HSS wall. Aprevious paper (Ref. 7) presented test results

leading to a conclusion that there was no

significant column strength reduction between

shear tab connections and through-plateconnections. However, this conclusion was

based on only four tests using HSS with a b/tratio of 16. Recently similar column tests were

conducted with b/t ratios of 29 and 40 (Ref. 3).This study with eight tests included symmetric

t_FIG.

61

2 - COLIIMN TEST SETUP

connectionr¡ on both sides of the HSS and unsymmetric connections on just one side. Both snugand tight bolts were included in the originel four tests, but only snug tightened bolts were usedin the eight laær æsts.

The æst setup for all the column tests is shown in Figure 2. In these tests, the beams wereloaded to about 70% of the connection capacity and then a load was applied to the top of thecolumn until a buckling failure occurred in the lower portion.

Table 3 presents the column strengths as ratios of the maximr¡sr experimental load divided bythe yield load given by area times the satic yield strength from a tension coupon taken from the

. wall of the HSS. The nondimensional wall slenderness of the HSS is defined as

(s)

In the U.S., a thin-walled tube is defined as one having a less than 0.67.

TABIÆ 3 - COLUMN STRENGTHS FOR TABS vs. THROUGH-PI-4,TE TESTS

blt d. CONNECTION Pr,/P,

TWO SIDES ONE SIDE

15 1.39 Through-Plate, TightShear Tab, TightThrough-Plate, SnugShear Tab, Snug

0.530.510.s00.49

29 0.89 Through-PlateShear Tab

0.630.61

0.420.46

40 0.60 Through-PlateShear Tab

0.580.45

0.420.42

connectron on two nThe tests unsymmetrrc testsfailed gradually in bending.

The conclusion from Table 3 is that shear tab connections used with HSS column rrat are notthin-walled will develop essentially the same column snength as those where the wall isreinforced with a through-plate. With thin-walled HSS, shear tabs may have a detrimental effecton the axial column capacity. For connections on only one side of the HSS column, there is nostrength reduction for using shear tabs. It is safe to assume that these conclusion hold for othertypes of simple connectioris that have smaller transverse strains.

SI.]MMARY AND CONCLUSIONS

The test programs have shown that the variety of simple framing connections typically used in

62

steel constn¡ction can confidently be used with HSS colum¡¡s that are not classified as thin-walled' The tabulated connections capacities and criteria for evaluating .àr.""tions that appearin the AISC Manual (Ref. a) can be applied when HSS columns "r.

ui"d. The only addirionallimit states that must be considere¿ are ã simple thickness criteria for punching shear of the HSSwall when shear t¿b connections are used anã a limit on maximum effective weld size based onthe HSS thickness.

connections that involve welding at the cenrer of an un¡einforced HSS wall will produce localstrains that exceed yield. However, the resulting wall distortions are urr.ly noticeable and notnearly as great as the distortions of the conn"cting elements. The local distortion in the HSSwall has negligible influence on the column capaci[ as long as rhe HSS is not classified as thin-walled' This applies to connections on one side oi the HSS or synmetric on both sides.

careful consideration should be given to the type of connecrion specified in a design, since rheconnection cost can vary by a factor of 2t/2.

ACKNOWLEDGEMENTS

The connection and column tests programs were supported by the steel rube Instirute of NorthAmerica and additional funds for the shear tau invästigarions were provided by the Society ofIron & steel Fabricators of wisconsin and Arsc. The HSS maærial was provided by theYd9"9 Tube company of America. Special thanks is due to Dave Mathews of Ace Iron &steel company of Milwaukee who fabiicated the connection material and provided the costestimates for fabrication. The work was conducted over several years by four graduate students;steve Herlasche, Joe Ales, ch¡is Haslam and Homyan Boloorchi.

REFERENCES

1. Korol, R.M.; Ghobarah, A.; and Mourad, s. 1993. Blind Bolting w_Shape Beams toH-SS columns. J. of gtpctural Eneineering AscE, ll9 (12): 3463-34g1.)3.

: A New Manufacturing process, Flowdrill bv, utrecht Netherlands.Sherman, D'R' 1995. sirnple Framing Connections to HSS Columns. proc. Narional$teel construction conference: 30-t to 30-16. American I^rilr;;s;"iä*rruction.

7.

6.

5.

4.

l:,r,'.T1r' j_^Y t.1"d Sherman, D. R- 19!9. Beam connections ro Recrangurar Tuburar

American Institute of Steel Construction 1993.

American Institute of Steel Constn¡ction 1994.edition. LRFD Vol. 2: Chicago IL.

[or Strucrural Steel Buildings: Chicago IL.Sherman, D. R.; and AIes, J. M. 1991. The Design ofcolumns. Proc. Nationar steer construction conferãnce:Institute of Steel Construcrion.

Shear Tabs With Tubular23-7 to 23-14. American

Construction.

Columns.

63

: 1-7 to l-22.

FATIGTIE OF HOLLOIV STRUCTURAL SECTION lryELDED CONNECTIONS

A. M. Yan VYingerde', J¡ A. Packer'

ABSTRACT

An overview of the two currently-available fatigue design methods is givèn. The preferred methodfor fatigue design of connections between hollow stn¡ctural sections is the'hot qpot stress method,rather than the classification method which appears ii mosi curent tru.*J .od; ;;specifications. Recommendations for S**- Nr lines are given for aII welded HSS connections,together with thickness colrection factors and references to parametic formulas for thedeærmination of stress concentration factors (SCFs), wherc available. The design philosophies aresupplemented by a practical design example, to show the use of the fatigue desìgn tools pìresentedin this paper.

AFMN,S

Sril*

bhrtpbte

o,

SYMBOLS AND NOTATION

Cross sectional area of member considered.Axial force in memberBending moment in memberNumber of cycles to failure.Elastic section modulus of member considered.Hot spot stress range: SCF.o,External width of member considered.External height of member considered (for square sections: h - b).Corner radius of member considered (for square and rectangular sections only)tl/all thickness of member considered.Brace to chord width ratio - br/bo.Width to wall thickness ratio of the chord: b/b.Angle berween brace(s) and chordNominal stress range (stess range according to beam rheory).Brace to chord wall thickness ratio : t/to.

0: chordI : bracea: axial stress

m: in-plane bending stess

a

SubscriptsMember

Loading

'Department of Civil Engineering, University of Toronto,35 St. George St., Toronto, Onta¡io M5S lA4, Canada

64

INTRODUCTION

Falisue is the process by which fluctuating loads cause local stresses and strains which aresufficient to induce localized micro tt*.turul changes resulting in the development of cracks. lnprinciple, all structures subject to variations in live lãad stress should be checkèd for fatigue.

A few major differences exist with regard to static srength:o Failure occurs slowly, by cracks growing with each load cycle. This allows for inspection andrepar.

o Failure occurs at sress levels which a¡e often an order of magnitude lower than the staticultimate stess.

o The local sness disribution is of major importance, unlike static behavior where the ductilityof the material 9l"n allows major favorãble srress redisributions. Th"r"f"*-,h; il;;,cannot use simplified skess disributions due to yielding as a basis for fatigue design.o l¡¡ a previous AwS announcement for the 7994 conference on fatigue of stuctures,approximately 90vo of stn¡cn¡ral failures were claimed to be caused by fatigue.

In spite of the frequent occulrences of fatigue failure, the amount of fatigue-related design rules,research and education is fairly limited. Existing HSS fatigue design rulei given by the IIW (Ref.1), or in Eurocode 3 GC3) (Ref. 2) or the Aws Dl.t design coãe (Ref. ã¡, .r" all based uponresearch results for circular hollow sbr¡ctural sections only.bther design spåciRcarions, such asthe general steel building design codes by the AISC (Ref. a) and CsÀ fn"r. sl and the bridgedesign codes by the NCHRP (Ref. 6) an¿ tt¡e MTo (Ref. 7), only contain a general classificationmethod' Even the best existing fatigue design rules a¡e fairly cruãe .o-p-rã to the overall levelof modern sbr¡ctural codes. These rules are based often on the nominal sness approach, or containinconsistent hot spot sFess defînitions, inadequate thickness correction and no (or primitive) SCFformulas' As such, these codes no longer reflect the current knowledge on the subject, which hasbeen extended by r€cent and ongoing tesearch programs, especially within the Europeancommunity (e.g. Refs- 8,9,10). As a reiult, a more prJ.ir. hot qpot ,i.r, method can now beestablished to rePlace the previous inaccurate nominal stress and hot spot stress approaches for thefatigue analysis of welded HSS connections.

PRINCIPAL FATIGUE DESIGN METHODS

This method simply uses the so-called nominal srress, which is determined f¡om simple beamtheory (or: F/A * Yfì, without taking the uneven stress distribution around the perimeter of theweld into account- This stress is than plotted on the S,.- N, line of the class of the connecdonconsidered' to arrive at the number of cycles to failure. Its great advankge over the hot spotmethod is its relative ease of use. It is most useful for consrructional details ior which the fatiguebehavior does not vary considerabry with the actuar geometry of the connection.

However, for connections made of hollow stn¡ctural sections, with widely varying fatiguebehavior, either very many classes and rules have to be defined or the rules must be based on the

65

connections in the group with the lowest fatigue resistance, leading to excessively conservativeresults for other connections. The classification methods in both AWS (Ref. 3) and EC3 (Ref. 2)grouP connections together, such as in the design example in this paper, which based upon the hotspot stress method have a factor l0 difference in allowable stress for a certain number of cycles tofailure. In other cases, ttre classification method has been found to be unconservative (Ref. I l).

Hot Spot Stress MethodIn the hot spot stress approach, the fatigue life is not directly related to the nominal stress, butthrough the so-called hot spot sEess, which is the maximum geometrical stress occurring in theconnection wherc the cracks are usually initiated. The ratio between the hot spot süess and thenominal stress which causes this hot spot stress is called the stress concentration factor (SCÐ. klthe case of welded connections between hollow stn¡ctural sections, the hot spot sEess occurs atthe toe of the weld. k¡ principle, one Sru- Nr line can now be used for all t¡pes of connections,since the SCF incorporatcs the differences in stress distribution around the perimeter of the weld.

A problem with this method is the determination of the SCF. In the past rwenty years, manyinærnational investigations have been carried out, leading to S¿.*- Nr lines, together with anumber of parametric formulas for determining the stress concentration factors (SCFÐ for variousqpes of connections. However, if parametric formulas do not exist, or the pammeters are outsidethe range of validity of the formulas, expensive numerical analyses or measurements onexperiments have to be carried out. Also, the various design guides do not have the samedefinition of the hot spot stress, some definitions yielding at least a factor of two difference withothers. It would not make a difference for the design if both SCFs and S,¡...- N¡lines in one designguide were a factor of two lower than in another, but it can cause errors if the SCF is based onformulas from one design guide line and the S*.r.- Nrline from another. Therefore, it is importantthat S'¡.r.- Nrlines and SCF formulas come from the same source, or are verified with each other.

HOT SPOT STRESS DEFINITION TO BE USED FOR HSS CONNECTIONS

In order to be able to determine the effect of combined loadings, it is necessary to establish fixedpositions where the SCFs a¡e determined. For circular HSS these are the crown and saddle on thechord and brace, whereas for rectangular HSS the stresses should be considered at five positionsA to E on the chord and brace (see Figure 1) (Ref. 8). kr order to exclude very local weld defectsin the case of experimental measurements, or numerical singularities in the case of Finite Elementanalyses, the stress at the weld toe should be determined by extrapolating sEesses measured at agiven distance from the weld. As the stress increase is generally non-linear with respect to thedistance from the weld toe, a "quadratic extrapolation" is recommended. The procedure isdescribed in (Refs. 8,12).

The hot spot stress is thus defined as the extrapolated stress at the toe of the weld, along thelines of meâsurement considered, (see Figure 1). In case the angle between brace and chord isnot 90", the brace is no longer symmetrical ourof-plane of the connection, and lines A to E willhave to be considered at both toe and heel. For K-connections with overlap, four more SCFs occurin the braces in the overlap area (lines A and E in either brace), resulting in a total of 14 SCFs.

66

The total hot spot stress along a line of measurement is the summation of all nominal stresses inall members of the connection multiplied by their respective sress concentration factors. h thecase of HSS T- and X-connections, loaded by member axial forces and in-plane bendingmoments, the total hot spot stress can be determined at all lines of measurement of Figure 1 by(Ref.8):

S,¡...: o.¡.SCF,, + o.r.SCF., + o'o.SCF',o + o.o.SCF.o

As a consequence of using fixed positions for SCFs, the hot spot stresses found may underesti-mate the true hot spot stresses in each member if the direction of the principal stresses deviatesfrom these lines, especially if the stress concentration is small. In that case, the stresses at otherpositions, or in other directions or at the inside of the members, may be higher. Therefore, aminimum value of 2.0 is specified for SCF., and SCF,, in the proposed design recommendations.

Figure 1. Fixed positions at which SCFs should be determined for HSS connections.

PROPOSED DESIGN RULES FOR HSS CONNECTIONS

Basic S*"-N, line to be used for circular HSSAn eÍtensive'investigadon on fâtigue by the UK Departmênt of Energ! has been ca¡ried outrecently on the basis of 400 welded circular HSS connection test results (Ref. l0). The resultingS^,.- N, line has been proposed for inclusion in the new DEn design guidelines and is also thebasis for the proposal in this paper. As the DEn line runs at a 1:3 slope until 10 million cycles andthen at a 1:5 slope until 100 million cycles, the general shape of the line is very similar to the EC3S^,.- Nr lines. To enable future inclusion in EC3, this line has been translated into an EC3classification of I 1'4, which means that a hot spot stress range of 114 MPa (16.4 ksi) is specifiedfor a fatigue life of two million cycles. This revised S.,.- N, line, which only differs from theproposed DEn line in the high cycle region (> 5 million cycles), is also suggested for the AWSDl.l code (Ref. 8). The recommended S*..- N,line is shown in Figure 2.

Basic S*,. -N' line to be used for rectangular HSSA statistical analysis of test daø based on welded square HSS connections, together with a

thickness correction, resulted in an EC3 classification of 90 (Refs. 8, I l). The slope of the S*,.- N,line is l:3 until 5 million cycles. For higher numbers of cycles the line becomes horizontal for

(l )

67

Circular HSS Rectangular or Square HSS

constant amplitude loading (no fatigue damage). For variable amplitude loading, it runs at a slopeof l:5 until 100 million cycles as adopted in EC3 and then becomes horizontal (see Figure 2).

Correction factors for wall thicknessThe basic S*-- N,lines (EC3 class 90/l 14) are for a wall thickness (t) of 16 mm only.l. For 4 S t < 16 mm, a positive correction factor is applied to the basic S,*- Nr lines benveen

N¡1,000 and Nr5 million. This is because thin HSS connections will exhibit a longer fatiguelife than thicker HSS connections, for a particular hot spot stress rÍtnge. For N,larger than 5million (variable amplinrde only) all lines are parallel to ttre basic lines in Figurc 2agan.

2. For t < 4 mm, the influence of the root might be governing, thercby reducing the fatiguestength, so these thicknesses are outside the range of validity of the thickness correction.

3. For t > 16 mm, the correction factor of the new DEn guidelines is followed, since the rcsearchprogram on square HSS connections (Ref. 12) did not include specimens wittr >l6mm.

The equations of the S**- Nr lines arc given in Refs. I and I l. However, the resulting S*-- N¡ linesfor various wall thicknesses, shown in Figure 2, are easier to use for the designer.

Figure 2. Recommended S*-- Nr lines for welded circular and rectangular HSS connections, withvarious wall thicknesses.

SCF Parametric formulas to be usedFormulas for circular HSS are given by Efthymiou (Ref. l3). The combination of these formulasand the S^n.- Nr lines of Figure 2 for circular HSS has been verified by van Delft et al. (Ref .9).For connections made of square HSS, SCF formulas are available for T- and X-connections and

given in Table l. By means of a tentative correction factor , they can also be used for non-90"connections (Ref. l4):Lines B,C,D :

Lines A,E :

SCF is the lesser of : SCF¡ormutaeoo and l.2.SCF¡or,'.,urr.-sin2(O)

SCF is the lesser of : SCF¡ormuraeoo and l.2.SCFr"*,ur.m".sin(O)

6À

-. ¡l(x)

vtlDEOÊÆ2ú6.Doct,

ã 1(x)ctatt

o!

l0- l0Numbet of Cìdes to Fallure t{,

6'o.

=-.400_a

a

(DEOcÆ 2oott,6c¡

a/,

Ë rooÊtU'o

!

Nufnber ol qrdes to Failure Nt

68

No data is available for rectangular hollow sb:uctural sections with h*b. However, it is believedthat chords with 0'5sh/bs2 would not show considerable differences in their scFs, so rheformulas given in Table I can be used for such connections too. Formulas for square HSS K-connections have been developed (ref. 15). These formulas ¿¡re as yet not verified against testresults and further simplification is necessary since the designer now has to use about 100complicated equations for the anarysis of a singie K-connection.

Table I SCF formulas for 90. T- and X_connections ular hollow structural sections

Line BLine CLine DLines A,E

!Çft for "onn..tionr

lord.d bSCF=(-0.0 1 I +0.085.p- o.ot z.gz¡.zy@SCF:( 0.952-3.062.þ+2.382.þ2+0.0228.2ry).Zlt-0.øso+s.Btl.F-4.68s.þ4.r0.7s

SCF=(-O.05 +0.332.þ-0.258.F\.2^y Q.o8+ t.062-þ+0.s27.F\.f 0.7 s

SCF=( 0. 3 90- 1 .05a. p+ I .t 1 5.þr.Z^y (-0. I s4+4.ss5.Þ-3.sor.p2¡Minimum SCF: SCF,, > 2.0Fillet welds: Lines A,E: SCF,,:1.4g.SCFr*h (if F = 1.0, line A cannot have a fillet weld)SCF fot

"onn".rionr, loud"d

Line BLine CLine DLines A,EMinimum SCF: SCF., > 2.0

SCF:( 0.1 43 -0.20a.þ+O.06 a.F\.2.y r@SCF:( 0.077-0.129.F+0.061.F2-O.OOO¡ .2^l).?.y(t.s65+r.874.p-r.028.þ2).ro.tsS CF:( 0.208 -0. 3 8 7 .þ +O.209.F\.Zl Q.e25 +2.3e8.p- ¡ . 88 ¡ . p2). r 0.75

S CF:( 0. 0 1 3+0. 69 3. þ -0.27 B. F2¡.2y Q.7 m+ t .8s8.þ -2. t @.F2)

X-conn. ,p:1.0: Line C: SCF=0.65.SCF,*, and Line D:Fillet welds: Lines A,E: SCF.,:I.40.SCFr*r" (if F =

scF:0.50.scFf*hi.0, Iine A cannot have a f,rllet weld)

@with loads on trr" "r,ori rscnffi

Line CLine D

s cF:O. 725 Q1 0.2a8.F.a o. t s

s cF: I .3 73 Q7 0.20s.F., o.za

4*g" of validiry:

-

0.35< Ê sl.01.0 < rltS4.0

12.5< 2y <25.00.5< ho/bo<2.0

0.25< ¡ <1.0h,,õ,: 1.9

)*i:::Tib:i:ltterms (excePt the 27 þrms for line c) and is no reflection of the accuracy or sensitiviry of the formulas.

69

DESIGN EXAMPLE

In a Vierendeel truss, a T-connection is loaded as shown in Figure 3. The loads shown are actuallyload ranges. Chord is 200x200x8 mm, Brace is 100x100x4 mm. The corner radius is nvice ailarge as the wall thickness and a partial penemtion groove weld is used. Required fatigue life:2million cycles.

J F:4O kN

M:l4kNmí. | ) M-r4kNm

Figure 3Case IChord: 200x200x8 mm, Brace: l00xl00x4 mm -) F-0.5, Znl:ZS,r:0.5.Ar:1495 mm2 and 50-362163 mm3 1A¡, So calculated from member dimensions, taking cornerradii into account).Nominal sEess ranges: o¿¡-p/{¡: 26.75 MPa, ono:lvflSo- 38.66 MPa.Determine the relevant SCFs for lines A to E using the SCF parametric formulas of Table l:SCF"r A: t 4.33, SCF"r B: I 8.55, SCF"TC- I 6.56, SCRI D-8. I 4, SCF.oC={.95, SCFToD: I .62.Note that the SCF of line E is equal to that of line A (same set of parametric formulas) and thatthe SCFs due to bending moment in the chord are 0 for all lines except for lines C and D.No axial forces on the chord or in-plane bending moments on the brace occur, so these SCFao andSCF ¡ do not need to be determined. The total hot spot stress in lines A to E follows from Eq. l:S,¡,.4 :14.33.26.75-384lvPa,highest hot spot stress r¿rnge in the brace.S,¡,.8 -18.55.26.75-496 MPa, highest hot spot stress range in the chord.S,¡,.C : I 6.5 6.26.7 5.+0.95.38.66-480 MPaS,,,'D - 8.1 4.26.7 5+1.62.38.66:280 MPa.Brace: see Figure 2, rectangular sections, t:4 mm, S*,".: 384 MPa :> Nr= 300,000 cycles.Chord: see Figure 2, rectangular sections, t-8 mm, S*,.:496 MPa:> Nr= 30,000 cycles.Therefore, the fatigue life of the connection is, determined by chord failure, only 30,000 cycles.

Case 2

I-et's double the wall thickness of the chord: 200x200x16 mm and keep the same brace: Ê:0.5,2t¡=12.5,t4.25,4r:1495 mm2 and 5o:607638 ñffi3, aar26.75 tvtpa anà omo-23.04 Mpa.SCFaTA-6.19, SCRrB:2.84, SCF.TC-2.46, SCRID:7.54,SCF.oC-0.76, SCF,¡6D-1.28.S,¡.4 :6. 1 9 -26.7 5 :l 66 MPa, S.-B :2.84 .26.7 5 -7 6 lvfPl a

S,¡*C :2.4ó.26.75ú.76.23.04:83 MPa, S,¡.D :1.54.26.75+1.28.23.04:71 lvpa.The highest hot spot stress range in the brace, 166 MPa is less than 50Vo of the previous example,even though the nominal stress range in the brace has not been changed. h the chord, the highesthot spot stress range is 83 MPa,less than 20vo of the value in the firsr üy.Brace: see Figure 2, rectangular sections, t:4 mm, S*,: 166:> Nr> 5,000,000 cycles (the fatiguelimit for constant amplitude loading).Chord: see Figure 2,rectangular sections, t:16 rnrn, S,¡".: 83 Mpa -t Nf = 2,000,000 cycles.

70

Just doubling the wall thickness of the chord results in about 70 times longer fatigue life or, forthe same fatigue life,4 times the load.

Case 3I-et's instead double the wall thickness of the brace: l00xl00x8 mm and keep the chord from caseI at200x200x8 mm. F:0.5, 2y25, r:r.00, Ar2779 mm2 and 5o:362163 mm3, o"r:r 4.39 Mpaand o'o:33.66 MPa.SCF"TA:14.33, ScRrB:31.2, ScR¡c:27.85, SCF"TD:l3.69,SCF'6C:1.0g, SCFT'D:1.91.56,.4 :l 4.33.14.39 :206 MPa, S¡,..8 :31.2.14.39 :449lvlt:aS,¡..C:27.85.14.39 +1.08.38.66:443 MPa, S,¡..D:13.6g.14.3g +1.91.3g.66 :Z7l lvpa.The highest hot spot stress range in the brace, 206 MPa, is about 50Vo of case l, purely due to thechange in the nominal stress range in the brace. In the chord, the highest hot spot stess range is449lvPa, almost the same as case 1. I-ooking at Figure 2 for wall thi.k r.rr.s of g mm for thebrace and the chord results in a fatigue life of 60O,000 cycles for the brace and slightly over40,000 for the chord, so there is hardly any improvement in iatigue life, compared to case l.

Case 4lnstead of just increasing the wall thickness, let's nry a chord of l00x200xg mm (bo : 100 mm)with a brace of l00xl00x4 mm. This chord has smaller cross sectional area (about T3va)compared to the original chord of case l,whereas the brace has the same dimensions as in case l.Analyzing this geomerry: F:l .0,2yr2.5,t:0.50, Arl495 mm2 and so:214621 mm3, o^r:26.75MPa and 0.o:65.23 MPa. The nominal stress in the chord is higher than in the first geome¡ry, butdue to chord and brace having the same width, Iow scFs occur:.SCF"¡A:1.85, sc&rB:0.27, sc&rc:I.38, scF"rD:0.68, SCFrec:1.19, SCFToD:1.95.Since SCF.I has a minimum value of 2.0, SCF",:2.0 for lines A to E.S*. A :2.0O.26.75:54 MPa, S,.,. B :2.00.26.75:54 Mpas*. c :2.00'26.75+1.19.65.23 :l3l MPa, S*,. D :2.00.26.75+1.95.65.23:tgl Mpa.Brace: see Figure 2, rectangular sections, t:4 mm, S*,.:54:) Nr) 5,000,000 cycles.Chord: see Figure 2,rectangular sections, t:8 rnm, S*".: lgl Mpa:> Nr= l,OJó,000 cycles.This connection is almost OK (a factor of two in fatigue life means about 25vo difference in stressrange), despite a smaller chord than in case l. The higher fabrication costs for this connectionmay well be justified by the improvement in fatigue strength.

Case 5Staning from case 4, let us again double the wall thickness of the chord, although such a largeincrease seems hardly necessary here. Try a chord with 100x200x16 mm, 50:3361õg mm3, p:i;,2t¡=6-25 (this is ourside the range of validity of the formulas), r:0.25, omo:41.65 Mpa.All SCRr are still 2.0, SCF,6C:0.88, SCF.6D:1.43.S*,. A :2.00'26.75:54 MPa, S*.. B :2.00.26.75:54 Mpas^". c :2.m.26.75+0.88.41.65 :90 Mpa, S*". D :2.0o.26.75+1.43.4L65 :l l3 Mpa.Brace: see Figure 2, rectangular sections, t:4 mm, S*.,.: 54:) Nr) 5,000,000 cycles.Chord: see Figure 2, rectangular sections, t:16 mm, S*.,.: I l3 Mpa -, Nr= I,000,000 cycles.No improvement over case 4.

71

o

a

Conclusions from the design exampleCompare cases I and, 2: a doubling of the chord wall thickness leads to a 70 fold increase infatigue life. On the other hand in cases 4 and 5, a doubling of the chord wall thickness yielded noimprovement in fatigue life, the lower hot spot stress range being completely negated by the lowerS*,- Nr line due to the thickness effect. The designer has to strive for low stress concentrationfactors:o l.arge values of p (above 0.8) lead to a direct force transfer from brace to chord and hence

lower stress concentration factors. One can use rectangular sections to obtain favorable pvalues.

Very small values of p would also help, due to an even stiffrress disuibution around the brace.For many connections this does not have a beneficial effect until P<0.3, which is outside therange of validity of many parametic formulas.Increasing the chord wall thickness causes lower nominal stresses in the chord. Moreimportant are the lower values of 2T (yielding lower SCFs in the whole connection) and ¡(lower SCFs in the chord). Increasing the chord wall thickness is often effective in raising thefatigue strength of a connection. But if the SCF is already low, as in case 4, the SCFs remainthe same and the thickness effect will often negate the effect of lower stress ftmges.Increasing the brace wall thickness is generally less effective.The main aim of the designer is to obtain low SCFs: with SCFs of about 20 or more, as incases I and 3, the allowable nominal stress range of the connection will almost certainly betoo small for practical application.

CONCLUSIONS

It should be noted that the position of the S*.- N, line is dependent on the definition of the hotspot str€ss, so it is important to use a specified combination of S*.- N, line and parametricformulas rather than picking them from different sources. The use of a S*.- N line withoutmatching parametric formulas, as is currently the case in AttrS D1.1, is therefore notrecommended.The new S**- Nr lines, as presented in this paper, were determined in conjunction with theparametric formulas recommended herein. The recommended design procedures are backed upby extensive tests as well as numerical analyses and are expected to avoid the current excessiveover- or underestimation of the fatigue capacity. In addition, fabrication costs can be loweredfor smaller wall thicknesses yet still utilize their inherently higher fatigue strength.Clever choices of the members will result in low SCFs, which is by far the most effective wayto increase the fatigue life of a connection.

ACKNOWLEDGMENTS

The research was carried out with the financial support of CIDECT (Comité International pour leDéveloppement et l'Étude de la Constn¡ction Tubulaire) Programs 7K and 7P, the NaturalSciences and Engineering Research Council of Canada and NATO (CRG No. 930101).

72

1.

2.

4.

5

6.

7.

8.

9.

REFERENCES

I¡ternational lnstitute of Welding, Subcommission XV-E. 1985. Recommended fatieue

design procedure for hollow section ioints, IfW doc. XV-582-85, IfW Annual Assembly,

Strasbourg, France.

European Committee for Standardization. 1992. Eurocode no. 3: Design of steel strucrures

- Part 1.1: General rules and rules for buildines, ENV 1993-1-l:1992 E, British Standards

lnstitution, London, UK.American Welding Society. 1994. Structural Weldine Code /Steel, ANSVAWS Dl-1-94,

14th edition, Miami, USA.American Institute of Steel Constn¡ction. 1993. Load and resistance factor desi8¡

specification for stn¡ctural steel buildinss, 2nd edition, AISC , Chicago, USA.

Ca¡¡adian Standards Association. 1994. Limit states desim of steel stn¡ctures, CAN/CSA-

Sl6.l -94, Rexdale, Canada.

National Cooperative Highway Research Program. 1993. Draft LRFD bridse design

specifications and commentary, NCHRP 12-33, Modjeski and Masters Inc., Consulting

Engineers, Harrisburg, USA.Ministry of Transportation of Ontario. 1991. Ontario hiehwav bridee desisn code,

OHBDC-91-01, 3rd edition, Downsview, Canada.\ü/ingerde, A.M. van; Packer, J.A.; and Wardenier, J. 1994. Cnteria for the fatigue

assessment of hollow structural section connections. Journal of Constructional Steel

Research,35: 7l-1 15.

Delft, D.R.V. van; Noordhoek, C.; and Da Re, M. L. 1987. The results of the European

fatigue tests on welded tubular joints compared with SCF formulae and design lines. Proc.

Steel In Marine Structures (SIMS '87), eds. C. Noordhoek, and C. de Back: 565-577,

Elsevier Applied Science Publishers Ltd.Thorpe, T.W.; and Sharp, J.V. 1989. The fatigue performance of rubular ioints in

air and sea water. MaTSU, Harwell l-aboratory, Oxfordshire, UK.V/ingerde, A.M. van; and Packer, J.A. 1994. Fatigue design of connections between

hollow structural sections. Proc. AWS-WIC lnternational Conference on Fatigue,

American Welding Society, Miami, USA.Wingerde, A.M. van. 1992. T\e fatigue behaviour of T- and X-joints made of square

hollow sections. Heron 37 (2): l-180.Efthymiou, M. 198S. Development of SCF formulae and generalised influence functions

for use in fatigue analysis. Proc. Offshore Tubular Joints Conference (OTJ '88), UEG

Offshore.Research, Englefield Green, UK (with subsequent colrections by Shell Co.).

Wingerde, A.M. van; Packer, J.A.; Strauch, L.; Selvitella, B.; and Wardenier, J. 1996-

Fatigue behaviour of non-90" square hollow section X-connections. Proc. 7ù Intemational

Svmposium on Tubular Stn¡ctures. Balkema, Rotterdam, the Netherlands.

Wingerde, A.M. van; Packer, J.A.; and Wardenier, J. 1996. Determination of stress

concentration factors for K-connections between square hollow sections. Proc. 6ù ISOPE

Conference, International Society of Offshore and Polar Engineers, Golden, USA.

10.

11.

t2.

r3.

t4.

15.

73

I'l

IttL

tr

EARTHQUAKE.RESISTANT DESIGN PROVISIONSFOR TTJBT]LAR STRUCTT]RES

Yoshiaki Kurobane* and Koji Ogawaf

ABSTRACTS

Essentials of seismic design are outlined first referring to topics like earthquake forces closely

related with the enrigv-"uiãoing capacity Jrtto"torãr and differences in design methodology

between American *î¡up*,"i"ão¿Jpt*isioni forbuilding.str.uctures' Then three main subjects

i:l';rråf ifjåT;fh jäf

"H'ïîH:xf:*n:lig,'r'";"å'iå"rl;l:i'f ä'.iiiffi :lô"-i

secúon girder.o*".tiãorln ,eiatioo t" ¿t"tiúl*quiremens for moment resisting frames; and 3'

LessonJþamed from the Kobe earthquake' .

ESSENTHI,S OF SEISMIC PROVIS|IONS

Many seismic design codes provide static seismic forces for simple a-p$reg{arluitting stn¡ctures'

However, when strucü¡res ¿ue irregular oii.g"t than ordinary br¡ilding stnrctures, the dyqamic^

analysis is the only *.ihod to deþãnine dl-sifr-seismic forces. Two representative examples of

statiô design forces are shown below.

The total lateral force due to earthquakes assumed to act at the base of a building is called the base

shear. The base rtt"*ïá..";dúË 6 the Uniform Building Code (Ref. l) can be calculated by

t{

ttrL

rt

tIL,

I

L

r'IIl.

(1ìv_zICWr- R.

whereZ = seismic zone factor| = importance factor.C = base shearcoefficient.W = the total seismic dead load supported at the base'

R = ieduction to.ï* tã u".o*t roi äirr.renr energy absorbing capacities of various stn¡cn¡resw in cyclic loading.

The Building Standard Law of Japan and its subsidiary laws (called the Japanese building code

¡rt\he¡eafter) specifies the base shear as v = DrFesZ C W \z)

whereþ- = reduction factor having a function similar to 114*'í:.= äiriprìäã":,iãliã"tot ioïcounr for vertical stiffnëss inegularity and horizontal torsional

inegularitY.Editorial modifications of the original formula are made in Eq. 2 so that the two formulas follow

the same format as much as possible.

The intensity and nature of ground motions assumed in these formulas are rePresentedby ZC'

x Professor, Faculty of Engineering, Kumamoto University, Kumamoto 860, Japan

ai

!

I

:L74

REQUIREMENTS FOR BRACED FRAMES

l,< 1890/tî

),,s 4giljí4gyJl < 1.3 891/\,-F"

or

ßilt+E s x

0.3<p3 0.7

Sglt\tl < À<1981/r,T 0.3<ps 0.7

Tabte I Comparison of Rn ÞllDr) Values for Special Moment Resisting Frames and BracedFrames between UBC and Japanese Building Code Provisions

Dr=l1R,

0.083

0.083

0.25

0.25

Japanese

Building

Code0.35

a) 4 = specified minimum yield stress of steel being used. MPaNL = no limitÀ = slenderness ratio

b ) Ê = rat" of lateral force resisted by bracings to the total design lateral forceWhen p = 0, frames concern the special moment resisting frame. *hen p > 0. framesconcern rhe dual system consisting of a SMRF plus bracings.

0.25

0.3

0.3

0.4

3.3

2.9

3.3

2.9

2.5

0.8

o 0'6N

0.4

o.2

0.0012345

I seconds

Fig. I Specified Ground Acceleration Spectra forDesign (S,J'SrJo denote the site coefücienlSoil becomes softer in this order.)

which is expressed as an acceleration responsespectrum of a single-degree-of-freedom elasticsystem with a damping capacity of 5 Vo critical.The values of ZC for Zone 4 (the zone ofhighest seismic risk) calculated from the twoformulas are compared in Fig. I This figureshows that the ground motions assumed in thetwo codes roughly coincide.

A great difference between these two codeslies in values of R =llD. These values fortwo typical buildiirg strructures, a momentresisting frame and a braced frame in steel,¿ue compared in Table l. As is evident in thisfigure, the Japanese building code isrecommending more conservative design thanthe UBC. The UBC. however. uses theallowable stress design criteria to proportionstructures. In addition it specifies designdetails to ensure sufficient ductility of thestructures. Therefore, R,, serves more

1.2

1.0

-s,-St-S'

s,

7s

0.6

o.5

0.4

0.3

o.2

o.1

0.0

0.6

0.5

0.4

0.3

o.2

0.1

0.0

RANK I

RANK IIRANK III

:

Es:rel¡betfol,tnrsecI-e:

MaHoanfsta

Th'she

wh

Th'her

Ed:the

Th,

5

(a) BEAM MECHANISM

Fig.Z D, Factor for Monent Resisting Fra¡¡es according to AU ßet 2)

1510 15

(b) coLUMN MECHANISM

PLASTIC DEFORMATION FACTOR OF MEMBERS

REQTJIRED FOR STRUCTTJRE OF C

RANKI RANKII RANKIII

3

6

0.75

r.5

0

0

5

8.4

2.75

3.9

",

2.4

N

FAILURE

TYPEA

ductile

bnrtle

Note:¿¡ The ductile failure shows a gradual load decay after reaching the peak load owing to plastic insabilitylike local buckling of plate elemenrs. The brittle failure shows a sudden loss of load-carrying capaciry at

the peak load. See Fig. (a) below. When ductile failure occurs, the energy absorbed in the decayingbranch of load deformation curves is taken into account for evaluating hysteretic damping capacity.

b) The both beam and column mechanisms form a panel mechanism sustaining plastic hinges either atbeam ends or at column ends. respectively. See Fig.(b) below. The girder to column connection shouldbe detailed to have a sufficient strength capable of developing plastic hinges either a¡ girder or columnends.

c) The plastic deformation facror is defined as the ratio of õol6,on the fictitious perfectly elastic plastic loaddeformation curve, which has the area under the load deformation curve OAB' equal to that under theactual load deformation curve OABC. The cumulative plastic deformation factor of a story under cyclicloads a¡e calculated based on the plastic deformation factors of members and connections. The stn¡cturesare classif¡ed in Ranks I. II, and III depending on their ductility. See Fig. (c) below.

DUCTILEFAILURE

wh

.F

DEFORMATION

lô"1 õe Ila.)la-.'....+l

(c)

76

Table 2 Rantrs of Structures Vùicd with their Ductilitr ßef.2)

DEFORMATION

(a)

functions than just a physical reducdon factor ro reduce the base shear. The selecrion of R*. values

have been made in. ä¿#äñ;;;äiã"¿¡"äg."enhl mannerbased on rhe pasr experences'

The Japanese building code, on the contrar¡" uses the ultimate-strength design criteria to proportion

structures. D, is linr?åä;"Iñ" -t'tr-e-ráauction

in responses io"ground-motio-ns' so far as sreel

struc$res "r. .on..r,i"d.

-in *äny .ur"r.tr,îiãl-uä or p, i, determiñed by considering the barance

of the energy input,".r,ä.*äïä¿iñ" ¿itiipä"J;;;i;$; anã inelastic äeformations of stn¡ctures

dwine earthquakes ¡" ãi.*-pl.orproporäJ;;i¿ñi9t *uüiiiotv moment res-isting frames is

i'ustiated in Figs. 2, iö;d i6i fnef, Zi. ît"r.'f,g*"s show titt uãfi"t of D' as functions of the

number of stories N;;äîË ì,iäilrc ¿"r";ä*" ;ä;iv ãr.tt.ottures' and ínclude (a) moment

frames with suong .fiä;"iä;;"ù;#;;ää õilñ;:iÑith weak columns and suong beams'

The prastic ¿"ror,outiol'.ãp*i,v or"*u*pìîriurir'i'-" ã"nneðin Table 2' Naturally, D. (=l/R*')

becomes srearer ., itä Ëräi;'"*;i ,n* iãiit'. iãÃ"ion"r ¡r.uui" áamases (namelv'-inelastic

deformations) tend,äT""à""*à on "

fewer stories in the latter ones'

Most of the buildings designed according to the uBC are governed by the story drift limitations'

which are a riure moie strin-gent than thos. i'"",iri¡ãl"i'"t.-u""iloinfcoaä. In consequence buildings

designed according; il# *" ¿inerent cf¿e. iËn¿ ro have á-uãut tt" same safety lever against

earthquakes, in "onrräiä;".e';;tãiff"r"n""r

in R*. factors between the two'

The frequency of earthquakes increases in inverse-propôrtion to an-exponential function of the

magnitude. Th. Jd;,"Ëî;ilffi;;4; specifies sêrviceab'itv limit siate criteria for moderare

earthquakes ,nu, ."'Ëïiä-.*ã'iã-õ3"u, f"."fi;ï;i;;; õ;úõä siruice period of each building'

The a'owabre srress ãå'rign is used *ith ;;;.ri,";;. th" Úriõ óo¿. spêcifies the serviceability

desisn crireria only implicitly. rire story ãriüii*¡t"tions.menìioned abou" are one example of

thesã criteria. oesigning struätures r9r rai.e i"iå"rã ""nh-quakgs

bv requiring rhat structures remarn

nea¡rv elastic i, gr";öi"å.äîãøä -¿ ""¡rî,nûl. fiom the irouáuitity of such an occufrence'

Re sp-onse s,o "unr,

q',jårî ä;ffi ;; ñ"*:i; ;;ä;ñ;ä " äåti uet v uv tne- gnlrgv di s s ipati on

rendered uv in"l"rtiJääårirî"iiä,i*rr*;;;;'ririr ttår'¿r"adv been discussed tully in connectron

with the R... factor. The both codes allow' it*"tu'ul aamages io buildings due to yielding during

severe earthquakes, à" .ã"¿i,ion rhat a .uä,^r"pil;.ãiiupí. oii*.ror.i reading io loss of life is

avoided.

Designershavetopredictfailure.modes.oftheirstnrcturestoevaluateseismicforcesfordesign'This is an essenrraä'"iä;;;ãr in. t"i;;;;;ry" ilñ the designof structures against other

loads tike gravity "r;;;-'";t-¿ *tãîl;;;' ih" subje*ofltnãplastic deformation capaciry

will be discussed *oiã tfttifically in the following Sections'

SEISMIC DESIG¡{ OF TRUSSES

Lattice girders afe sÛonger and lighter-than l-section girdersìn general' Loads other than seismic

forces frequently gfîïråìirJä";;ö;i Ë.'-;o* ñ;;es' Ho-üeuer' there exist trusses seriouslv

affected by earthquakes. High-rise rp""å;"ì"Ñdü;r:,h trussed frames damaged during the

recenr Kobe earthquake are oñe outstandüñ;öi;ïä;11¡^frussed structures are more difficult

than special ,nom.r,tì.sisring rram.s to-ã"iign ^fo, èarthquakes owing to. a greater difficult-v- in

predicting faiture ,";ä;t:'ïËï"ii,o¿ f- ,f,. t?lt*. desigir-of hollow section trusses' however' rs

now advancrng rapidty based on ,..rn, .*,înt-iä:lLlì:íJese stu¿ies include tests of several

complere steel trussei(Refs. 4.5) as well as compositt t*tttiãither with concrete slabs or with

conirete-fired chords (Refs. o,zl. reni'a;'";ä¿ñ;ild.li;;; háve recently been proposed by

Architectural Insritute of Japan fRef. 8).. îunt.r r.íitiõn of the guidelines are now in progress ano

w'r be incruded i";;1ii'R"commendaìions tt¡at wilr be t.uläa in the nea¡ future (Ref' 9)' The

77

-j

following part of this section first reviews the behavior of planar, triangulated, directly weldedsreel trusses under cyclic loads. Then, AII proposals for the seismic design of SHS lattice girderswill be discussed.

Behavior of Circular Tubular Trusses under Cyclic Loads

An example of load vs. deflection curves of tn¡sses under cyclic loads, extracted from a series oftests of 15 complete tn¡sses (Ref. 5), is shown in Fig. 3. The tn¡ss was a Warren type cantileveredtn¡ss under a cyclic shear load applied at the loading end. The tn¡ss first sustained out-of-planebuckling in one of the braces, accompanying a sudden drop in load. The buckledbrace is indicatedby the B symbol. After this, the deflection increased at a nearly constant loa{ showing a stablehysteretic curve, because the chords carried a pan of the shear load as beams. The small open dotsin the figure indicate formations of plastic hinges.

Ar this stage, the K-joints sustained a shell bending chord failure at positions denoted by the Ssymbol. This was caused by a redistribution of loads in members. Axial forces in members framinginto K-joints were balanced before braces buckled. After a brace buckled, however, the axial forcein the brace was quickly lost. Then the K-joints ca¡ne under combined inplane bending and axialIoads and failed at a load lower than the capacity of the K-joint under balanced loads. This sequentialfailure of the K-joint wz¡s confirmed by drawing a load path of the measured axial forces in the twobraces framing into the K-joint. An example of such load paths is schematically shown by the pathA in Fig.4.

The load path of the axial forces in the two braces first follows a 45 degree line because the twobraces make the same angle with the chord. Compare the load path A and the K-joint on the lefthand side in Fig. 3. The axial force in the compression brace suddenly increases while that in the

r50

-ïESTAMLYSIS

P(k

,' R(rad.l-0.03 -0.0 0.03 0.04

-r50

Fig. 3 Load rs. Deflection Relationships for Truss

78

tension brace suddenly decreases as soon. as the other compression brace immediately adjacent tothe tension brace buckles. As the.load path reaches the ultiriãte capaciry polygon shown by dashedlines, a shell bending.failure of the joint occurs. The ultimare-capacity polygon for K-joints hasalready been discussed by the authori (Ref. l0). when tne axi¡ forôe in íi,å t.iíion brace decreases,the K.-joint capacity decreases afo-ng the line. segmenlfreaain! ior trre vJoini.ãf^"ìij, i" compression.Another example of sequential failure is sheñ bending chõ¡d failu¡ã åi; Kj;i;ï?orro*in! iãieiuibuckling of a chord, although the test results are nor ri¡ãrn ¡ere. The braces iustaine¿ out-of-planebending loads after the chord buckied laterally.. The K-joinr failed ar a load irg"ii=r.-uv lower rhanthe capaciry of the K-joint under balanced ¡oa¿s because ãf comuined load "îr""t.'

The P symbol in Fig. 3 denotes punching shear cracks in the joint. The C symbol denores crackinitiation and extension in the brace wdlJdong the wel¿ ioes.'rnese cracls úãiãiouna only afterthe joints sustained significant shell bending a''enection "iiuur

walls, either due io shell bendingfailure of the chord wãus or local bucklingäf tn. .;;p;r;i;n ut^ð. fS.e i,g. 5iì These cracksappeared to be ductile tensile cracks accoñrpanying shåarilip_planes ùur.iæî¿"å rapidly undercyclic loading, frequently having led to a conipleíe sõpatation äf a brace from a chord. The materialat hot-spots along the weld toes sustains.larle plasti.ìttàiniwell into. rrráin-t-¿ening range.The material's toushness deteriorates owing ío i.p""iLã .ãiã-*otting. rtriiitroül¿ u. rhe reasonfor quick developrñentt or.tuóLi,ã,iã"ghîo c¡ie¡on nái u".n identified to predict initiation andextension of ductile cracks at weld toes.

The dashed lines in Fig. 3 show the ¡esults of a point-hinge frame analysis, in which the plasticdeformation over a Ie¡efh of a member is in-corpoíat.Jin täoial and rorarional deformations of aplastic.hinge. The elastic and inelastic deformationr ãr¡ái*r a¡e also taken into accounr in thellaJrsis' The figure shows that the analysis represents aótual behavior observed in the test well.Although the analysis s.ho¡v1 herewere nlrrgrmLa i" iõ8ilhr method of point-hinge anatysis hassince been improved to include strain-haràenin_g effecis, *iìån ã"¿e possiblË ió å""urit.lv reproducethe post-buckiing behavior of tubular struts (ñ.ef. I I ).

In the tests of l5 trusses' some of the joints failed before members buckled. An example of loadpaths observed when K-joints failed beÏore buckJing o¡;;;b;;t ir ill*tiãü üvì¡. p",¡, B in Fig.

N2M\/ GOVERNING CRITERIA

K.JOINT

XP.JOINT

Y-JOINT

LOCAL BUCKLING

xN ¿ ¡1N s.

@?rNsxNt

K

XP

Y

LB

@ puNcHrNG

'HEARCHORD WALL FAILURE

DUE TO PALSTIFICATION

OR PUNCHING SHEAR

Fig"l Load Paths of Axial Forces in Two Braces Framing into one K-Joints and ultimateCapacity Pol-vgon for Joints

79

Fig. 5 Failure of K-Joints under Cyclic Loads Showing Cracks at \{eld Toes

4. After the load pa¡h reached rhe ultimate capacity polygon, the a,rial load in the com-pressionbrace remained constant while that in the tension brace increased further along one of the linesegmenrsofthepolygon. Inallthesejoints,thecapacitiesobservedintn¡sstestscoincidedaccurarelywiìh those predìcted by the ultimate capacity formulas derived from the results of isolated jointrests. Namèly, no significant effects due to different boundary conditions be¡veen actual jointsinrrusses and iiolated joints (e.g. secondary bending moments and end restraint) were found. Theuidmate capaciry of the K-joint is governed , unless tensile fracmre occurs, e-itherby localized shellbending deflection of the êhord wall or by local buckling of the compresfion brace in the reglon

adjacent to the joint. Theultirr-rate capacity *N Vwhich are most accuratelypredicted by the formulasof Kurobane et al. (Refs.12,13), can be representedby the equation

t{u= min(f,¡fs, xlv¿)

(3)where

rìL = the capacity ofthe joint deter-mined by chordwall shell bend-ing failure.

iYt = the caPacitY ofthe Jornt cleter-mined by bracelocal buckling.

Proposed Design Criteria

Conclusions drawn from the t5 rn¡ss tests may be summarized as follows:l. When trusses are under static loads like graviry or snow loads. existing capacity equations based

on isolated joinr rests are effective to predict the ultimate bebavior of joints in tn¡sses. There isno need ro consider sequential failu¡es ofjoints following buckling of members. Trusses may bedesi_ened either to have stronger joints than members or vice versa However, appropriate valuesfor the resistance factor should be assumed with due considerarions on failure modes. K-jointswith an excessively small _eap size may sustain prematue tensile failures with insufflrcient ductiliry(See Ref. l4). Trusses may fail more suddenly than the example shown in Fig. 3, when failuresare soverned by buckling ofslenderchords.

2. Two merhods are applicable to the seismic design of trusses. The first method is to desi-sn thetrusses to have suffrcient strengh so that both the joints and members resist the maximum possibleload effects. However, the crack growth along the weld toes under cyclic loads must be avoided.One of strategies to prevent these cracks is to keep a reserve ofstrength forjoints so that chordshell bendin_e or brace local buckiing failures do not occlu at the maximum seismic loads. It istentativelv proposed, from the 95Va confidence limit in verv low-cycle fatigue test results for T-joints(SeeRef. 15),todesi,rnthejointstobe25Vc sfongerthanthemrximumloadeffects.

3. The second merhod of seismic design is to desi_en the trusses to have sufficient ductiliry so thatthey w-ili not collapse under the most unusual external excitations. In this latter case, jointsshould be designed againsr sequenrial failures includin-s tensile f¡actures. The rest results indicatethat such sequenúal failures couid be avoided. rvhen thejoin$ are 25Vo stxonger than the buckling

80

Table3PlasticDeformationCapacit¡ofLstt¡ceGirdersandLimitingDimensions

RANKPLASTIC

DEFORMATIONFACTOR

xT

I n III

5.00 2.75 2.00

x¿ll8 ll8>x¿lll4 x>1120

i.s0.23+-t )'s0.23+2x

l{"", " = L/L: Length of the end segment divided by the total

length of lower chord

I : slenderness rario ofthe lower chord

loads of members. Inorder to Perform thesecond design method,however, the energYabsorbing caPacity oftn¡sses have to be eval-uated based on buck-line and post-bucklinghvíteretii behavior oftíusses. The Point-hinse frame analysismeihod is one of thefeasible waYs for thispurpose.

ALI Desisn Guidelines

The recent Au guiderines (Ref. g-) propose definite crireria for the anarysis and design of rrusses'

which are applicabþärhJ ñ;r d;;igi úñä;;ntioned above lttre sirengttr design method for

earrhquake loads). ñrË;ã-;ir".t"in" r.nÀtt i"rtors for truss members *ere derived on the

assurnption tnat¡oints åËf,;;ä "Ë-riäry.-îñ;; tt

" rtt"ngtr, design method requires that joints are

25 va sûonger rhan ,hä;ñ-uãloui'"nä,i: rÑ effãctive lãngth factors-can be used in the

sEength design.

TheAlJguidelinesProposesaductilitydlsiencriterionfortrussesbeingusedashorizontalmembersin soecial moment ;:üJn!Ë"*Ë;. Þ;rï,;;T;;ruji, in¿ir"te thai rattice girders under anti-

svrnmetrical bending roads usua'). sus,ui;';Ëstic ã1ial aefolmarions of chords concenrraring at

their end porrions,..il iîî;r*"IîJ.ffiË óf ine chor¿s goverm the capacity of th-e girders' Further'

the upper chords oruuiiv do not buckre "ïiü;"-ä;*iiing "n .tr srippüéd by floor systems. It is

possibre ro assume ,$'i.ï"äià,î;Ë;;;ã. ãiul"rti.e firder ttrat ttrê loweichord buckles in the

end segmenr ar one JnJ,iril. the rower 9I"ø viãros in tõnsion ar the orher end (see Fig' 6). The

duct'iry is further i";;;;ã;nen tarer¿ u',*iJrif ;itó*er chords are resrrained by concrete slabs'

When this failure *;ã;-it "t;umeq,

tn" ¿.îotmätion capacity of lattice girders can be g1l* 1:

Table 3. The pf"rti.-ã"fãr.ution factors in Table 3 conespìnd to those for brittle stnrctures ln

Table 2, because tuúuøt it*rs show a quictc load decay after local buckling starts to occur at

flexurally buckled sections'

when the ductility design is performed using the d:llySli:i::p-":i:iî:,:1?Il'll?lli,?; tiHI}iÏ:;Ï:.:Ï:iäT.ï"Ääi'iìfi¿i;iü;ü;'**{T¡:::::::Ì";1îil3:*ïî:,3:'Hå#lli;ät";;.qtitements Te: 1. Both thé upper th:It T9 Ïltt"til;;Ñ;ki.lä. Âil tte joints are iuong enough; and.3'

rensire ihe lo*er ciiãtot t u'. an *, 1f1i::,:d,1f?yi:îÎ:p,1iliîl'if , e

i,îË;; i i i ;r:' I t.'j:ll, ::: : J:,'j -ql.::'l'. l,"f llb;.kli;ti;ad. The last requirement imp'o':: i t",tl:lTlh

(emax - ey) I€>tag L"

Lii"î."1i'"*ing ãiu*.,rt to thickness ratio ón chords. Less

ri¡"Ëã"t design-criteria for Iess ductile latdce girders are

now under deliberation.

When the girders are designed t-o þe. stron-ger- than the

columns, tñe strength desig'n metho-d is applicabl.t t: ll!lattice girders. In this latter case' D, factors sho'*'n ln I aÞle

fÁã2"ã"¿ Fig.2 can be used becãuse plastic hinges form

inìft. .ofumns-- The latter design is more popular than theFig. 6 Plastic Rotations at Girder Ends

81

i

-l

ductility design, especially in low-rise large-span buildings. The UBC also recommends the latterdesign method.

The dynamic analysis is the only method to proportion more comp.lextruss structures. The APIrecommendationJ(Ref. l6) a¡e proposing dynamic analyses for the design of offshore n¡bula¡stn¡crures. The point-hinge frame analysis can be combined with the time history respons€ unqly{:for this purpose. Emphasis is placed on that joints should be designed to be strong enough to fulfillthe suength or ductility requirements mentioned previously.

RESPONSES OF MTILTISTORY FRAMES1VITH RIIS COLT]MNS

TO STRONG EARTHQUAKES

More than 90 per cent of steel multistory building frames inJapan use box-section colurnns due to their excellent cross-sectional prop€rties to resist biaxial bending loads. Cold-formed sections are cheapest and used most frequently. Thesesections are classified in two types by manufacturing process.When plate thickness is greater than about 20 mm, plate isbent at 4 corners and welded longitudinally by submergedor gÍìs metal arc welding. Lighter sections a¡e manufacturedby continuous cold rolling and electric resistance welding.Hereafter, the former type is called the pressed section, whilethe latter is called the rolled section. Cold-rolled RHSsections experience cold working in both the longitudinaland transverse directions, resulting in final material propertieswith a high yield stress and high yield to ultimate tensilestrength ratio (of about 90 7o). Pressed sections experiencecold working only in the corner regions.

The Japanese building code requires that girder to columnconnections in special moment resisting frames are strongenough to wa¡rant formations of plastic hinges at the girderor column ends. The most rypical details of girder to columnconnections designed to fulfill this requirement is shown inFig.7. These connections have through continuity plates(called the through diaphragm hereafter) at the positions ofgrrder flanges. Recently these connections are fabricated bywelding robots, resulting in a significant reduction infabrication cost. However, the amount of weld deposits isconsiderably large.

In 1987 a test performed at the Universiry of Tokyo revealedthat pressed RHS sections with artificial notches on thecorners sustained brinle fracture under bending load reversalsof a few cycles. Later tests of RHS columns with the throughdiaphragms showed that brittle fracture could start fromductile thumb nail cracks that developed at the weld toes onthe corners of columns during as early as the first or secondhalf cycles of load reversals (Refs. 17, l8). Material deterio-raúon due to cold working and high heat input during welding

l6ml6ml6ml6ml

Fig. E Example of Frames

82

were idenrified as lwo main causes lo1-?rly developments and-grow$,of cr19!¡' ^Disputes

alose

about the suitability;'f ;Já:iorrnãã nH9 ,J.rion, ai columns in-special momenr resisting frames'

The fottowing inu..tigaãä,i(iùï;. 8, rgl is Jn; ãithãìôt"*ortt'y "tþt"t's that discuss the ductiliry

;ñ;;ilË*dng írames with RHS columns'

AseriesofdynamicanalyseswereperformedonseveralmultistorybuildingframeswithRHScolumns and r_secrio";;j;: ù"ìå,íJ"o;ü;;;ry;ã p-¿aa"neáts *"t" taken into accounr bv

using a ooint_hinge ;ålË;il1'rrïã.'îrËl*,iå-rn"tã"ror*"tions of connection panels were also

ããoä¿,i"¿. on. ;r,lÍidpiÈ i#i;ä;liffi;J; Ëis I. rh" **Y;¿i:i:ï"1';,ff ? iffi ::

5 STORIES 10 STORIES 15 STORIES

0 0.01 0.02 0.03B¡(rad)

l-J-oEI

t-.LouJI

t--at¡lJ-

0

1

0

1

l-IotrJ-

FJ-IulI

l--oEJ-

t-.LoúJ.

l-J-otrt

t-otu

l-.J-oIJJ)-

l--J.IUJ-

l-rou,:E

0.01 0.02 0.03F¡(rad)

200

: COLUMN

i -ietnoer

---;----i.----:----_-:----:--------

10200rl

TYPE O FRAMES

0.01 0.o2 0.03F¡ (rad)

All the girder to columnconnections are designed as

shown in Fig. 7. Theframes consist of three tYPes

as follows:l. Type O frames Propor-donðd by the Plastic design

method in acco¡dance withthe Japanese building code

"ssuming a D., factor of

0.25. Member ilimensionsare determined just to fulftllstrength requirementsreeardless of the dimen-siõnal standard' The driftIimitations are i gnored-Z.TyWD frames in whichthe bending strength of

sirders are increase arc JTíim"s; the shea¡ strength ofconnection Panels areincreased to 1.2 times, thestrengths of members and

connection Panels, resPec-

tively, of TYPe O frames.TvpeD frames are designedtó'see the behavior offrames when the directionof horizontal groundmotions makes an angle of45 degrees with the girders(oblique earthquakes).3. Tvbe R frames designedaccõtding to the allowablestress design method with aD-factor of 0.2 as sPecified

in' the Japanese buildingcode. Members are ProPor-tioned following the normaldesign practice. The driftlimitations a¡e considered.

20

r 10 20 ,r 1o 20 n1o 20

: COLUMN' Vvbv.rrt r

--.i----i----!-'-

10 201

TYPE D FRAMES

. i IGIRDEF

l¡

f -"i-" i -'

+---.i----i---

0n1020

TYPE R FRAMES

Fig. 9 Responses of Example Frames to Strong Earthquakes

83

The ground motions usedforthe analysis are the 1940El Cenuo NS component,the 1952 Taft EW compo-nent and artificial groundmotions. Thetwoobservedground motions are scaledto conform with the desþspecra shown inFig. I (themaximum velocity ofground motions is equal to50 cm/sec). The artificialground motions show asmooth velocity responsesPectn¡m curve at 120 cmlsec over the period rangegreater than 0.6 second for2per cent damping.

The results of analyses areshown in Fig. 9. The heightof stories from the base isshown as the ratio to the

total height of frames. The 3 graphs on the first row show the story drifts R, for the 3 types offra¡nes. Story drifu of Types O and D frames lookthe same. Story drifu of Tyþ R frames exceed1/100 slightly.

The graphs in the second to fourth rows plot the cumulative plastic deformation factor at eachstory, which is the sum of plastic defonnation factors in the positive and negative directions, sustainedduring the earthquakes- The response plastic deformation factor defined in the above is denoted by4. As seen in these graphs, plastic deformations occur mainiy in the connection panels, with themaximum 4 values being less than 20, while plastification of columns and girders is less. Thisstatement is applicabie even to Type D frames that have the suengthened girders and connections.Plastic deformations in the columns concenrate only on the lower ends in the first story, with the 4values being less than 10 in Type D frame and less than 6 in the other frames. These values of 4 caneasily be accommodated with by Rank I columns specified in Table 2. Plastic deformations in thegirders ¿ìre greater than those in columns. This fact, combined with extensive yieiding in the con-nection panels, helps avoidin-s concentrations of damages to the columns on a few limited stories.

LESSONS LEARNED FROM KOBE EARTHQUAKES

The Kobe earthquake recorded ground motions significantly stronger than those assumed in thedesign spectra shown in Fig. 1. One of the most unusual damage patterns found in steel multistorybuilding frames after the 1995 Kobe earthquake is a tensile fracture of lower flanges of girders.Cracks started from roots of cope holes (See Fig. 7), at toes of girder flange to diaphragm welds orat notches formed by welding steel run off tabs on the both sides of each girder flange. Thesecracks frequently changed to low-energy fast failures as they grow (See Fi-e. 10). Although rensilefracrure at the welds between RHS columns and through diaphragms were found in manlr low-risebuildings, lack of penetration existed in all of these cases. No tensile failure rhat was expected rooccur at the weld toes on the corners of cold-formed pressed RHS columns was wirnessed. as fa¡ assound full penetration welding was performed.

Fig. 10 Cracks initiated at a root of the cope hole ran acnoss the lowerflange in a brittle Eânner.

84

The damage parrern described above, however, is found reasonable from the numerical analysis

results deõri'UeA in rhe previous section. Bending moments at the column ends are bounded by

yielding of panel zones in the connections unless panel zones-are reinforced. The increased yield

rtt"rs oirn"i.rials in cold-rolled RHS sections also helped avoiding tensile failures in these columns,

because more energy was dissipated in girders. Since girders are frequenlly weaker than columns,

details causing stress concentrations at the girder ends should be avoided.

REFERENCES

l. Uniþrm Building Code.199l.International Conference of Building Officials- Whinier, Ca.

2. Il¡i¡nate Strength and Deþrnation Capaciry of Buildings in Seismic Design.1990. Architecn¡ral Institute

of Japan, Toþo, Japan (in JaPanese).

3. ReconnaitsoÃr" Ràport on Damage to Steel Building Structures Observed Jrom the 1995 Hyogoken'

Nanbu Earthquake. 1995. Steel Committee of Kinki Branch, Architecrural Institute of Japan, Osaka,

Japan.4. Inoue, K., Yamamoto, K., Matsumoto, K., and Wakiyama, K. 1987. Nonlinear analysis to tubular truss

tower subjected cyclic horizontal force. Safet-,r Criteria in Design ofTubular Stntctures. eds. Y. Kurobane,

and Y. Makino: 47-56: Kumamoto Univ. , Kumamoto, Japan.

5. Kurobane, Y., and Ogawa, K. 1993. New criteria for ductility design of joints based on complete CHS

rn¡ss tesrs. Tubular Structures V. eds. M.G. Coutie, and G. Davies: 57G581: E & FN Spon, London, UK.

6. Kurobane, Y., Ogawa, K., and Sakae, K. 1994. Behavior and design of composite lattice girders with

concrere slabs. Tubular Structures V/. eds. P. Grundy, A. Holgate, and B. Wang: 69'76: A.A. Balkema,

Ronerdam, The Netherlands.7. Matsui, C., and Kawano, A. 1988. Strength and behavior of concrete filled tubula¡ trusses. Proc. Int.

Speciatry Conf. on Concrete Filled Steel Tubular Structures,.4SCCS: I l3-l19.8. Recent Research Developments in the Behavior and Design of Tubular Structures. 1994. Architectural

Institute ofJapan, Tokyo, Japan. (in Japanese).

9. Recommendations for the Design and Fabrication of Tubular Structures in Steel. 1990. ArchitecturalInstirute ofJapan, Tokyo, Japan. (in Japanese).

10. Kurobane, K., Ogawa, K., and Ochi, K. 1989. Recent research developments in the design of rubular

structures. J. Constuct. Steel Researcå l3: 169-188.I l. Ogawa, K., Kurobane, Y., and Maeda, T. t 995. Post-buckling behavior of circular tubular struts. J.

Struct. Construct. Eng., AH 475:137-144 (in Japanese).

12. Kurobane, Y, Makino, Y., and Ochi, K. 1984. Ultimate resistance of unstiffened rubularjoints. J. Struct.

Eng., ASCE I l0: 385-400.13. Kurobane, Y., Ogawa, K., Ochi, K., and Makino, Y. 1986. Local buckling of braces in tubula¡ K-joints.

Thin-Walled Structures 4: 234O.14. Kurobane, Y., Makino, Y., and Ogaw4 K. 1990. Further ultimate limit state criteria for design of tubula¡

K-joints. Tubular Structures. eds. E. Niemi, and P. Makelainen: 65-72: Elsvier, London, UK.15. Kuroba¡re, Y. 1989. Recent developments in the fatigrre design rules in Japan. Fatigue Aspects in Strucrural

Design. eds. J. Wa¡denier, and J.H. Reusink: 173-183: Delft Univ. Press, the Netherlands.16. Recommended Practice for Planning, Designing and Consrructing Fixed Ofrshore Plarforms. I 993. API

RP2A-LRFD. American Petroleum In stirute, rù/ash in gton DC.17. Kuwamura H., and Akiyama. H. 1994. Brittle fracrure under repeated high stresses. J. Construct. Steel

Research 29:5-19.18. Toyoda, M., Hagiwara, Y., Kagawa, H., and Nakano, Y.1992. Deformability of cold formed heavy gage

rectangular hollow sections: Deformation and fracture of columns under monotonic and cyclic bendingload. Tubular Structures V. eds. M.G. Coutie, and G. Davies: I43-150: E & FN Spon, London, UK.

19. Inoue, K., Ogawa, K. Tada, M., and Yanagihara. H.1994. Earthquake responses of member plasticdeformation of rigid frame with RHS column , J. Construct. Steel 2, JSSC: 9- I ó (in Japanese).

85

FIRE PERFORMANCE OF CONCRETE-FILLED TTJBUI"AR COLT'MNS

T.T. L¡e end V.IiR. Kodur'

ABSTRACT

The fire resistance performance of concrete-filled hollow stn¡ctural steel coh¡mns is presented

for tbree types of concrete filling, namely plain concrete, bar-reinforced concrete and fibre-

reinforced rooct te. Results from experimental and theoretical sn¡dies indicate that any required

fire resistance, in the practical range for most buildings, can be obtained for hollorr steel

columns through the three types of concrete filling. The important pararneters that determine the

fire resistance of ¡þs sqlnmns are discussed. A fire resistance design equation, zuitable forgeneral application and incorporation into codes is presented. Also presenæd is how the

ãesigner can select various parameters to satisfy fire resistance requirements.

KEllilORDS: Fire iesistancc design, HSS columns, Concrete-filled

INTRODUCTION

Steel hollow structural section (HSS) columns ÍLre very efficient structnrally in resisting

compression loads and are widely used in the constn¡ction of fr¿Ined structures in inúsnialbuilåings. HSS columns. like other structu¡al members, are to be designed to satisff the

requireãrenrs of serviceability aud safety limit states. One of the major safety reçirements in

UuitCing design is the provision of appropriate fire protection to structural members.' The basis

for this rcquirement can be attributed to tbe fact that, when other measures for containing the fire

fail. stn¡cn¡ral integrity is the last line of defence.

HSS columns are often filled with concrete in order to achieve increased load-bearing capacity.

Concrete filling also increases fire resista¡ce. Through the utilization of a concrete core,

external fire protection required for the steel can be eliminated, thus increasing the usable space

in the building. Further, properly designed concrete-filled hollow steel columns can lead, in an

economic ,""y, to the reaùzaiion of architectural and structural design with visible steel without

any restrictions on fire safetY.

For a number of years, the Nationat Fire Laboratory (NFL), Institute for Research in

Construction, National Research Council of Canada, has been engaged in sh¡dies, whicb were

supported by the Canadian Steel Construction Council, aimed at developing guidelines forthe

design and construction of concrete-fìlled HSS columns. Both experimental and theoretical

studies, using numerical techniques, were carried out to investigate the influence of concrete

filling on the fire resistance of HSS columns.

'National Fire l¿boratory. Insritutc for Research in Construction, National Resea¡ch Council of

Canada, Ottawa, Canarja KIA 0R6

86

EXPERIMENTAL ST I.JD IES

Test Soecimens

Fiffy-eightconcrete-filledHsScolumns*t*lî::1.1:*ttuttbvexposingthecolumnstofire'The corumns were of circurar and square cross sectioñs and wóre infilred with three tj?es of

concrete; nanrely, plain concrete (Pq; bar-reinfo¡ced- concrete (RC) a¡d fibre-reinforced

concrete (FC). No åtemal fire protection was provided for the steel.

A' corumns were 3gl0 mm long, from end plate to end prate. The ouside diameters (width) of

the columns varied from l4l mm to 406 mm and the walr thicknesses varied ftom 4'8 mm to

r2.7 mm. parameters investigate¿ ¡orjuJrl end conditions, concrete strength, load intensity,

aggregatc and reinforcement. -rigure l shows erevation and cross-sectionar details of tlpical

H-lS ãolumns fitled with three tlçes of concrete'

./mÅ*o*-*Þ gã. Ñ"^ ã"-

A4ntr¡ll.O 295Ír1

Figurc I Elevdion asd cfoss section of concre¡e-Frlled srecl columns t¡sod in firc Tests

(c) Cohíln(t) Colutîn PC (b) Coù.úûri FC

87

The hollow steel columns wcre filled by pouring concrete into the colu¡nn througb the topopening and vibrators $¡ere used to consolidate the concrete. The average 28day cylinderstrength of concrcte varied from24 to 49 MPa, while the corresponding sFength on the test day,

which was four months or more later. varied from 24 to 59 MPa.

The reinforcement for FC filling consisted of steel fibres, with the percentage of steel fibres inthe concrete mix being 1.77% by mass. For the RC-filling, lateral and Eansverse reinforcementwas provided according to CSA-423.3-M84 (Ref. l). The main bars and ties æ requiredspacing, were tied to form a steel cage which was placed inside the HSS coh¡mn.

Test Conditions

The tests were c¿rried out by exposing the concrete-filled gslumns to heat in a furnace speciallybuilt for testing loaded columns. The test funrace was designed to produce conditions, zuch as

temperature, sün¡cffal loads, heat transfet to which a member might be exposed during a fire.It consists of a steel framework supported by four steel coh¡mns, with the fi¡mace chamber

inside the framework. The hydraulic loading system has a capacity of 1,000 t. Full details on

the characteristics and instn¡mentation of the column furnace are provided in Ref. 2.

Most of the HSS columns tested were subjected to a concentric load. Only three columns were

tested for eccentric loads. The applied load on the colurnns varied from about 600/o to 140% ofthe factored compressive resistances of the concrete corc and about l0 to 45o/o of the factored

compressive resistances of the composite column calculated according to the specifications ofCSA/CAì.I3-S I 6. l -M89 (Ref. 3).

The load was applied approximately 45 min before the start of the fire test and was rnaintained

until a condition was reached at which no fi¡rther increase of the ærial deformation could be

measured. This was selected as the initial condition for the æcial defonnation of 1þs çshrmn.

During the test. the column was exposed to heating controlled in such a way that the average

temperature in the fumace followcd, as closely as possible, the standard temperature-time curye

of ASTM El l9-88 (Ref. a) or CANÂJLC-Sl0l (Ref. 5)-

The load was maintained constant throughout tbe test. The columns were considered to bave

failed and the tests rvere terminated when the hydraulic jack, which has a maximum speed of76 mm/min, could no longer maintain the load.

The furnace, concrete and steel temperatures, æ well as the æcial deformations and rotations,

were recordedat} min intervals.

Results

Full results of the fire tests on HSS columns, filled with PC, RC and FC are given in Refs' 6' 7

urd 8. Results from the fire te.sts indicate that the fire resistance of PC-filled HSS columns is

about I to} h, as compared to about 15 min for unprotected HSS columns. For RC-filled

columns and FC-filled columns fire resistances as high as 3 h were obtained.

88

The failure of the columns varied from compression to buckling depending on the size of thecolumn and the type of infill. The majority of the PC-filled columns failed by buckling.Buckling was significant in columns with sectional dimensions less than 203 mm. Generally,the failure of PC-filled columns was by sudden contraction, while RC-filled and FC-filledcolumns failed by gradual contraction.

The behaviour of concrete-filled HSS coh¡mns under fire conditions is illustrated in Figure 2,whicb shows the variation of the ærial deformation with time for the three types of concrete-filling (Ref. 9). These three columns had similar characteristics and were subjected to similarload levels. As expected, thc columns expand in the initial stages and then contract leading tofailure. The deformation in these columns results from several factors such as load, thermalexpansion and creep. rWhile the effect of load and thermal expansion is significant in the earlystages, the effect of creep becomes pronounced in the later stages.

It can be seen from the figure that the deformation behaviour of the FC-filled steel column issimilar. during tbe later stages of the test, to that of the RC-filled steel column. The initialhigher deformations in fibrc-reinforced concrete-filled columns might be due to the higberthermal expansion of fi bre-reinforced concrete.

NUMERICAL MODELS

Tbe main objective of the experimental studies was to generate fire resistance data for immediateuse by the construction industry and to provide information for the development of generalmethods of calculating the fire resistance of concrete-filled steel columns.

æ

û

EtoEc'ooaaE 'roo

toaE-

{o

Petm¡gFClll¡ttgnÞt¡l¡¡og

\*r,ta \

t00

Tlnq mlnutlt

Egure 2 Comparison of Axial Deflections forComete-Filled Hollow $¡¿sl ÇslrrmnsExposed to F¡rc

cu. llts(ssú r G¡)

Clr. Hgg

(!2. r C¡l

l+ HtCFAr6r,

Figr¡¡e 3 Conparison of Fr¡c Resista¡ce for Conc¡eæ-

Filled Hollow Steel Coh¡m¡s

89

Mathematical models were developed for predicting the behaviour of PC, RC and FC-filled steel

columns in fire (Refs. 10, I I , 12, l3'r. The steps associated in the developrnent of the models

involved the calculation of the fire temperatures and the temperature, deformation and strength

of the concrcte-steel composite construction. The calculation procedure was incorporated intocomputer programs. The validity of these computer prograns has been established by

comparing the predictions from the models to test datå. The models can accormt for the

important parameters that influence the fire performance of concrete-filled IISS columns.

The computer programs were used to carry out detailed numerical studies (Ref. 9) to compare

the fire resistance of HSS columns with three ttpes of concrete filling. The fire resistance ofsimilar circular and square columns, as obtained from computer models, is corryared for three

types of concrete filling in Figure 3. The fire resista¡ces ofthe PC-filled steel cohrmns aremuchless than the fire resistances of the RC and FC-filled columns. The fire resistances of the FC-filled HSS column is almost the same as that of the RC-filled HSS column.

Although it is possible to use the mathematical models for fire resistance design, the calculation

procedure is elaborate and requires considerable skill and effort. A method more nritable forgeneral application and incorporation into codes, is the use of design formulas in line withionventional design procedures. The development of zuch design equations for calculating the

fire resistance of plain concrete-filled HSS colunns, is illustrated in the following sections.

FACTORS IIVFLT,]ENCING FIRE RESISTANCE

The computer programs developed above were used to carry-out deøiled paranetric studies to

generate a large a¡nount of data on the fire resistance of concrete-filled HSS columns. The

lnfluence of various factors on the fire resistance of concrete-filled HSS columns was

investigated through computer-simulated fire tests. The effect of various parameters on fireresistance for PC-filled HSS column is presented in this section.

The influence of the variables wÍrs assessed by comparing the fire resista¡ces calculated for the

various conditions sn¡died, with that of a reference column (Ref. l4). For this pl¡{pose, a

column, with an intermediate diameter of 273.1 mrn a steel wall thickness of 6.35 mrU an

effective leng¡h of 2.5 m and siticcous concrete filling with a strength of 35 MPa" was selected

as a referencè column. Two refercnce loads were selected for the fire resistance comparisons,

nanrely I 150 kN which corresponds to a fire resistance of the reference column of 60 min and

330 kll which corresponds to a fire resistance of 120 min-

The influence of the various study variablcs is shown in Figures 49 and discussed below.

Outside Diameter of the Steel Section

In Fig. 4, the fire resistance of the columns is shown as a function of the steel outside diameter

for thi wo loads of 330 kN and I 150 kN. It can be seen from the figure tbat the colurnn outside

diameter. which is a measure of the column section size, has a major influence on the fire

90

resistance of the column. The curves in this figure indicate that tbe fire resistance increases

more than quadratically with thc column outside diameter.

ThickTress of the SteelWall

The influence of the thickness of tbe steel wall on the fire resistance of the columns is shown in

Fig. 5. It can be seen that, for thc smaller colum¡ diameters, the fire resistance tends to increase

Ñ, for the larger sizes, ro dccrease with increasing wall thickness. The influence of the wall

thickness is smãll, however, in comparison with that of the column section diameter. For

practical purposes, it seems warranted to neglect the influence of thickness of the steel wall on

the fire resistance of the column.

e,

E 160q;0E9 120g0oeE)t¡.

Ê25oEg;

Ê 20066

Et*o:¡¡ 100

toül (30 kN)l¡¡¡t (1150 kN)

100 150 ã)0 250 300 350

Outside diarnoler. mn

-ffi

-3s€-r,

3?arñn

T3Íñ

i ;il-ttt*l¡al EÍr

4681012Wallthidgless. mm

Frgure 5 Fl¡e Resistæce as a Function of HSS WallThickness

160L

0

Figr¡re 4 Fl¡e Resista¡ce as a Function of Column

Outside Dianeær

Lo¡d

In Fig. 6, the fire resistance of the columns is shown as a function of the load for tbe reference

co¡¡rirn, the smallest column and the largest column considered in the paranretric stt¡dy. For fire

resistances abovc 45 min. which lie in tbe practical region, the fire resistances of the sslrrmng

increasc sharply with decreasing load. Tbe influence of load on fire resista¡ce is relatively

higher for thè iurg.r columns. For the colu¡nn with an outside diameter of 406.4 mm' for

r*-.rpt.. a rcduct-ion in load of about 35% from 3000 lcl'{ to 2000 k}'I will double the fire

resistance of the column from approximately I to 2 h. For the reference column, which has a

diameter of 2ß.1mm, the loaðhas to be reduced by about 70o/oto double the fire resistance

fromlto2h.

Effective Leneth

In Fig. 7, the fire resistance of the columns is shown as a function of the effective length of the

colu¡in for the two selecred reference loads of 330 kN and I150 kN and two strengths of the

91

concrete filling, namely,20 MPa and 35 MPa. The curves show that in the range of effective

lengths of 2.5 to 4.5 m, the fire resistance is approximately inversely proportional to the

effective length.

The influence of the effective length is somewhat greater for low loads than for high loads. The

influence of the compressive strength, howeveç is relatively grËter for the higher loads. It can

be seen in Fig. 7 thaL for low loads and higbervalues of the effective length" the influence of the

compressive strength on the fire resistance of the column becomes very small.

'l¿10

120

_E 100É,

o't80g.98æotr 4{t

æ

0

300

?fi

Eæoo'(,c€ iso6gE rool¡.

50

o

O¡¡ilr d¡rll.r (ta1.3 rlrn,o|¡td¡tr d¡n* ln3.l ¡rút lO¡¡¡i¡a¡¡r: (4OA,llrrn)

æ00 ,O(Xt 60æ 8000

Load, kN

35l¡Fr

20lPr

3!lMP¡ rr\\\

'a2ori'r ______ì.

t¡d(3!!¡tl)t¡d(fisorto

O LO ZO 3.0 '+.0 5.0

Efieaivs len$t' m

Figr¡re 6 Flre Resisance as a Function of Load Frgr¡re 7 Flre Resis¡ance as aFrmction of EffectiveI-ength

Concrete Streneth

The influence of tbe concrete strength on the fire resistance of the column is shown in Fig. 8 for

the two selected reference loads of 330 kN and I150 kN. The curves show a moderate influence

of the concrete strength on the fire resistance of the column-

The influence of the compressive strength is greater for the higher loads than for the lower loads.

For the lower loads, the fire resistancã of the colum¡ increases by approximately 40Yo if tbe

concrete strength is roughly tripled and for the higber load by about l00o/o.

Tvoe of Aeeree¡te

In Fig. 9, the fire resistance of the reference colurnn is shown as a fi¡nction of the load' for a

siliceous aggregate and for a carbonate aggregate concrete filling. The curves in Fig. 9 show

that the nrãieslstance of the column filled with carbonate aggregate concrete is higher than that

of the column filled with siliceous aggregate concrete. In the practical regior¡ namely, for fire

resistances above 45 min, the diffeiencã in fire resistance between carbonate aggregate and

siliceous aggregate .on.rrt, filling varies from approximately 20o/o to 4O%- The difference in

fire resistañõe provided by the two tlpes of concrete tends to increase with lower loads or higher

f¡re resistances.

92

þad (330 kN)

Load (1150 kN)

cEooc,63tttt@

ol.L

250 I

I

200

150

100

ccdocao

Ilt@

@

l.L

äño 'so 2oo ?so 3oo

Goncrete strength' kN

Frer¡re I Ftre Resistance as a Funcüon of Concrete F¡sure n ff"ffiï^ï#"#fi"åiJ#åt"tSuength

DESIGN EQUATIONS FOR FIRE RESISTANCE

Based on the data from the parametric studies, expressions were develo¡ed for the calculation of

the fire resistance oi circ,rfa, *a ,quur.îa's .äturnor fiiled with pläin coDcrete' As shown

above, the most important parameters ,ú ñt-tne tue fire resistanõe of hotlow steel columns

filled with Plain concrete are:

o The outside diameter or the outside width of the column

o The load on thc column

o The effective length of the column

o Concrete strength

. Type ofaggregate

Based on the relationships between_the fire resistance and the above paraneters, found in the

paramctric studies. thc iollowing. ro*ulu for the fire resistance of pc-filted Hss colurnn

ffiËä;";i;iild;s, *" "iãbtitt'"d empiricallv (Ref' I5):

R=rffi;o'

where:

R = fire resistance in minutes; f. = specified zL-day concrete strength in MPa; K = effective

length factor; L = unsupported length of tbe column in mm; D = outside diameter of the column

(l)

93

t siliceous aggr€gar'

\ ----- Carbonate aggrsgats

IìI

in mm; C = applied load in kN; and f, = a constant to account for the t¡pe of aggregate and the

crcss-sectional shape of the HSS column. For circular columns, the value of f, is equal to

0.07 for siliceous and 0.08 for carbonate aggregate concrete, while for coh¡mns with square

cross-section, the corresponding value of {, is 0.06 and 0.07 for siliceous and carbonate

aggre1ate concretes, respectively.

Equation (l) is deemed to be applicable when the following limits are set on the parameters thatdetermine the fire resistance of the column:¡ Loads are not greater than the factored resistance of the concrete core deterrrined in

accordance with CAN/CSA-SI6.I-M89 [Ref. 3].. Firc resistance not greater than 2 h.r Specified compressive strenglh of concrete at 28 days in the range of 2040 MPa.

r Effective lenglh of column (KL) in the range of2000-4000 mm.o Outside diameter (width) of the column in the range of 14&410 q¡n (l't0-305 rnm).o Width (D) to thickness (t) ratio not to orceed Class 3 section according to CAN/CSA-S.16.l-

M.89 (Ref. 3).

In the above equation, the fire resistance is expressed in terms of structr¡ral desig parameten¡.

This offers a convenient method of integrating the fire resistance design with stn¡ctural design.

Using these equations, a designer can arrive at a desired fire resistance value by varying differefitstn¡ctural parameters, such as length, load, diameter (width), and concrete strenglh. The r¡se ofthese equations leads to an optimum design that is not only economical but is also based on

rational design principles.

ER3ãFgg3ããEfætiva t¡ngü, KL mm

Round Hollæ Sleel Columns't h Fire Resistarice

Souare Hollov Steel Colum¡rs'' thFireResistanco

Figr¡re l0 Fre Resistance Design Graphs forconcreæ'Fiued Hollow coh¡mns

94

I h BstlngTpo S ConctÊttl'(¡S.Cr r:a t-.,rrs¡s¡-ã€,,""**t<r^rn

.",eli RELù

FR:ËËsB3ããEísctit€ L€îgñ KL n¡m

The fire resistance equations evolving from these studies are incorporated into the National

Auitaing Code of Canäda (NBCC, nef. tO). In the NBCC, Equation (l) is rearranged so ¿ls to

calculate the mærimum load carrying capacity, c*, for the required fire ¡esistance rating of a

pc-filled HSS column. In order to make the âesign process simpler, the NBCC contains design

charts, for different fire resistance ratings, wherãin C* is ploned as a function of effective

lengttr for various column dimensions and concrete strengths'

Figure 10 shows two such design g,uphs for circular and square pc-filled HSS columns and

having a I h fire resistan"" rriing. For hollow structural sections commonly available in

C*"å, the C-, for concret" ,t "rrgths

of 30 MPa and 40 MPa can be read from the design

charts.

SUMMARY

Concrete filling offers a practical solution for providing fire protection to hollow structural steel

columns without -v .*"-al protection. Results from the experimenø] and numerical studies

indicate that any "-oun,

of fire resistance, in the practical range for building constn¡ction, can

be obtained for HSi columns through three t¡pes óf concrete filling. The use of fire resistance

design equations reads to an optimim design that is not only economical but is also based on

rational design PrinciPles-

REFERENCES

canadian Standards Association. 19g4. Design of concrete structures for buildings'

CAN3-423.3-M84. Toronto, Canada'

Lie, T.T. 1980. New facility to determine fire resistance of columns' canadian Journal of

Civil Engineering 7(3): 551-558'

Canadian Standar¿s Ássociation. 1989. Limit state design of steel stn¡ctures- cAN/cSA-

S 16. l -M89. Toronto, Canada'

American Society for Testing ar¡d Materials. 1988. Standard methods of fire tests on

Lritaing ro*t u.iion and matãrials. ASTM El l9-88. Philadelphia, P{, USA'

Underwriters' Laboratories of Canada. 1989. Standard methods of fi¡e endurance tests of

building construction and materials. CAN/IJLC-S l0l' Scarborough' Canada'

Lie, T.T.; and chabot, M. 1992. Experimental studies on the fire resistance of hollow steel

columns filled wittr;i;ir concrete, IRC Internal Report No. 6l l, National Research Council

of Canada, Institute ior Research in Construction. Ottawa, Canada.

chabot, M.; and Lie, T.T. 1992. Experimenøl studies on the fire resistance of hollow steel

columns filled with bar-reinforced concrete, IRC Internal Report No- 628, National Research

Council of Canada, Instirute for Research in Construction' Ottawa, Canada'

Kodur, V.KR.; -¿ lù, f.f. 1995. Experimental ludies on the fire resistance of circular

bollow steel columns frlled with ,t.ól-fibr.-reinforced concrete, IRC Internal Report

No. 6g l, National Research council of canada, Institute for Research in constnrction'

Ottawa, Canada.

l.

,)

3.

4.

5.

6.

7.

8.

95

g. Kodur, V.KR; and Lie, T.T. 1995. Fire resistance ofhollow steel coh¡mns filled with steel-

fibre-reinforced concrete, Proc. Second Univemity-Indr¡stry Wo¡lahop On Fibre Reinforced

Concrete fuid Other Composites. Toronûo, Canadæ 289'3V2-

10. Lie, T.T.; and Chabot, M. 1990. A method to predict the fire resistance of circula¡ concrete

filled hollow steel columns. Journal of Fire Protection Engineering 2(4): lll'126.ll.Kodur, V.ILR.; and Lie, T.T. 1996. Fi¡e resistance of circular steel coh¡rnns filled with

steel-fibre reinfo¡ced concrete. ASCE Journal of Strucn¡ral Engineering (in press).

lZ.Lie,T.T.; and lrwiri" RJ. 1995. Fire resjsance of rectangular sæelcoh¡mns filld with bar-

reinforcedconcrete. ASCE Joumal of Strucn¡ral Engineering l2l(5): 797'805.

t3.Ifudur, V.KR; an¿ tie; tt: Performance of concrete-fitld steel coh¡mns etçosd to fire.

Joumal Of Fire Protection Engineering 7(2): l-9-l4.Lie,T.T.; Invin, R.J.; and Cbabot, M. 1991. Factors affecting the fire resistance of circular

hollow steel colurn¡s filled with plain concrete. IRC Internd Report No. 612, National

Resea¡ch Council Of Ca¡ada, Institute ForResearch In Constn¡ction. Onawa, Canada.

15. Lie, T.T.; and Sfinger, D.C. 1994. Calculation of fire resistance of steel hollow structural

steel columns filled with plain concrete. Canadian Jor¡rnal of Civil Engineering 2l: 382-385.

16. National Building Cods of Canada. 1995. Appendix D. Fire Performance Ratings.

National Resea¡ch Council of Canada Ottawa' CaraÅL

96

ABSTHACT

The following aspects of tubular offshore structures are covered in this paper:functionality, -fabri'cation & erection sequence, early failure/survival lessons, designforces, simþle joints, fatigue, fracture, weiding, inspection, bigger & better, and structuralintegrity.

INTRODUCTION

Offshore structures are usually designed by teams of e_ngineers, involving severaldifferent technologíes. Althougñ the téam leàder is usually a structural engineer, thefollowing other spebialties are álso ínvolved:

environrnental loadings (wind, wave, and current),oil field operations and topside.safety considerations,economic venture and risk evaluation,foundatíon design (e.9. laterally loaded piles),construction oPerations,inspectíon and repair.

Design is usually governed by considerations other llan in-place gravity loading.Conitructíon opeiaiíons often dÏctate the layout and architecture of thes_e^lqçê lubularspace frames,'which may be transported and launched in modules of 50,000 tonnes.Guidance for þlanning, designing, and construction fixed gffs.hgre platforms can be foundin API RP 2A, which-is the deJacto international standard, incorporated into the firstedition of ISO DIS 13819 Part2. Marshall 1992 gives a broad introduction to the subject.Many other key references can be found in prôceedings of the Offshore TechnologyConference (OTC).

FUNCTIONALITY

The most common type of offshore platform is the fixed, steel, pile-supported qtructure.Over 3,000 of theseÏave been built-worldwide, in water depths up to 400-m. These arepermanent structures, built to support up to 60 oil 4 gq.ç wells, together.with theässociated drilling and production e{uipmerit, over a servicé life of several decades.

TUBULAR OFFSHORE STRUCTURES

by Peter W. Marshall

' MHP Systems Engineering, 1711 Woodland vista, kingwood, Texas 77339

97

(713)358 641s

Mobile offshore drilling units, such as jack-ups, semi-submersibles, or ship-shape rigs,are used for exploratdry Oritling (to proVe oil ilep.osits.large enough to justify permanentplatforms), or tb drill isblateO éâteli¡te wells which will be tied by pipeline to a nearbyplatform.

FABRICATION & ERECTION SEQUENCE

Fixed offshore platforms consist of the following major elements:1. jacket2. piling3. deck

The main structure is a welded tubular steel space frame, also calted jacke-t or template,which extends from the seafloor to just above the water surface. This is designed to. reslstthe lateral loads imposed bV wind, wave, and current, as well as vertical gravity loads.The jacket is assembled oñshore, usually lay.ing on its side. Tf¡e tubes are customfabrióated to size, and wetded together iñto ilafplanar bents. These are lifted into avertical position a¡iO t¡eO together w¡tn aAO¡tional bracing to complele.the space frâme.. ltis then d¡<¡OOeO onto a bargè in one piece, torryed offshore, launched at sea, and set 9n thgsea floor by ballasting, often with the assistance of a large seagoing crane or derrickbarge.

The platform foundation is established by driving tu.bular steel piling through the.jacketlegs (or in deep water, through sleeves wh¡ch exténd only a sh.ort distance above the seano-o4. The pilðs penetrate 3b to 120-m into the sea floor, and are attachedJg t|1e jacketlegs'by weläing åbove water (or to the sleeveg.by gtoyling.the annulus). Veûical and,ovértuining loals on the struiture are resisted b.y axial loads in the piling. Lateral.andtorsional lõads at the base of the jacket are carriled into the soil via portal action of thelaterally loaded piles.

A superstructure, or deck section, is set gn lop to complete the structure. lt carries thefunctional; loads ior which the structure is built,'keeping-men and equipment out 9f harm'sway, above the waves. The superstructure is.typicâlly a composite of tubular, plategirder, and wide flange beam and truss construction.

EARLY FAILURE / SURVIVAL LESSONS

The first steel offshore platform in the Gulf of Mexico was built in 1947. Its constructionwas described in the motion picture Thunder Bay, slarnng Jimmy Stuart. ln.the hap.p[ànding, he defuses environméntal opposition, sulvives a hurricane, gets a gusher, and ofcourse the girl.

Hurricanes Audrey (1957) and Carla (1961) caused great destruction and death onshore,but only minor damäge óffshore, aloirg wiÎh usefulðgta-pl pile performance and wavefoices. 'Hurricane Hilða (1964) cáusedihe failure of 22 offshore piatforms, many.of whichcompletely disintegrated due io failure of their tubular joints. No lives were lost becauseof a'políóy of delmanning the platforms when theie was. warning. of an imminenthurricäne.'However, the eniuing investigations led to the modem punching shear.criteriafor tubular joint desi(¡n, and the fìrst use óf improved steels in joint cans (Carter, Marshallet al 1969).

98

lmplementation of these _de.sign ìmprovements led to platform designs with considerablereserue strength (Bea & Marshall, 1976). ln hurricane Camille (1969), a platformdesigned tor 57-ft waves survived an 80-fl wave. ln hurricane Andrevù (1ggj), a'numberof older pç-] 964 platforms were again weeded out; more remarkable, howeúer, was thesurviv.al of platforms which were exposed to 2 to 3 times their allowable desijn loads,including some which were completeiy overtopped by waves (see OTC 7470-74i5).

DESIGN FORCES

9nSq a platform site has been selected, experienced specialists should be consulted indefining the met-ocean conditions from whibh operatinçj and extreme design criteria willbe drawn.

Wind forces are important to designing above water porlions of the platform, and for thethe drilling and production equipment. A turbulent atinospheric boundary layer near thesea sudace has a rather co-mplex structure in space and time. Wind speeôs increase withng¡gl,t above the w_ater surtacg,. and gusts can be up to 1.7 times the hourly mean speeà.wind contributes 10 to 20"/" o'nthe totál lateral load ón a pratform.

Empirical relationships for estimatìng significant wave heights, given the wind field, weredevelo.ped during the second world war, to assist in planñing a-mphibious landiñgË. l; ãnatural sea state, wave heights vary ¡q1Qomly, as. de'scribedly tfie Rayleigh d¡stiiOution.The.most likely extreme wave out of .100_0 w_aves (about a 3-hour storm) Èi.eO1imeînäslgnificant wave height, or 3.7 times the RMS water surface fluctuation.

Water particles ín deep water waves travel in circular orbits, rotatinq in the direction ofwaYe travel, with.the, magnitude of motion decaying.with depth. Horizóntal velocity peaksat the wave crest, while horizontal acceleration-an-d pressuie gradient peaks So-rJegreeé(1/4 wave lengtl¡) ahead of the crest. For steep extieme storÉr waves, n¡gh óioði"wã"etheories,. e.g. Stokes Sth, must be used to.describe water particÍe üelocities andaccelerations, requiring the use of a computer.

ln addition to wave action, tidal currents, wind-driven currents, and ocean circulationcurrents contribute to the total water particle velocities.

When a vertical cylindrical (9.g. jactet leg).is subjected to a horizontal pressure gradient,lateral forces analog,ous to buoyancy result; furtfiermore, since the boäy partiall"y ¡tockjthe flow, an "added mass" eff-ect creates additional forces in phase'r,iritn ü¡é lateralpressure gradient and water particle acceleration. A turbulent ivake behind the bodvcreates a drag force. which is proportional to water parlicle velocity squared. lt has beeílempirically.observed that reasonable-design forces äre obtained by suþerimposingthesàeffects, us.!g the'Morison eguation for eãch incremental length ót eácn mãmOei ñ théplatform. The drag and inertia coefficients in this equation hãve been calibrated on iuilscale wave force measurements. Forces increase in the presence of marine growth.

Although there are .many,-computer -programs for analyzing space frames, specialfe_atures,are required for ófrshorb platfbrmé. Wave theory'is uËualty integráteã'w¡if' nestructural analysis, to avoid having io manually transfer huþe volumes of däta. Distributédgra)/lty, þuoyancy, and wave forces along the members are collected into fixed-end forcesand moments at the. nodes, prior to solving the. structural matrix. During stress recovery,these same distributed forces must be recälled in order to get correct ËenOing moreÁídalong the entire length of the members, as the critical momeñt is often near mij-spil. -

99

Since the behavior of the laterally loaded pile foundatio¡t is highly non-linear, specialtechniques are required to actiieve compatible solutions for both structure andfoundation.

ln some areas, otfshore platforms must also be designed for the effects of floating sea ice,earthquakes, or mud slides.

SIMPLE JOINTS

ln offshore structures, most of the structural connections are tubular io¡nls involvingcircular tubes. These have been extensively reviewed by the Underwater EngineeringGroup (1985) and by the present author (Marshall 1992a).

Most connections are made by simply welding the-branch member (e.g. bracing). to- themain member (e.g. jacket leg). ln the US, welds in T.-, Y-, and K+onnections, made fromone side withoi¡t óa'cfing acõbrding to AWS prequalified practices, may be presumed.todevelop the full strengthõf members joined. _Mogt design problems aripe, not in the weld,but in ihe main merñber, which muðt transfer loads from one member to another viaóuncning shear and shell bending stresses. A locally lhickened section of the mainñ.tËrOãrlol¡oint can, is usually prÑiOeO for this purposé. Typically, this is about twice asthick as the attached braces.

The static strength design of such connections is described in a companion paper(Marshall 1996) ãnd the Aþpendix thereto (Marshall 1989).

FATIGUE

Fatique may be defined as damage which results in fracture after a sufficient number ofstreõs fluctúations. Fatigue performance or capacity.is characterized by S-l¡_99rves, plotsof total stress range (óeak+o trough) versu-s cyðles to failure (at say 97o/o lur_vival).Referring to Figuré l,'fatigue anatysis for offshore structures includes the followingelements:

(1) Long term wave ctimate is the starting point for estimating the demand.side of cyclictbáOing.'fhis is the ensemble of all seã states occurring yearly (or for the structurelifetime).

(Z) Global space frame analysis is perfo-rmed to obtain structural response in terms ofcybtic membér stress for each sea state of interest.

(3) Geometric stress concentrations at all potential hot spot locations.within the tubularòonnections must be considered, since fatigue failure initiates as a localphenomenon.

(4) Accumulated stress cycles are then counted, and applied against suitable S-Nòúrves, using Mine/s rule of cumulative damage.

100

Filt?:"r"?'f#ffi å'3ål:?"î,î"1"?å1'J3,ï'å::?Ëül:!'-ä,"'Fl"f;l'!Éy"l'iü'd3äläiriãtã'¡å1'r,"'iõt"t ranle to stíess or strain, on the outside surface of the intersection

ìi;Ëä'äùoulo'o'" rããrrrãð artei shakebown by a. strain gaugg..adjacent to and

oeroendicutar to tñã ¡ntJrsection weld. Hot spot stress places mãnyïifferent connectionäffiäi;i"î.oñ ä-cöärrrón uat¡s. tre rhicroscopic l''o!.cþ effecls, .metallursica.lËääãåiñ;, jib ¡nc¡p¡enf ciácks at the toe of the weld'are built into the S-N curves, which

ãi"'brpír¡ãä¡ry OaseO on a large data base of tests on as-welded haró¡vare.

Hot spot stress can be derived from nominal member stress using a parametric SCF

lèireãä concentration factor) formula, such as Alpha Kellogg:

SCF-1.8atsin0vt

where T ¡s the ratio of branch to main member thickne.ss, 0 is.the angle between member

aré, T is ma¡n rLr¡ãi mdius/thickness, and a is thg,oyqll-=ing parameter 99 given in

ÃwS ol.l-go r"Ër":ã.e (i:o for k, t.7 lor'TN,2.4tor X. 0.67 toi lpg. and 1.5 for.-oPÐ-

Âúèmãt¡uety, cr rãv oË ,èø q"giqqtaJe effectivg.cygl¡ç punching shear for use with S-N

tilÑåJxt aírd t<z (éee Note 5 of AWS Dl .1-96 Table 2-6).

Figures 2 compares Atpha tfllggp with othe.r morg scPhlsticated.parametric formulae

áäanãVtical ápprããctiãslo Sdr_"té,g. finite elemg!'¡l). ftre ordinate is the ratig of þo!.spot

"rõ"d õ'Ëñ;ñ¡ís lnË.i Èigrre s stïows how well tiot spot strg¡s, calculated with Alpha

Ëiãdúãáiòts i"t¡guã peñormance of tubutar conneit¡ons. lt does about as well as

méà;-uîdd not spot êiie!ð ¡n the original database, with similar scatter and bias on the

safe side (Marshall 1993).

The foregoing sc F reflect the overall ge.ometry oj lhe tubular connectiot¡I¡:TgÏ["notãf,lftËct o:t tne weld itseff is not expäcitly cal-culated, it does enter inlo the cholce ot s-ñãruË. rtiJ upp"iðurväãxl áo kf appiy to joints in which the weldsnersg il-Tihyivrti irrð aojoiniñg tase meial, for joints niifn oranch members up to.2S-mm thick. Lower s-Ñ';ñ;;-åpöiy-¡itiã *elOs'are not sc profiled.. For thicker'weldments, a size gfecJlào-ùóúóñ ¡riial¡glä ãtiåáõn ãppries. Th'e combined effect of thickness and profile i:;hõñ iá riguie?. This salme apþróach to fatigue, SCF, and S-N curves appears in API

RP 2A and ISO DIS 13819 PartZ.

FRACTURE

Most fracture control problems in offshore structures occur in the tubular connections or

nodes. As these "p-prãrén

inéi utt¡rate strelgth, tþe hot spot.repiqp exoerience triaxial

stresses, loca¡iz$Ëldil'g; iüiy Ëi"rtíð ðñ"lr "uónding, f+tg" defiect¡on.eifects, and.load

iå¡ði¡¡îtión. rnesé occunences in the presence of ùeld tóe notches place extraordinary

õ-rãã; on the ;"t;ñ rõrgf'ness of thä main member at tubular cohnectio¡.s.Typi$ld;¡d p*e¡"" ¡siã,]åe-ñ¡gi quality heat treated steel for the joint can, e.9. API Speç 2H'

2W, or2Y.

101

Conventional practices for the control of brittle fracture are based on Charpy jmp?cltestinq (AWS D1.1-96 sections 2.42 and C4.12.4.4). These are admittedly qualitative, butmay O'e correlated empirically to more definitive.éngineering^lqProaghês, such as theNRL fracture analysiö diaqiam (Carter, Marshall et al 1969). ln order to avoidcatastrophic propagätion of s-mdl ciagks at stresses approachi¡g the UT.Ç, anÇ to prwidecrack ar'rest for lbcá brittle zones in the H{Z,joint cans and other critical locations shouldhave the Nil Ductility Transition 30-deg-C below the Lowest Anticipated ServiceTemperature (NDT below l-ASÐ. ln addltion to this high level of notch toughness,struciural redundancy is used as a secondary level of defense.

Weld metal and heat atfected zones should have notch toughness requirementscompatible with the base metal in which they occur, enforced via consumable selectionand procedure qualification.

More advanced fracture control procedures, e.g. dA/dN and OTOD , are described inMarshall (1990).

WELDING

ln the USA and most of the rest of the world, bent fabrication is the preferred method ofãsðemOty for offshore jackets (Marshall,.1984). The intersection welds in Tl/- and K-ðonnectións are made from the òutside only, as the entire brace, point-to-point, is broughtinto position. AWS D1 1-96 section 2.39.2.2 describes.prequalified gogrplete iointoeneiration oroove weld details welded from one side witl"rout backing in Tl/- and K-bonnections.-The joint geometry and welding position vary.continuously _as o.le prgceedsaround the connéction with çjroove dimensions being defined as a function of localà¡hedral angle. Braces áre give--n a saddte-shaped cope éo that the lD weld root conformsto ne OD-of main membär, with a properroot gap all around. Special.proc_edurequalífication requirements, inctuding sainpie joints.o-ia tubular nlog|-uPr are described inÄWS D1.1-96 dection 4.12.4. Speõid wéldei qualifications, includingthe 6GR test andacute angle heel test, are presö¡bed in AWS.D1.1-96 sections 4.26 (5) and 4.12.1-2,iespectivé[. Less oneroui provisions for. partial penetration and fillet welds are alsogivån; thesê are particularly useful tqt-"_ta!!"ally loaded trusses in onshore applications-.Éárliér (g7Z-94) editions'of the AWS Codé had these tubular provisions groupedtogether in Chapter 10.

ln the nodal method of fabrication as practiced in the UK, nodes are prefabricated qs¡f.gpressure vessel practices, including'repositioning the work piece, welding.from bothb¡des, and PWHÏ. However, this is much more onerous and expensive tha-n the practiceOescúOeO in AWS, and single-sided closure welds are still required as the nodes are¡ñãrp"r"ted into the space irame, and service failures emanating from root defects in theclosure welds have been rePoñed.

INSPECTION

Three nondestructive inspection methods are routinely used on fabricated structures.These methods include visual, ultrasonics (UT), and radiography (RT). The magneticoäñi"1" inspectíon technique (MT) and thd liciuid penetrant ieihnique are generallyðonsidered as enhanced visual inspection techniques.

102

.1

All these techniques have procedural requirements which should be followed if they areused. An approüed quality control plan,.with procedures for each inspection method,should'be developed for each job application.

Visual. Visual inspection is always required. The visualtechnique is used either by itselfor as an integral pãrt of other ND-E techniques. Visual inspection should.be co.nducted..byqualified inspectórs, for inspection of w-orkmansþlp anO technique prior to and during theri,èlOing proéess, and for in'spection of final weld for completeness, size, contour, cracks,and other discontinuities.

Penetrant technique. The liquid penetrant inspection technique (PT) is useful fordetecting discontinuities such as craiks, porosity, etc. that are open to the surface.

Magnetic Particle Technique. The magnetic pafticle technique. (MT) is useful fordeiõcting discontinuities that are open tolhe su'rface o.r.a.re- slightly subsurface. TheoroceOuie for MT should conform to ä written procedure which follows ASTM F-709

ãnO npl RP 2X (third edítion, when issued), ôr similar national standards which providegu¡ãance specifiòally for the inspection of as-welded components, including provisions forihe resolutibn of indícations by light grinding.

Radiographic Technique. The radiogrqphic technique {RT) is useful for detectinqburied-or'thru-thicknesð discontinuitíes ¡h butt welds of simple geometry. The RT

þrocãàures in AWS D1.1 cover qualification of insp.ectors, standard practices andïechniques, image quality control via penetrameters, film and source tYPgs: geometriclimitatións (e.g. õOgé bloóks), and disposition; as well as providing.appropriate separatecriteria for non-tubu'iar static,'non tubuiar dynamic, and tubular structures.

Ultrasonic Technique. The ultrasonic technique (UT) is also.usefulfor detecting buriedor thru-thickness dis'continuities, and is particularly useful in identifying and.s.izing planardiscontinuities. lt is the only method apþlicable to internal inspection of welds in tubularT/Y and K connections, due to their complex geometry.

All UT should be in accordance with an approved written procedure which describes theappl¡caOle range of geometries, acceptäñce criteria for each type and size of weld,sbäcific UT insirumeñtation, transducdr characteristics (frequency, size, shape-, beamáñgle, etc). surface preparation and couplant, calibration test block and referenceiétieciors,'instrumeni cá¡¡bration methodé and interval, base metal checking,. ygldgeometry determination (e.g. indexing root location), scanning pq¡grn and sensitivity,t?ãnsfer'correction, coriec-tion for õurvature efféct on skþ distance,, method.ofO¡scont¡nuity length ând height determination, and protocol for defect verification duringexcavát¡on ánd répair. Sepairate procedures for tubular and non tubular structures shouldbe considered.

ln addition to the usual national certification schemes, UT technicians should be requiredto demonstrate their ability to execute the full scope of these testing procedu.res, using ap}ãciical test or mock-úp which incorporatei.weld types, local dihedral .angles,ih¡cknesses and discontini¡ity sizes of inierest. Their pedbrmance assessment shouldconsider false alarms as well as defects found. Aòceptable level of performance(pio¡ãoii¡tv ótã"1eðtioñ) shoutd be evatuated in the contexi of structural reliability issues,e.g. fractuie criticality vs. structural redundancy.

103

The foregoing procedure and qualification requireme¡ts,- as well as repofting of results,should bé in the context of applicable standards. Techniquqs a¡d reject criteria aredifferent for non-tubular (AWS D1.1-96 section 6?6)_an4-tubular (section 6.27)applications. Other applicable standards include API RP 2X tor tubular structurescöñstructed by the beni or point-to-point method, and A.Çft{E for prefabricated nodeswhich are welðed from both-sides and stress relieved as if they were pressure vessels.Note that for API RP 2X, the user defines the accepUreject criteria according to theservice requirements of his structure.

Reject Criteria. For simple (unstiffened) tubular joints in bent-fabricated structures, theweids are made from oné side without backing. Fortunately, the hot spot areas at tubularTll and K intersections occur at the outside surface, with reduced stresses at the root ofthe weld. ln view of the difficulty and undesirability of repairing innocuous root defects inthis situation, both AWS D1.1 añd API RP 2X provide separate criteria for the root area ofwelds in tubular Tl/ and K-connections. These allow somewhat larger discontinuities,based on experience-based fitness-for-purpose consid.erations (Marshall 1984a)..Nosuch relaxatión is allowed for the root area of butt joints (i.e. closure welds), nor should itbe applied at footprint crossings in stiffened nodes.

ln the acute angle region of simple T/Y and K-connections, the first root passes are. sonarrowly confined thai sound quality weld cannot be assured. These are.designated.asthe "ba-ck-up" weld, and excll¡ded from the theoretical weld throat. Nondestructiveinspection is'not applicable to the back-up weld, any more than it would be to the root landin a partial penetration weld.

BIGGER AND BETTER

The worlds deepest fixed offshore platform is "Bullwinke," in 490q water depth in the Gulfof Mexico (Digre et al 1989). ln water deeperthan 400-m, floating platforms arè-beingintroduced'tol¡ll the traditiónal role of fixed platforms, such as tension-leg platforms,spars, jumbo semi-submersibles, and turrèt-moored ships. -fhq"g are high-techväntureõ, with higher unit costs per well or per ton of payload than fixed platforms. Wherethere are targe -numbers of w'ells and high payloads, th-e ec_o_n9.mic.s of scale makecompliant towérs (which share many desqablecharac-teristics with fixed platforms) viablein water depths up to 900-m (Marshall & Smolinsk¡ 1992).

STRUCTURAL INTEGRITY

Designers must be involved in assuring lifetime structural integrity for his designs, forseveial reasons. They know the structure better than anyone else, which parts oreredundant of secondary, and which parts are primary of critical, as well as assumed levelsof performance (e.g.'ófìo¡ce of faiigue S-N curv-e in relation to weld profile. To Þqcorhprehensivety'in-control of his pioject,.the designer must be involvéd in materialseleötion, welded joint design, welder and proceduie qualification, fab.rication qualitycontrol, ând inspâction. These issues háve a humän side as well as technicalconsíderations. Bôth receive detailed coverage in the AWS StructuralWelding Code.

144

SUMMARY & CONCLUSIONS

This oape r can onry give a very brief introduction to the subject of tuburar offshore

;iñtüä.:'nä r"läiänËei which iollow provide more depth.

BEFERENCES

1.APl(1993),FìecommendedPracticeforPlanning,Designing,andConstructingHxedoffshoreptaflorms, epi-nÞ zÄ-inro ano epr nÞ e¡-wsD,-Ame¡cän Èetroleum lnstitute, washington Dc

2.Bea,Fl.G.andMarshall.P.W.(r9]Q,-fa!greModes.forotfshorePtaflorms'Proclstlntlcontoneãñåùórt ot ón-snore Plaflorms, 8055-76' Trondheim

3.carter.R.W.,Marshall,P.W..-e!qll]9-89),MaterialsProblemsinoffshorestructures'Proclstõtrtñoie tecir cont, Houston' oTo 1043

4_ Diqre, K.A.. Brasted, L.k., andlvlarshalt, p.w. (1989), Design of the Bullwinkle Plafform, Proc

OtfËñåte f"ch Conf, Houston, OTO 6050

5.lso(1994),DrafilnternalstandardlsolDlsl3Slg,PetroleumandNaturalGaslndustries-Offshore Structures, Part 1: eãnãrãl Requimments, and Pa¡ 2: Fixed Steel Structures'

lntemat¡onal Standards Organization' London

Marshall, P.W. (1984), connections for welded Tubular structures' 11W Houdremont Lecture'

Peçamon Press, Boston

Marshall. P.W. (1984a), Experience Bâsgd Fitness-lor-Purpose ultrasonic Reject criteria for

Tubutar Stwctures, präÀñõWñönrrróãnt on FFP in Welded Construct¡on' Atlanta

Marshall, P.w. (1989), Designing Tubular connections with AWS Dl '1, welding Joumal' March

Marshall, P.W. (1990), Adyq¡ce Fracture control Procedures for Deepwater otfshore Platforms

ãnO Compl¡ant Towers, Welding Joumal' Janualy

Marshalt, P.W. (1992), Offshore Stluctures, Chapter Q'A ¡1 Constructional Steel Design:'an' i Åiàäîii) i

"i' G u ia ;, eËävie r Ápp tied science Publishers, London.

Marshall, P.W. (1992a), Design of Wetded Tubutar Connections: Basis and IJse otAWS 01'1'

Elsevier Science Publishers' Amsteruam

Marshall.P.W.(1993),APlProvisionsforStressConce'ntraliorrFactor(ScÐ'S.NCurves'andõìlåipr.]ii¡å Ët¡äòs,-p-i* offshore Tech Conf, Houston, oTc 71s5

Marshall, P.w. (1996), welded circular Hollow Truss Connections, Proc AWS lntl cont on welded

Tubular Structures' vancouver

Marshall, P.W. and smolinski, s.L. (1992),The Mother ol All Rqsjlient structures: Fixed Base Tower

in 3ooo-fr water, and'ðä;äö;Ëia;ãrõ rrä"áålËióbËn nsÓE conf on civil Enss in the ocean'

Austih

uEG (1985) , Design of Tubular Joints for offshore structures (3 vols), underwater Engineering

Group, ClRlA, London

7.

8.

o

10.

11.

12.

13.

14.

105

I

lI

I

30 r ,750BRAN.CII

(a)

GLOBAL

.750', wALLBRANCH

(b)

CONNECTIONAS A STRUCTURE

-50 KSIG) HoT sPoT

NOTCH STRESS at

(c)

LOCAL ORMICROSCOPIC

go xs¡ G¡Hof SPoT c6

1.50'WALL'JOINT CAN

Ft 4u Kg

o:ron(f-lIzzrrlU)an-{õail

n-l

o

HOT SPOT STRESSsc: /r SIN e

€>---l -pI;¡C)>OX3rnTl r

=br.i c>UC)p

IN CHORDa Hs/vp

20

108ø 4.00

t¡l¡rlcta

û74q

108ø 2,75

ñæ

n

à

c)ïof,)(f

ou,q

2\---

%.

\t

\

108ø 1.00

H l-avt o(v lr'tþHÉx

l-- ub v,Øu|tOt¿l fftÉ.t-- o4v, ttl

ÀþonlL lr,Ut l-5l- o.o¡c:coU

oæ

(scF)chord ' | '8 d' f 'ff sln g wlrH ø( = l'o(scF)b'c..1.0 + 0.6 q,.

Ir0+{F'lscr).¡'o¡a]r_t,--ì¡[ôr-Â

Þatlt^.fìol¡[ ,luutÊl Çrtod rr]¡ûT]t lr¡ttþttf)IT

- dt$lt ,tol{T,

r - (¡rdltJ'ráo

a - ovtßtåF

'-¡ù-á

î3d-:oo:-ìp-clt{-o r_l'\- o F.Â-"

I

,-ô tA â'-t

API-X

o,-otË

O--t

rt4utL+Ø

ïl-()Z LLJLrJ Jü<I.- CJØv)lr l¡nfO()Jtrtr

X

X'

o\0

IMPROVEDPROI.ILE

BASICPROFILT

1/4

SECONDARYINFLUENCE OF

CIJORD THICKNESSNEGLECTED

3/B 5/B

BRANCH THICKNESS

{-l .zs

2

INCHES

6

FIÇ URE

B 10

(Ð

DESIGN OF HSS COLI,JMNS AND BEAM.COLI.JMNS

D. R. Shermanr

ABSTRACT

The developmenr by AISC of a specification specifically for the design of stn¡cnual rubing

provides an oppornrnity to reexamine design criteria for HSS columns and beam-columns.

Comparisons with other national standards are also possible. The emphasis is on the use ofmultiple coluru¡ curves based on method of manufacture and design criæria for thin-walled

sections. In addition, the results of a pilot test program on the cyclic behavior of ærially loaded

HSS braces illustrate the requiremeût to limit the width/thickness raúo in seismic applications

to prevent local buckling and subsequent fracture.

INTRODUCTION

Structural steel design specifications for buildings have historically been developed for hot-rolled

open shapes and built-up plates members. Even though circular n¡bes were used in some of the

earliest steel stn¡ctures, the trend for widespread use of tubular members and the development

of specific design requirements for n¡bes began in the 1940s. In the case of round n¡bes, the

motivation came from the offshore industries where the circular shape was effrrcient inminimizing the forces on exposed frameworks in a flowing fluid environment. Manufacnrring

technology also produce efficient methods of mass producing square and rectangular ubes as

well as circular without the expensive mills required for hot forured shapes. As a result, a

considerable body of research on the behavior of tubular members has been generated and design

criteria for n¡bes have gradually appeared in specifications. However, in some cases for the

sake of simplicity, conservative criteria for other shapes were applied and the full advantages

of n¡bular behavior were not always achieved. The AISC has initiated an effort to produce a

specification specifically for the use of strucn¡ral tubing in building applications. The

consolidation of this material will simplify the design of n¡bular members and permit the most

eff,rcient use.

Stn¡ctural nrbing can be manufactured by several different processes which can i¡lfluence

properties that affect structural behavior. Consequently, design criteria for different rypes ofi.rUing can vary. At the same time, the designer must be aware of availability so that the criteria

used ln design is for a type of tube that the fabricator can obtain. Another complication in

tubular member design is that many of the standard sizes are classified as thin-walled, so that

a comprehensive design criteria must consider local buckling and not assume tha¡ sections willbe compact.

This paper presents the current state-of-the-art of design criteria for both round and rectangular

1 Universit,y of Wisconsin-Milwaukee, Milwaukee, Wf 5320L, USA

110

colunns' beam-columns and beams (square tubes are included in the rectangurar category.) 1Ianer is included since beams are-o:e anchor oranj interacrion;ì;;i* for beam-corum¡rs.brief background is presented on tube manufacrur¡nä ,n¿ how ir influences sructurar behavi<The emphasis is on the criteria u*.Jin-rli'ü"iä'öilt"s, arthough some comparison with orhnational or regional sandards are made. More à.ør, on ,n.ou?""rJrìng ano the data base fdesign criteria are contained in Ref. 1 il;-;;Aäron or ru¡ui"r-rr,rmber design criteriaspecifications from different parts of n. **rãlñ;;r, in Ref. 2.

MANUFACTTIRE AND AVAILABILITYBoth round and rectangular sm¡ctural ubes can be manufacn¡red seamress or with one or morcontinuous seam welds along the length. eot uon rypes, the tubes

"rn u, further crassified ahot-formed or cold-formed' cold-formø impllsì¡rii ar reast trre nnais¿ing and shaping takerplace at ambient temperatures- ln.trot-foÃä;br;, the sizing and shapíng are performed aelevated temperatures to reduce the sriffner" of trr; materiar. cord-formed tubes that are

åffiäily stress relieved at temperature of approximately 450'c *"

"lr" classified as hot-

In addition to affecting the yield and, ultimate strengths, the method of manufacture infruencesthe level of residual sfesses-in the rube, t¡e uariat-i-oiî yierd srrengh and thickness around rheiäiri:tåtriiåff"htness' ro some exænr, a, or,i,.r. prop"ïi"rl'inu"n . rhe structurar

Generally hot-formed tubes have negligible residuar sûess whire through-thickness residuars incold-formed tubes are very high, especially for *.uJ à.tangurar rube-s. Tubura¡ products areJiilffiò*::#i.i'asured out-or-straien,"Á i"'äå-r"'*?¿L.ã"órar rubes in the ranse

Even though hot-formed and cold-formed rubes may have the same chemistry, the cord-formedmembers will have higher yields and ultimate. rr *á ,, roun¿ea stress-strain curves due to cordworking' cold working alio cause some variation irryirrd srength ".;;;rh. perimeter orcor¿lformed tubes' especialfu at the **.r, of recranguií rì.rio*. Heating ro 450"c w'l rerieveil#å,i:ff 1i:'H;iïË,1*;';x*i'ffihlä;ffiî,',åu.ing,r,,,*."e,¡

Seamless rubes will have some variation in thickness around the perimeter. welded tubes aremade from plate or strip resulting in very uniform thickness.;..p, ;;; a thickening ar thecorners of cold-forrned rectangulai n¡bes.'l-.:r ;;rr'"î¿ strip can u. åuäin.d with precisethicknesses, it is common practice *g1g u s. prod;."r, ,o make Þbes near the minimumthickness permined in the Ëryil*. rp".in..rio", il;-ü; r0äo berow the nominar thickness.trå'ïiiiiïî:iåï*lmt:$"oiäo*'ä"'('.ïJ, momenr or inertia, etc ) shourd be

The information on the.four types of round tubes (welded, seamless, hot- and cold_formed) andfour types of rectangula, ruúËs i, irpon nt in a gtobai sense. However, from a regional

111

viewpoint, availability becomes a par¿ìmount consideration. For example, in the U.S.

recøngular n¡bes are only produced as cold-formed welded. A design specification that included

criteria for hot-formed rectangular tubes would be misleading. The designer who based a design

on provisions for hot-formed tubes would be embarr¿ssed to find that these sections are not

available. On the other hand, recent practice in Canada was that producers could provide a

degree of stress relief so that rectangular tubular colurnns could be designed in accordance withmore favorable provisions for hot-formed members. This raises the problem of insuring tbat the

fabricator obtairs the correct product. Hot-formed seamless rectangUlar tubes are produced inEurope and, therefore, a justification exists for a wider range of design provisions in that region

of the world.

Again considering U.S. practice, round stn¡ctural tubes are produced under the 4500 cold-

formed specification (Ref. 3). However, there are large quantities of hot-formed pþ available

in distribution centers. Currently U.S. design specifications do not distinguish be¡*,een the pipe

and 4500 tr¡bing, so that there is a poæntial problem of acçisition of the t¡.pe of material

inænded in the design or insuring that substin¡æ materials do not require a redesign.

ROT,'IYD TT.JBES

Àxial CompressionThere are three considerations in design criæria for round tubes in axial compression.

1. column buckling curves or equatioris

2. local buckling equations3. inæraction between local and column buckling

Specifications that contain multiple column curyes assign hot-formed n¡bes to the highest and

cõld-formed rubes to the next highest curve. The basis for this is the extensive series of column

tests conducted in the CIDECT (Comite International pour le Develppement et I'Etude de la

Constn¡ction Tubulaire) program in the 1970s. In the U.S., a decision was made by AISC to

use just one column curve. This was to simplify the design process by having only one set ofcolumn load øbles and to avoid potential problems of a design being based on the higher curve

while the material obtained for fabrication is cold-formed. These considerations apply not only

to n¡bes but also to other shapes that could be assigned to various column curyes. The tests data

indicates that the AISC column curve (Ref. 4) is conservative for round n¡bular sections, and

slightly more so for hot-formed members.

Elastic local buckling of circular cylinders is known to be highly imperfection sensitive and the

strength drops rapidly with the diameter/thickness ratio, D/t. U.S. practice of excluding round

rubes that would buckling elastically from building specifications stems from the 1968 AISI

specification (Ref. 5) or the earlier edition. This is accomplished by specifying a maximum D/t

oi O.++Aemr. Although inelastic local buckling is not as imperfection sensitive as elastic

buckling, thére is still considerable scaner in test data. A number of empirically based equations

for predicting the strength have been proposed. The AISC (Ref. 4) equation is based on the

allowable rrés crireria of Ref. 5. This equation is a reasonable lower bound to post 1950 test

data on the local buckling of round tubes under axial compression'

112

PPv

D-lH,

0.03798/F, 2E--3'

D/t s o.rr4E/F..

0.TI E/Fv s D/t s o.44gE/Fv

-l some specification consider an interaction between local and column while others just use thelower of the two critical loads. AJSC,uses rh. ¡;;;;îppr*rr, by modifying the yield srress

il ff ::,'Hi.equation bv a local buckling re¿ucrion-iaoL;, q, which is trrá equivarent of p/p,

for À"1õ> 1.5

= a, for D/t ¿ 0. 0714 n/Fv

for O.O7r4E/Fy

for 0 .3098/F, <

.D/'t s 0.3ogE/Fy (3)

D/T s o.44BE/Fy

Q)

À^= KJIE" rtrl E

Mu

T,

Mu-M"vMu-Mv

. iJ,

0.330D/t

P", = As(o .6}sa^2) eFy, for À"1fis l_.5

P",=WÞ",

one modification- to this approach that appears in some specifìcations is that the e reductioncapacity is applied only to thè critical colu'mn lo.J"nJnot ro rhe corumn srenderness par¿rmercr,À.

Bendingwhen criteria for round tubes first appeared in specifications, the format was atowabre stressdesign' The allowable bending stresses in compression were specified to be the same as foraxial compression' After postlelastic strength was recog nizedin codes, the ly%increase inallowable stress for compact shapes was extended to include rubes that mo n, local bucklingIimit in Equation 1' rni rcn increase was based on rhe minimum ,t"p. rãrr"r for wide flangesections' even though the shape factor for compac, rã*ã sections .*..Ëd. 1.30. when urtimarestrength criteria were developed, circular tuues rrao t" ur reexamined to determine if the samecompacmess limit would apply to develop the full pl*i. .paciry of the round tube. There wasalso a question as to whethir the locai uucni"g rri""in roi a tuue i" oiãi.ompression wourdapply to a member in bending where a sress gradient exists. The resurts of severar experimenarprograms (Ref' 1) were used m develop'ne Tusc

"qu.iìo* for the urtimaæ bending momenr.

(1)

7E/F,t

20t

.0

113

A short range of D/t for elastic buckling is included in order to maintain the same maximumlimit for D/t as for axial compression and still be consistent with the test data. As with anycriteria that is based on empirical results, other equatioru for bending capacþ have beenproposed for other specifications. Some specifrcations do not provide for a transition berweenthe plastic moment and the yield moment and, therefore, contain a significant discontinuity instrength at the D/t which defines a compact shape. No lateral-torsional buckling criteria are

required for round ubes.

Combined Compression and BendingThe rezults of over a hundred beam column tests (Ref. 6) indicaæ that the inæraction equationused in the AISC Specification (Ref. 4) reasonably predicæ the capacity of round n¡bular beam-columns even when local buckling is considered. A linear interaction equation used in otherspecifications is conservative.

RECTANGTJLAR TTJBES

Axial CompressionThe difference in the normalized column strengths benreen hot-formed and cold-formedrectangular tubes in the CIDECT programs is more distinct than for round tubes, causing cold-formed übes to be assigned to lower column curves in specifications with multiple curves. Thehigh levels of residual stresses is a major factor for the lower strength. In the U.S. where asingle column curve is used, much of the data falls below the curve, indicating somewhatunconservative design. However, this situation is not as severe as accepted practice with heavilywelded open shapes, where normalized test data is even lower than that for cold-formedrectangular hrbes. As noæd earlier, only cold-forrred rectangular tubes are produced in theU.S., and with a single column curve, there is no design benefit for speciffing any stress

relieving operation.

The unconservative design of cold-formed recungular columns is not as severe as it appears.

Much of the test data was normalized by the offset yield of the section obtained from subcoluÍrn tests. This reflects the ir¡herent high yield stress in the corners of the tube resultingfrom cold working. U.S. practice is to determine the yield strength with a coupon taken fromthe middle of a side of the finished nrbe. The yield load calculaæd by the material yield strength

times the gross area will be less than the weighted average that includes higher strengths in the

corners. Some European specifications perrrit the yield to be determine from a weightedaverage and other specifications base the design on the virgin yield strength of the plate or stripprior to forming the tube. Thus the appropriate column curve depends on the method ofdetermining the yield strength. With all these refinements, U.S. practice does not result indesign suengths that are significantly different than those of other specifications.

Local buckling of rectangular tubes is almost universally treated with the effective widthconcept. This concept was theoretically proposed by von Karman and later empirically modif,red

by Winter (Ref. 5) to account for inelastic action and imperfections. The concept pertains tothe force carried by a long plate supported on two edges parallel to an axial force. A uniformsrress, which has the same magnitude as the true stress at the edge, acting on the effective width

114

will result in the same post-buckling force using the true stress distribution. The effective widthequation for the case when the side supports have the same thickness as the buckled plate is used

by AISC for local buckling of a rube wall.

b"/t = r.rrrlrft11 - o .3s!{l/r (b/t)l = ot,

In this equation, b is the flat width of the side of the tube and f is the average stress based onthe total gross area, usually the critical stress for the column. A reduction factor Q is the ratioof the remaining effective area divided by the gross area and Equation 2 is used to determinethe column buckling load, which reflects local buckling interaction. Since AISC bases f on thefull section properties of the section rather than the effective properties, iteration to determinethe critical load is avoided.

In other specif,rcations, both the effective width equation and the column curve may differ fromAISC, producing different critical column loads. However, using the concept of effective widthto provide the interaction between local and column buckling is the same.

BendineThin walled rectangular tubes in bending are designed with the effective width concept ofEquation 4 for the compression flange. In this case the stress, f, is taken as the yield stresssince failure occurs when the yield suess is reached in the corners. Using just the effectivewidth for the compression flange causes a shift of the neutral axis away from the flange, as wellas a change in the moment of inertia and the section modulus. The limit moment is determinedby setting the bending stress calculated with the effective section modulus equal to the yieldstress.

Using f as the yield stress and sening Equation 1 equal to the full width, the width/thicknessratio that defines a thin wall section is I .4}JF,/Fy. For sections that have b/t less thanl.L2JElFy. AISC permits the full plastic moment. rrt/hen b/t is between these limits, themoment capaciry is based on a linear transition between the plastic moment and the yieldmoment. Other ultimaæ suength specifications have similar provisions. The limits definingcompact, noncompact and thin walled sections are nearly the same in various specifications,although the definition of width may be the outside dimension, inside dimension or the flatwidth.

Square tubes are not subject to lateral-torsional buckling and, therefore, do not require lateralbracing. Rectangular tubes bending about the major axis could buckle laterally and AISCcurrently has provisions for the unbraced length. However, for tubular sections the unbracedlengths are so large that realistic designs would be controlled by deflection or the reduction ofthe section moment capacity caused by lateral-torsional buckling is negligible. Therefore, thenew consolidated specifrcation will not contain lateral bracing provisions for elastic analysis,although provisions will be included when a plastic analysis is used for the moment distributionand some hinges must sustain finite plastic rotations to develop the failure mechanism. Themaximum unbraced length from the hinge is

(4)

115

LN= ryrrzo.rc r!-rr,(s)

In Equation 5, M, is the plastic moment of the section, M, is the smaller moment at the end ofthe unbraced length, and r, is the radius of gyration about the minor axis.

Combined Compression and BendineAISC uses the same interaction criteria for axial compression and bending as for any othersection. There is some recent evidence (Ref. 7) that this criteria may be slightly unconservarivefor rectangular beam-columns with eccentric end loads when the eccentricity is the same at bothends and produces single curvature. For unequal eccentricities and reversed curvanrres, thecriteria may be overly conservative.

Cyclic Axial LoadingRectangular nrbular braces have been know to fracture catastrophically in earthquakes. A pilotprogram consisting of nine tests of members zubject to æcial end displacement reversals wasconducæd to investigate the failure mode (Ref. 8). The nvo tubes sizes had bit of 36 aú23,with the former being classified as thin walled. Initial column tests showed that since theslenderness ratios of the test members were the same, the two sizes buckled at the same enddisplacement but subsequent local buckles formed at substantially different end displacements.The cyclic test progr¿rm was planned so that there would be no local buckling in one tests whilelocal buckles would forrr in all other tests. Test variables were the axial displacement range,the mean axial displacement and the rate of loading as determined by the period for a cycle.Tests with local buckles follow a similar pattern of behavior. Column buckling is followed bya local buckle which leaves "horns" at the corners. After several cyclés with tensionexcursions, cracks initiate at the HSS corners on both horns and propagate through the thicknessand away from the corners in subsequent cycles. As section is lost at the cracks resulting in aneccentric load, the lateral deflection reverses during the tension pan of the cycle but returns tothe original direction during compression, producing a snap-through behavior. Eventually thecrack pops across the local buckle, resulting in increased lateral deflection that creates a largeenough eccentricity to reverse the direction of column buckling in the subsequent compression.

Although it was possible to make conclusions regarding the influence of the variables, the overriding conclusion concerned the effect of local buckling. The test with no local buckle was

stopped after 50O cycles and all other tests fractured between 18 and 41 cycles. This justifiesthe AISC provision (Ref. 9) that n¡bular braces should have b/t < 0.65V8/F, (about 15) inseismic applicatiorrs. This would preclude the formation of local buckles even under extremeaxial distortion. With further study, it may be possible to relax this restriction to some extentif axial distortion levels can be predicted.

CONCLUSIONS

Sufficient information now exists on the behavior of round and rectangular tubular members to

116

1.

formulate reliable design criteria rhat will take advantage of the properties of the closed shapes'

AISC is preparing a specifrcation that will consolidate provisions for tubular members and,

hopefully, simplify the design process. This specifications will reflect the tyPes of rubes

.uãinulè in the U.S. as well as rhe general philosophy regarding steel design. other

specifications in differenr parrs of rhe world may differ considerably due to the availability of

diff.t nt types of tubes and the acceprance of refined design concepts, such as multiple column

curves.

REFERENCES

Sherman, D.R. 1992. Tubular Members. Constructional Steel Desien-An International

Guide eds. P.J. Dowling, J.H. Harding and R. Bjorhovde: Chap. 2.4,gL-lM. [,ondon:

Elsevier Applied Science.

Kato, B. and Sherman, D.R. eds. 1991. Tubular Structures. Stabiliw of Metal Suuctures-

A World View ed. L.S. Beedle: Chap. 9,495-536. Structural Stabiliry Research Council,

Bethlehem, Pa. : I-ehigh University.American Sociery for Testing and Materials 1993. 4500 Specification for Cold-Formed

Welded and Seamless Carbon Steel Structural Shapes in Rounds and Shapes: Philadelphia

PA.American Institute of Steel Construction 1993. l,oad and Resistance Design Specifìcation

for Structural Steel Buildinss: Chicago IL-A¡erican Iron and Steel Instin¡te 1968. Comrnentary on the Specification for the Design

of Cold-Formed Steel Members: Washington, D.C.Strr.tn*, D.R. 1990. Cyclic and Post-Buckling Behavior of Tubular Beam-Colum¡s.

Tubular Structures eds. E. Niemi and P Måikeläinen: 388-395, Elsevier Applied Science.

Sutty, R.M. and Hancock, G.J. 1994. Behaviour of Cold-Formed SHS Beam-Colum¡s.

Proóeedings 12th Specialtv Cor¡ference on Cold-Formed Steel Suuctures eds' W-W. Yu

and R.A. I¡Boube: University of Missouri-Rolla.8. Sherman, D.R. 1995. Stabiliry Related Deterioration of Structures. 1995 Theme

Conference, 1-9. Structural Stability Research Council, Bethlehem PA: I-ehigh

University.g. American Instirute of Steel Constn¡ction t992. Seismic Provisions for Stn¡cn¡ral Steel

Buildines: Chicago IL.

3.

4.

5.

6.

7.

117

GTIIDE TO TITE HOLLOW STRUCTT'RAL SECTION GT]IDES AND CODES

J.A. Packer'and S. Kitipornchait

ABSTRACT

The principal reference sources or specifications which guide or govem the design of onshore

structures with steel Hollow Structural Sections (HSS) a¡e reviewed. This contemporary(1996) overview of available codes and recommendations is intended as a directory ofauthoritative resource material for the practising structural engineer. The scope of the reviewis international and covers both multinational and national documents, with the latterconcentrating on literature published in the U.S., Canada, Japan, Germany and Australia.

KEYWORDS

Hollow Stn¡ctural Sections, tubes, standards, codes, specifications, design guides

BACKGROTJND

Hollow Stn¡ctural Sections (HSS) were first produced by Stewarts & Lloyds Ltd. in the U.K.and one of the first guides for their use in design was by Abrahams (Ref. 1), 'n L962. Mostresearch on HSS connections in the 1960s took place at Sheffield University under the

direction of Eastwood and Wood (Refs. 2 and 3) and the results of this were quicklyimplemented in Canada and publicized by Stelco in the world's first HSS connections manual

in l97l (Ref. a). Stelco maintained the pre-eminent marketing role for HSS in NonhAmerica throughout the 1970s and 1980s, and the popularity of the product in Canada now islargely a result of this company's efforts. Eastwood and Wood's connection strcngthformulas were also included in the Canadian Institute of Steel Constntction (CISC) Liru¡States Design Steel Manual in 1977 (Ref. 5), but have not appeared in later Manual editions.

A large amount of research and development work on HSS took place during the 1970s,

particularly with regard to connection behavior and static stren$h. Much of this \ryas co-ordinated by the Comité International pour Ie Développement et I'Etude de la ConstructionTubulaire (CIDECT), which is a group of HSS producers with the aim of collectivelydeveloping the market for manufactured tubing. The CIDECT Technical Secretariat has

recently moved to Paris, France, but readers interested in purchasing CIDECT documents(referred to later) in 1996 can most easily do so from Mr. D. Dutta, CIDECT, Marggrafstrasse13, 40878 Ratingen, Germany. Alternatively, most CIDECT member companies carry a

reasonable library of CIDECT technical reports as well as design guides. The only NorthAmerican member is IPSCO Inc., P.O. Box 1670, Regina, Saskatchewan S4P 3C7, Canada-

*Department of Civil Engineering, University of Toronto, 35 St. George St., Toronto, Ontario M5S I44, Canada

#Department of Civil Engineering, University of Queensland, Brisbane, Queensland 4072, Ausualia

118

A new "state-of-the-aft" approach to welded HSS connection design was under preParation in

l9B0 by CIDECT (Monogiaph No. 6) (Ref. 6), but its publication was being continually

deferred so stelco in the meantime proceeded with the publication of its second connections

manual in l98l (Ref. 7). This guide was expressed in a Limit states Design (LSD), or Load

and Resistance Facror óesign (|RFD) formai, and was the first englishlanguage HSS design

guide to do so. The 1980s then saw a period of consolidation of resea¡ch knowledge and

experience commencing with the landmark treadse by v/ardenier in 1982 (Ref' 8)' soon

afterwards followed ',tt," cloEcr book" on HSS design and construction in 1984 (Ref' 9)'

and GIDECT Monograph No. 6 on welded connection static design in 1986 (Ref' 6)'

Tlte International Institute of Wetding (llv), a learned group comprised of -national

welding

societies from around the world with headquafters dso ln Þa¡is, has played a major role in

assessing and assimilating HSS connection åesign knowledge into specification format' This

function is executed Uy Jotunt"er members of IIW's Subcommission XV-E on Welded Joints

in Tubular srrucrures. IIW is cunently approved as an official body for drafting ISo

srandards, so subcommission XV-E will'ükåþ play a key role in influencing international

standards relating to HSS connection design. - To ãate, the two principal connection design

documenrs which this subcommission has iisued relate to static (Refs' 10 and ll) and fatigue

(Ref. 12) design of welded, truss-type connections. IIW documents, which are predominantly

in English, can be obtained from lr¿fr. M. Bramat, secretary General, International Institute of

Weldirg, c/o Institut de Soudure, B.P. 50362,F95942 Roissy CDG Cedex, France'

COI.ïTEMPORARY INTERNATIONAL GUIDES AND SPECIFI CATIONS

ilwThe current, second edition, design recommendations for statically-loaded' welded' planar'

truss-rype, HSS connections (Ref. ll) achieved a wide international consensus and have since

been adopted worldwide by all national or regional specifications and guides, for square and

rectangular sections. For circular sections, tñe same is true except for the U'S' (Ref' 13)'

The fatigue design recommendations, published in 1985 (Ref' l2)' are based on the modern

approach of using the hot-spot stress method rather than the classification method' and are

scheduled for updating in the near future (1996197). Another recent' valuable' IIW

publication dealing witlifatigue definitions, analysis methods and recommendations - although

nor limited solely to HSS co-nnections - is the IIW special report edited by Niemi (Ref' l4)'

CIDECTCIDEC]T has recently adopted the poticy of promoting and disseminating its wealth of

accrued advice by publishing a series of design guìdts on various asPects of HSS

construcrion. These iuides suiersede all previous õpËCf [terature' To date the following

have been published, in the following order:.Design Guide for circula¡ Hollow Secdon (cHS) Joints under Predominantly static Loading,

by Wardenier et al., 1991 (Ref' 15)

,étructural Stability of Hollow Sections, by Rondal et al., 1992 (Ref' ló).Design Guide for Rectangular Hollow Section (RHS) Joints under Predominantly static

Loading, by Packer et ^1.,

1992 (Ref' l7)

119

.Design Guide for Structural Hollow Section Columns exposed to Fire, by Twilt et al., 1994

(Ref. l8).Design Guide for Concrete-Filled Hollow Section Columns, by Bergmann et al., 1995 (Ref.

l9). (This is based on Eurocode 4 for Composite Steel and Concrete Structures).

These five guides have been published in Germany in separate English, Frcnch and German

edirions and can be purchased either directty from the publisher (Verlag TUV Rheinland

GmbH, Köln, Germany) or specific steel construction organizations (e.g. Australian Institute

of Steel Construction (AISC), P.O. Box 6366, North Sydney, N.S.W. 2059, Aust¡alia Fax:

+61-2-9955 5406). Spanish editions should also be forthcoming very soon too. Two ft¡rther

design guides are planned for the near future:.Design Guide for Structural Hollow Sections in Mechanical Applications.Design Guide for Circular and Rectangular Hollow Section Joins under Fatigue Loading.

Another recent initiative has been to produce a comPuter Progfam for perforrring checks on

rhe LSD/LRFD resistance of planar, welded and bolted, truss-tyPe, statically-loade4

connections made from circular, square or rectangular HSS. This program, called CIDIOINT(Ref. 20), follows the rules set out in the two relevant CIDECT design guides above (Refs. 15

and l7). It is available in DOS and Windows vl.l editions, is in LSD/LRFD format, and

has a choice of different secúon databases for different countries. Sales to most countries are

now being handled by a software vendo¡: Computer Services Consultants (UK) Ltd., New

Street, Pudsey, Leeds, West Yorkshire LS28 8YS, U.K. (Fax: t4'1-l 13'236 0546).

Eurocode¡urocode 3 for steel structures (Ref. 2l), to be soon adopted throughout Western Europe, willprove to be a very influential force in international standardisation. Like the CIDECT design

guides for statically-loaded, welded, connections (Refs. 15 and l7), it conforms closely in

Ànne* K to rhe recommendations set out by IIW Subcommission XV-E (Ref. I l). On the

other hand, for fatigue design of HSS welded connections, and for the practical wall thickness

range of up to l2.5mm, the current version of EC3 permits the use of both the classification

an¿ tt "

hoi-spot stress methods. This generates some serious inconsistencies in the EC3 rules

(Ref.22), so this specification should be treated with caution for fatigue design.

Researchetttrougtr not in the coherent form of a guide or specif,tcation, advice and guidance resulting

from new or innovative research in HSS construction can be best found in the Proceedings ofthe International Symposia on Tubular Structures. This series of symposia began in Boston,

U.S.A. (1984) and have since been held in Tokyo, Japan (1986), Lappeenranta, Finland

(1989), Delft, The Netherlands (1991), Nottingham, U.K. (1993), Melboume, Australia (1994)

and Miskolc, Hungary (August 1996), under the organization of IIW Subcommission XV-Eand the sponsorship of CIDECT. The single-volume proceedings from each symposium acts

as an excellent collation of the latest, leading-edge, research on HSS worldwide. The

Proceedings of the 5th. Symposium (Nottingham) were published by E. & F.N. SPon'

London, U.K. (ISBN O 419 18770 7), and the 6th. Symposium (Melboume) by A.A'Balkema, Rotterdam, The Netherlands (ISBN 90 5410 520 8).

120

CONTEMPORARY NATIONAL GUIDES AND SPECIFICATIONS

U.S.A.Surprisingly. considering the size of-the market, there has been little direction given to

designing onshore ,,ru.Iur., with HSS in the U'S., and technical marketing and promotion

have been very modesr. Ar presenr, the American welding Society (AwS) Dl'l code (Ref'

13) covers the static design of welded truss-type connections - in both LRFD and AsD

formats - between "box sections" (square and rectangula¡ HSS) and tubular sections' As

mentioned previously, the connection design rules for rectangular HSS generally conform to

owcIDECTlEC3,but those for circular HîS ¿o not. FatiguJ design is also covered, by both

the hot-spot stÍess and punching shear methods' but a recent comparison between these two

design merhods in AWd Dl.l shows that connections can have very different allowable force

(orstress)rangesdependingonwhichmethodisused(Ref.23).

For the design of members, ties, columns and beam-columns (for both unfilled and concrete-

filled sections), are covered by Íhe American Institute of Steel Construction (NSC)

Specification for Structural Stee-l Buildings (Ref. 2Ð' - .

AISC is now in the process of

producing a separate Specification on iouctural Tubing, which is being drafted by

Subcomminee ll8. Thii will cover both member and connection design and may be

available in 1997. Some HSS promotional material, mainly consisting of safeload tables and

case studies, has been publis'hed by the Pittsburgh-basód American Institute for HoIIow

Struuural Sections (/IIHSS). This Institut" '"p'"'"nted several American tube maufacturers

but has now been closed. Its role has been iargely assumed by the Cleveland-based Steel

Tube Institute of North America lsIl), which has the support of many tube manufacturers

across the u.s. and canada. structural design aids fromsTl have not yet been generated but

a connecrion design guide conforming to the"pending AISC LRFD Specification on St¡uctural

Tubing is planned.

one should aiso be aware of the American specification regulating the geometric properties of

cold-formed HSS used in the U.S. ASTM dr*¿"t¿ 4500 (Ref' 25) permits a hollow section

wall thickness as much as lOVo below the nominal wall thickness, without specifying any

mass (or weight or cross-sectional area) tolerance' This can have a major negative effect on

the assumed (nominat) structural properties (Ref. 26). Most HSS manufacturers now tend to

produce undersized sections, but still within these excessively-generous ASTM tolerances'

Conformiry to nominal member dimensions can be ensured by adding supplementary

specifications to contract documents. A range of HSS grades is produced to ASTM 4500

(Ref. 25), with yield stresses ranging from 2ã8 to 317 MPa for round HSS and from 250 to

345 MPa for square/rcctangular HSS'

CanadaThe prime resource for HSS connection design in C¿nada is the CISC Guide by Packer and

Henderson (Ref. 27), which follows canadiaã specifications and is sold by both GISC (Fax:

+1416491-æ61)andtheu.s.AlsC.Originallypublishedinlgg2'thefirsteditionwasreprinred with some minor improvements in l-ate l-99S, an¿ a revised second edition is due in

late 1996. This book has also recently been translated into chinese and this edition is also

scheduled for publication in Beijing in 1996 (Ref' 28)'

121

The design of unfilled and concrete-filled HSS members is covered by the CSA Standa¡d forsreel structures (Ref.29). HSS in Canada is produced to CAN/CSA-G40.21-M92 (Ref. 30)

with a specified yield strength of 350 MPa. These products conform to CAN/CS A-G4O.âO'

M92 (Ref. 3l) Class C (cold-formed) or Class H (either hot-formed to hnal shape, or cold-

formed ro final shape and stress relieved), of which Class C is now the more popular. One

very imporrant fearure of the CAN/CSA-G4O.2O-M92 specification, especially with regard to

the fa¡ more liberal American ASTM 4500 counterpart, is that it specifies that the mass (or

weight, or effectively cross-sectional area) shall not differ from the published mass by more

¡ha¡t -3.5%. In addition, therc is a -SVo tolerance on wall thickness, but the mass tolerance

will generally govern.

Japanftre Aesign of tubula¡ stn¡ctures in Japan is regulated by the AII (Ref. 32). It is notable that

Japanese standards for cold-formed HSS permit a wall thickness tolerance of -107o, for the

cornmon range of thicknesses between 3mm and l2mm, with no masVweight/arca tolerance

(Refs. 33,34,35 and 36).

GermanvA pt"*i*nt reference source has been the handbook in 1988 by Ðuna and Würker (Ref. 37)'

atthough the recent CIDECT Guides (Refs. 15, 16, 17, l8 and 19) have been very popular in

Germany. There has been a German standard for steel structures made from hollow sections

(Ref. 38) but, like in most other Western European countries, this is destined for replacement

by parts of Eurocode 3 (Ref. 2l). Draft European standa¡ds a¡e already in place for the

mar¡r¡facturing requirements of hot-formed and cold-formed hollow sections (Refs. 39 and 40)'

and these allow for local thickness tolerances of up to -lOVo (depending on size) but are

accompanied by a mass tolerance of -67o. Considering the broad influence that these

EuroNorms will have, this mass tolerance is still far too liberal, especially in view of today's'

modern manufacturing capabilities.

Australiaftr"¡oign of HSS members (for wall thicknesses of 3mm and greater) and rypical

compon"n6 is prescribed by the national limit states steel stn¡ctures specification (Ref. 41).

As än aid to HSS connection design, the Australian Institute of Steel Construction (AISC) is

currently in the process of producing a "pre-engineered" connecúons manual. This will be

publishéd in rwo volumes, ihe f,irst dealing with "Design Models" which is imminent (Ref.

42) and the second dealing with "Design Tables". Cold-formed HSS are produced in

Australia to minimum specified yield strengths of 250, 35O and 450 MPa, with a permitted

local thickness rolerancé of -lOVo but accompanied by a mass tolerance of 4Vo (Ref. 43).

The 450 MPa yield strength is only available at present for square and rectangular HSS with

perimeters up io a00mm. This grade (C45O1C45OL0) is manufactured by Tubemakers ofÀustralia Ltd., by in-line galvanising to a mechanically (shot-blasted) and chemically-cleaned,

bright meral (Rei. 44). Innovative products such as this, combining high strength steels with

,urfu." pre-treatment, plus being aðcompanied by inclusion in relevant national or regional

structural specifìcations, will quickly increase the popularity, market share, and export

potential for Hollow Structural Sections.

122

l.

2.

4.

5.

REFERENCES

Abrahams, F.H. 1962. The use of steel tubes in structural design.Allied Technicians' Association, Richmond, Surrey, U.K.

Draughtsmen's and

Eastwood, V/.; and Wood, A.A. 1970. Welded joints in tubular strucrures involvingrectangular sections. Proc. Conference on Joints in Structures. University of Sheffield,U.K.: Session A Paper 2.Eastwood, W.; and Wood, A.A. 1970. Recent resea¡ch on joints in tubula¡ structures.P¡oc. Canadian Structural Engineerine Conference. Toronto, Onta¡io, Canada.Stelco. l9Tl.Hollow structural sections - design manual for connections. lst. ed.,Stelco Inc., Hamilton, Ontario, Canada.Canadian Institute of Steel Construcúon. 1977. Limit states desien steel manual.CISC, Willowdale, Ontario, Canada.Giddings, T.W.; and Wa¡denier, J. (eds.). 1986. The streneth and behaviour ofstaticallv loaded welded connections in structural hollow sections. CIDECTMonograph No.6, British Steel plc, Corbl', Northants., U.K.Stelco. 198l.Hollo* structural seciions - desien manual for connections. 2nd. ed.,Stelco Inc., Hamilton, Ontario, Canada.Wa¡denier, J. 1982. Hollow section ioints. Deift Universiry Press, Delft, TheNetherlands.CIDECT. 1984. Construction with hollow steel sections. British Steel plc, Corby,Northants., U.K.International Institute of Welding, Subcommission xv-E. l9gl. Designreco¡nmendations for hollow section joints - predominantly statically loaded. lst. ed.,Irw Doc. xv-491-81 (Revised), Irw Annual Assembly, oporro, portugal.International Institute of welding, Subcommission xv-E. 19g9. Designrecommendations for hollow section joints - predominantly sratically loaded. 2nd. ed.,IfW Doc. XV-701-89, IfW Annual Assembly, Helsinki, Finland.International Institute of Welding, Subcommission XV-E. 19g5. Recommendedfatigue design procedure for hollow section joints: part I - hot spot stress method fornodal joints. lst. ed., Irw Doc. xv-582-85, Irw Annual Assembly, strasbourg,France.American V/elding Society. 1996. Structural V/eldine Code - Steel. ANSVAWS Dl.l-96, l5th. edition, AWS, Miami, Florida, U.S.A.Niemi, E. (ed.) 1995. Stress determination for fatisue analvsis of welded componenrs.Abington Publishing, Abington, Cambridge, U. K.

15.

13.

14.

t2.

IL

'wardenier, J.; Kurobane, Y.; Packer, J.A.; Dutta, D.; and yeomans, N. 1991. Desien

7.

9.

10.

t6.

17.

euide fbr circular hollow sectiCIDECT (ed.) aird Verlag TüV

ircularRheinland GmbH, Köln, Germany.

Rondal, J.; wurker, K.-G.; Durra, D.; wardenier, J.; and yeomans, N. 1992. structuralstabilitv of hollow sections. CIDECT (ed.) and verlag TüV Rheinlund c*ffi]Germany.Packer, J.A.; Wa¡denier, J.; Kurobane, Y.; D.; and Yeomans, N. 1992. Desisn

Ia¡ hollow section (RHSCIDECT (ed.) and Verlag TüV Rheinland GmbH, Köln, Germany.Twilt, L.; Hass, R.; Klingsch, W.; Edwards, M.; and Durra, D. lgg4. Desien euide for18.

123

2t.

22.

23.

19.

20.

26.

27.

28.

29.

30.

structural hollow section columns exposed to fire. CIDECT (ed.) and Verlag TUVRheinland GmbH, Köln, Germany.Bergmann, R.; Dutta, D.; Matsui, C.; Meinsma, C.; and Tsuda, T. 1995. Þign guiAe

for õoncrete-filled hollow section columns. CIDECT (ed.) and Verlag tÜV Rtleinland

GmbH, Köln, Germany.Parik, J.; Dutta, D.; and Yeomans, N. 1994. User suide for PC-proeram CIDJOINT forhollow section ioints under predominantlv static loadinq. CIDECT (ed.) and Ing.-Software Dlubal GmbH, Tiefenbach, Germany.

European Committee for Standardization. 1992. Eurocode No.3: Desim of steel

structures - Part l.l: General rules and rules for buildines. ENV 1993-l-I:L99?5,,British Standards Institution, London, U.K.Wingerde, A.M. van; Packer, J.A.; and Vy'ardenier, J. 1995. Criteria for the fatigue

assessment of hollow structural section connections. Journal of Constructional Steel

Research.35: 71-115.Wingerde, A.M. van; Packer, J.A.; and Wardenier, J. 1996. New guidelines forfatigue design of HSS connections. Journal of Structural Ensineerine. AmericanSociety of Civil Engineers, 122(2).American Institute of Steel Constn¡ction. 1993. Lo-ad and resistance factor desimspecification for structural steel buildines. AISC, Chicago, Illinois, U.S.A.American Sociery for Testing and Materials. 1993. Standard soecification for cold-

formed welded and seamless carbon steel structural tubine in rounds and shaDes.

ASTM 4500-93, ASTM, Phitadelphia, Pennsylvania, U.S-A'Packer, J.A. 1993. Overview of current international design guidance on hollowstructural section connections. Proc. 3rd. International Offshore and Polar Eneineerine

Conference. Singapore, IV: l-7.Packer, J.A.; and Henderson, J.E. L992. Desien suide for hollow structural section

connections. lst. ed., Canadian Institute of Steel Construction, 'Willowdale, Ontario,

Canada.Packer, J.A.; Henderson, J.E.; and Cao, J.J. 1996. Desien suide for hollow structural

section connections - Chinese edition. Science Press, Beijing, P.R. China.

Canadian Standa¡ds Association. 1994. Limit states desim of steel structures.

CAN/CSA-Sl6.l-94, CSA, Rexdale, Ontario, Canada-

Canadian Standards Association. 1992. Structural qualiw steels. CA}I/CSA-G4.21-M92, CSA, Rexdale, Ontario, Canada.

Canadian Standards Association. 1992. General requirements for rolled or welded

structural qualitv steel. CAN/CSA-G40.20-M92, CSA, Rexdale, Ontario, Canada.

A¡chitectural Institute of Japan. 1990. Recommendations for the desier and fabricationof tubular structures in steel. 3rd. ed., AU, Tokyo, Japan.

Japanese Industrial Standards. 1988. Carbon steel tubes for general structural purDoses.

JIS G3444-1988, JIS, Tokyo, Japan.

Japanese Industrial Standards. 1988. Carbon steel squa¡e pipes for seneral structuralpurposes. JIS G3466-1988, JIS, Tokyo, Japan.

l"pãn"r. Society of Steel Construction. 1988. Cold-formed ca¡bon steel square and

iectansular hollow sections (box section columns). JSS n-10-1988, Toþo, Japan'

Architectural Institute of Japan. 1991. Japanese architectural standard specification

JASS 6 steelwork. AIJ, Tokyo, Japan.

24.

25.

31.

32.

33.

35.

34.

3ó.

124

37.

38.

39.

TÜV Rfreinland GmbH, Köln, GermanY'

DeutscheslnstitutfürNormung.lgS4.Stahlb?qlenl!¡'4e}'eßeaushohlprofilenunterDIN l8 808, DIN, Berlin, GermanY'

Institution, London' U.K.

European Committee for Standa¡dization' 1992'

(Draft Doc' No' 92146922)'

40.

41.

' 92146923)' British

rnctitrriion- I-ondon. U.K.

Standards

Institution, London, U.

Standards Association of Australia. l99O' Steel structures' AS4I0O-1990' Standards

Áusu¿ia, North Sydney, New South Wales' Australia'

iïìiliil;"il;;ä";i' 'sì.", -ðîirt

u",ion. Lsss. pre-ensineered eonneetþns rora ^. ^.a A tan Nnr.fhlst. ed., AISC, North

I 42.

43.SyO*y, New South Wales, Australia-

Standards Association of eustralia. 1991. Structural Stpçl hollow sections' 4S1163-

1991, Standards Australia, North Sydney' New South Wales' Australia'

Tubemakers Structural and Engineering Products' 1994' Desien capaçitv- -tables for

Ourue"l ,t."1 r,oilo*lrciiont iuU"miL"rs of Australia Ltd" Newcastle' New South

Wales, Australia.

125

-lI

l

.I

I

i

CONCRETE.FILLED HOLLOW STEEL SECTIONS

Eelmut G.L Priont

ABSTRACT

Concrete-filled steel tubes are shown to be an efficient means of carrying comparatively high ardal

loads and moments and are a viable construction method for both buildings and bridges- An

overview is given on the application of concrete-filled steel nrbes as columns with a brief

description oñ"arious coae deiign approaches. The topic of connections to concrete-filled steel

columns is discussed with referãnce to low rise and high rise building applications. Va¡ious

practical methods are described, ranging from simple shear connections that connect to the steel

rtt tt onty, to connections that transfer large bearing loads and moments into the concrete core-

A brief overview is given on the use of concrete-filled tubes as a rebabilitation method for

deficient reinforced c,Jncrete columns and beam to column joints. Strong emphasis is placed on

the suitability of this method for applications in high risk earthquake zones'

INTRODUCTION

Engineers long ago realized the potential for combining the tensile capacity of steel with the

coñpressive streãgrh of concrete in the construction of composite structurd members ofexceþtionalty high load carrying efficiency. Several construction methods have evolved, including

conventioná t"infot ed concrõte and pre-stressed concrete members, composite floor systems,

and composite columns. The latter generally consist of steel members encased in concretg which

not odylfficiently utilize the two materials, but also produce fire-resistant structr¡ral members.

After hollou/ structural steel sections became more readily available, engineers realized the

advantages of filling these with concrete. The two components of the member complement each

other idãally, in thát the steel casing confines the concrete laterally, allowing it to develop its

optimum cómpressive strengt[ whilã the concrete, in turr¡ prevents elastic local buckling in the

steel wall. Another advantate ís that formwork is not required, resulting in a significant saving in

construction cost and time] Athough the concrete core somewhat enhances the fi¡e resistance

above an empty tube, steel reinforðement is typically added to the concrete core to retain the

favourable fire resistance of encased sections.

For beam to column connections, many proven connection methods can be employed. Since the

load carrying capacity of concrete-fiileâ tolumns is significantly higher than for unfilled sections,

however, .nd ,inr. most of the a,xial load is carried by the concrete core, the design of

connections to transfer the high beam shear forces into the columns remains a challenge. These

I Dept, of Civil Engineering, University of British Columbia' Vancouver,B-C.' V6T lZ4, Caruida

126

probrelsl't:o',î,*:'iï'.:ffi"]i'|,üîffiTöTJ;i'"l'1'Hi1981; Dunberry' leminimalguidance,,;;;u"ingprouidjinJ,,igncodes.iii'"l'inii'iA

rngenera,.o::l1Hii"åiïJ:ä"il1"ï*.ii*"i:î"î;å'fffiïjmembers in much t

beam-corumn, ." lt"îïä ,ilr,J"öüio*I':l^Î" not rall within

the limits, specific o-r imphd, or "urr"ni

dËsign specifications' Questions

ffi.Jîiîäxr;ix$'*'git#Ïrn'"::"m'Lffi i::ïl##îï."iitrl"äo"'

::iflt:ïideifo;il" i" tr'" post-ultimate phase'

Design codes over the world are n9t consistent when dearing with concrete-filled hollow steel

members. Many ,"ää;, J..r i,i'î¡r, type or.o,nooíi " section at all' white others are

often very *nr"*ui',JJ "r¿',*ãr: ,,tï, ;;:;"eriar siröhs and/or section slenderness ot

the steel casing. rr;i.- u" pårceived;, ilil;;" "r,rr. itrñied amount of test data available

to enable coae *riie'1o p'ôp"i:j^:'ä *,';ï*;"i*;;"'- l" '¿¿ition

different design

ohilosophies exist in different "ountri"r, în uorrnaule stress;;tg; ri"s, safety iactors are applied

ät oe materiar strength level wherear,nurri¡*,ion factors.i" tñ" i"r¿sand/or.member resistance

æe appried ," *u*r'i*rï;;';-*ot, #öril;;õ;v t.r'ã "arãLoad and Resistance Factor

Design). Also,.*i,å;;;*"lr; t. iniä"'ld ffiFj;;;* "pplv a lower bound criterion

to the data pornts, whefeas ,o*, ur.î;,- value ,o ;iilt* Ëol -¿ resistance factors'

Depending on the scatter of test results, these tw.o ænt-*ti"t-"- d"tiu"r different levels of

reliability uno a.r,gn.r, ur, ,*tion.a îãi ;;- ¿itrerent"iä;;;;åt; without considering the

"täï"ããr"l of ciibration of the codes'

COLUMNS

The most common use of concrete-fired holrow sections is as ærially loaded columns in low to

medium rise buildiner - l" a few.cases, ,tî, :î;;;;;:$ ;;;;;"sed successtullv in high

rise buildingr, ,.ki,,ig ur. of higr¡ _streruth concrete, *jrät'*iii ,ttr aim of increasing the

stiffiress and controrilirï.ä ålie-¿-.ll and Foot, lese)'

previous research on ho'ow structurar sections filled with¡ormal strength concrete (characteristic

strengh of less than 40 Mpa) tu, puurã'itä;;i;t turther developments' especially in the case

127

of circular sections, where considerable strength can potentially be gained from triærial

co¡¡finement of the concrete. The interaction between the concrete core and the steel casing has

been investigated since 1957 (Klöpper and Goder), while Knorvles and Park (1969, lg7Û),Neogi

et al (1969) -d chen (lg7o¡ ri..inrally addressed the relationship between slenderness and

confinement. Since the concrete has to reach about 95Yo of its compressive strength before the

confinement is activated, only stocþ cotumns tpicatly achieve this state before overall buckling

dictates the ultimate strength. such an increase in compressive strength was observed

experimentally as the slendeñess ratio of the column was decreased, but no consens¡¡s has been

reached to define a limiting slenderness ratio.

To achieve full confinement, the steel is best utilized in the circumferential direction and should

preferably not be loaded longinrdinally (Knowlesand Park 1969). In practice, howwer, this is

""ry Aim*k to achievg sincã bond stresses a¡rd frictional forces between the concrete and steel

"",tL longitgdinal straining of the steel, thereby reducing the yield strength in both the

circumferential and tongitud¡nal directions (Furlong 1968, Virdi and Dowling 1980). This was

demonstrated in t."6 ãy Gardener and Jacobson (1967), which have shown no increase in

strength when only the concrete was loaded, compared to full load application to the concrete and

steel.

Consequently, most equations for the ultimate load of composite sections assume that the

component materials ."t ind"p"ndently. Knowles and fark (1969, 1970) and Tomii (1977)

assumed that the steel and the concrete interact by adding ductility and stabilþ to the columr¡ but

collectively do not add strength to the column beyond their individual contributions. The a"tial

strength oithe column is modelled by using a summed tang€nt modulus approach, which assumes

the steel to reach full yield before -buckling;

due to the lateral s¡¡pport of the concrete. The

ultimate strength is thus the sum of the rt.il *d conøete strengths, ignoring both the 'triardal

effects and bond.

For circular sections, confinement of the concrete through hoop stresses in the steel shell resr¡lts

in a significant increase of the concrete strength. The steel itself will, however, experience abi-

axial Jress condition and a reduction of the sieel resistance has to be taken into account. Taking

the above into account, an expression of the following nature is tlpically found in the design

codes (CanadiarL 1995):

Po=crÇ5+pCc

where c¿ represents a reduction in the steel capacity Cs, while the concrete capacity Cc is

increased by a factor p. The factors cr and p depend on the diameter-to-thickness (D/t) and

lengrh-to-diameter (L/D) ratios of the tube and the ratio of the steel yield and concrete

coripressive strengtùs. They both remain unity for rectangular sections and for circular sections

with length-to-diameter ratio UD > 25-

To address some of the concerns and extend the existing knowledge to more slender steel tubes

filled with high strength concrete, an experimental program was initiated which employs steel

tubes with diameter-tõ-thickness ratio @7t) of 92 and yield stress of F, = 262 - 328 MPa' filled

128

with concrete of characteristic strength f "= 73 '92lvPa(Prion and Boehme 1994)' A full raîge

of road combinations (axial load versus åorlri r"", "ppriàJïo "ttrr"o"rize

the road capacities

and load-deformation ùehaviour up ro -äî;;tíu urtimaie.-Further work has subsequently been

done by Rangan ;;;"t;. (tggz), *io-íur"rsstullv t";;;; test results with anal¡ical

predictions

BEAMS

Although concrete-filled holrow sections are serdom used as beams, most columns will experience

some amount of bending in combinationï;i ;h" a¡<iar forJe--ä'it thus important that reliable

expressions ro, tr,"--iã*ent resistanr; b; available, to be used in appropriate interaction

equations-

For rectangula¡ hollow sections with flange wall slenderness up to 720i{Gr), Lu and Kennedy

(1994) have shown irru, *' plaslig ur.lîio"tr are deveroped in the steel and in the concrete'

Excellent "gr**"n, *.,

""f,iä""d br*;;rãri t*tltt -d ti;;roposed T?gtl'based

on zuch

stress blocks, when;;;;** i"r"t in ttr-]i.;ñ;;i.k"" t" i. "q,i¿

io thevield value, Fr' and the

concrete stress level was taken to be equal to the concrete urã"Ëi i" , at túe time of testing' The

rwo compon.n , ,upfoi ã*, other in-tttlärr""r ,"rtt1i1,r ã7 "onnn"r

the concrete, increasing

its compresri.,,. ,"riTluî;""ä;;;'À;tiil; ;;rh* than 0'8i of it' as used in reinforced concrete

theory. At the same time, the concret';;;t*t' inwa¡d Utdti"g tithe steel wall' thus increasing

the steel strain at *r,¡.i'ro.¿ brrkli"J;;;.-';ht"f";;, rãrtiont exceeding the slenderness

requirement, orcnïrï;;;r, in u.nliöîT"ur.i" ã"".lop tulr plastic moment resistance'

Similu results have been achieved for circular ho'ow sections @rion and Boehme' 1994)' who

tested very thin steel tubes fiÏed with r,igt, ur.ogtr, "on.r.irl när" it was found that, although

the bond between a smooth

steel tube and the concrete

seems to be unreliable, the

ã;,i." generated during

bending ihrough Pinching

was su-fñcient to develoP the

combined section strengÍtU

using the steel Yield strength

æd the concrete characte-

ristic strength (Fig' 2)' Since

t¡picallY onlY a small Portion

óf ttre concrete is relied uPon

to provide a balancing com-

pression bloclq the moment

resistance is not very sensl-

tive to the exact shaPe of the

concrete stress block'

P-P

D buckling

Tlever arm

I

_t-

Figure 2: Internal forces in a concrete-filled section

ruPture

(c")

129

BEAM COLUMNS

The interaction of moment and axial forces on a concrete-filled steel column is not much unlike

the behaviour of reinforced concrete beam-columns, with a distinct "nose" in the low arial

load/trigh moment region. This is based on a plane strain model assuming perfect bond between

the concrete and steel. Since

considerable slip can occurbetween the two materials,

various codes have adoPted

different design approaches.

In lapan one can design ac-

cording to an elastic metho{which is uzually strength orstiftress governed, or consider

the ultimate strength which istlpically the case for earth-quake resistant design. In thelatter case it is left uP to the

designer to decide how theloads are divided uP betweenthe concrete and steel, whichimplies that strain compatibilitybetween the concrete and steel

is not required. This is shownin Figure 3 as the suPer-

position model which rePre-

sents a band of accePtable

solutions, the most beneficialinteraction being when the

steel tube does not contributein carrying the a,xial load. Test

results (Prion and Boehme,

1994) have shown that the

compatible strain model more

realistically represents the

behaviour of concrete-filledsteel tubes, although a rela-

tively wide scatter of results

indicates that still more

research is required (FiS. 4).

The Canadian code (Canadian,

1995) is more conservative in

that, for rectangular sections, it

Figure 3: Superposition of intertal forces to resist applied loads

B=Cr+C.-T,Mu= M. + M.

concrele glfeates

Fiqr¡re 4: Moment-a¡ial toad interaction

=o.CLxo

Èoo

xo]UN

=cEo

-+.-.i¡¡_¡.-.

STEEL SECNON

f***\' rp : .¡o'þ l

:,: à !,I,o'2r'12.1 .q-í- -N'A' d:-:

q lç {

CONCREIE SECT¡ON

a Boshme,1989Y Boehme;1988A Tldy,1988

superposttlon model Ca=Q

superposltlon model Cs=Py

NORMALIZED MOMENT lMexp/Muol

130

uses an approach not much unlike thatfor compact l-sections, where a modestincrease in the interaction diagram existswith a limit on the moment componentequal to the pure moment case. Forthose sections where the moment isentirely carried by the steel tube (e.g.circr¡la¡ sections), a linear interaction isrecommended (Fig.a). The benefit of thea¡rial load to expand the moment capacitybeyond the pure moment resistance isthus not permitted.

The European code @urocode 4) hasadopted an approach not much unlikethe reinforced concrete model. Roik andBergmann (1984) have introduced aprocedure where the interaction curve isdetermined by calculation of a few criticalpoints through a plastic plane sectionsanalysis model (Fig. 5), resulting in acurve similar to that of the compatiblestrain model.

Npl. Rd

Npm.R

Npm.Rd

Figure 5: Compatible strain plane secfion interacfion model(Bergmann,1990)

From the above it is evident that the moment-axial load interaction has not been fully accepted inanl final form by va¡ious codes and that continual change in the codes is to be expected in ñ¡turerevisions.

CONNECTIONS

Connections to concrete-filled hollow sections typically range from standard steel connections forsmall loads to elaborate details that are requiréd to transfer loads into the concrete core. Thelevel of complexity depends to alarge extent on the purpose of the concrete in the steel tube. Ifthe concrete has been added for stiffness, fire ptot""iion or to prevent the tube from crushing atthe connectio4 standard connection details ."

"ppropriate since the load in the member isprimarily carried by.the tube.

when the concrete, and possibly additional reinforcement, a¡e called upon to carry asubstantialpart of the axial (or bending) load, it must be assured that a proper load transfer ossurs from theadjoining beams into the concrete core. Since the concrete is ablä to carry a suUstantiA load in anefficiently designed member, the required wall thickness of the hollow structural section is often aslender (class 4) element, with limited capacity to accept targe shear and moment forces.

Multi-storey concrete-filled sections, although carrying a significant load in the bottom storeys, donot experience very large connection forces at each-storey level, as the total load is graáudly

I

I

I

I

I

II

I

131

introduced over atl the storey connections. For such moderate moment a¡rd shea¡ loads, it is often

possible to design a "skin connectiori" with no direct load transfer to the concrete. lhe usr.¡al

steps must be taken to assure that the steel wall will be able to carry the conc€ntrated loads.

Since local buckling is largely prevented by the concrete core, it is tlryically acceptable to use the

full yield capacity of the steel wall. The amount of friction between the steel wall and the

concrete core is uzuqlly adequate to transfer the load into the concr€te over the height of a storcy.

For low-rise buildings where the majority of the arial load is transfened into a column through a

small number of connections, it might not be sufficient to connect to the steel wall only. A direct

load transfer to the concrete core through bearing will be required. This obviousþ requires the

penetration of the steel wall, which adds to the complexity of the connection. Several methods

have been proposed in the literature.

Breit and Roik (1981) have proposed a method where vertical steel tabs pass through the steel

tube, often with cirq¡lar holes cut out of the steel plates to enhance the bearing area on the

concrete core. This method is especially sr¡itable for simple connections where only the web of a

beam is connected to the tab. Becar¡se of the relatively long vertical cr¡t into the steel u¡bg

confinement of the concrete in this critical region will be lost unless the tube is welded to the steel

tabs along the slot.

One way to avoid the loss of confinement is to use circr¡lar ba¡s to penetrate the steel tube, thus

leaving a significant part of the tube intact for confinement of the concrete (Maclellaq 1989).

The steel bars are then welded to whatever connector element is required, uzually a vertical tab

for connection to a beam web. In both cases mentioned, it has been shown that concrete bearing

stresses far in excess ofthe concrete cylinder strength ca¡r be generated, due to the confinement ofthe concrete by the steel tube.

The through-bolt connection methodhas been shown to be very effective,especially when moment forces have tobe transmitted. In this method, ordi-nary steel connestion details are used

in combination with long bolts thatpass through the column section(Fig.6). The concrete prevents crush-

extended endPlateconnection

ing of the steel section and thus endplate connection

permits the bolts to be pre-tensionedwhich increases the stiffness of the

connection, especially when subjected

to moments. Confinement of the con-crete through the steel shell and thepre-tensioning greatly enhance thecompression strength, enabling transferof the vertical shear loads by bearing ofthe bolts on the surrounding concrete. Figure 6: Through-bolt connection for concrete'filled columns

steel beams

concrettfilled HSS

132

I't

I

Tests have shown @rion and Mcl-ellan,1994) that failure tlpically occurs by shearing of the boltsand a more ductile failure mode will have to be assured through detailing of the beam connectionhardware. Slip of the concrete in the steel tube very seldom occurs because of the relatively smallload carried by the steel shell and the additional friction resistance generated by the bolt pre-tensioning.

RETROFTT

The concept of concrete-filled steel tubes presents an efficient means of repairing or retrofittingreinforced concrete structures. V/ith the advancement of knowledge about the respon* oireinforced concrete structures to earthquake motion" the requirements for confiningreinforcement have increased significantly, leaving thousands of buildings and bridges without Ihenecessary reinforcement to withstand a strong earthquake. Encasing such members *ittt circr¡la¡(and sometimes rectangular) steel tubes and filling the gap with cement grout has proven to be acost-effective method of upgrading such deficient structures (Fig.7). The same method has beenused ercensively to repair strustures, mainly bridges, after moderate damage was encounteredduring earthquakes in North¡i dge Q99a)and Kobe (1995).

This method of retrofir has recently beenshown to be an effective method ofretrofining beam-to-column rein-forcedconcrete joints (Hoffschild et al., 1993;Prion and Barak4 1995), which oftenwere constructed without any tiereinforcement at all. Round and squareretrofit were both shown to be adequateto strengthen the joint beyond what wasrequired. Although the round retrofitexhibited more favourable strength andduaility characterisrics, the significantlyhigher cost to fabricate such complexjoint sections is probably not justified bythe somewhat superior performance. Ifnecessary, local reinforcement in re-gionsofhigh sress experienced with the squareretrofit, was shown to signi-ficantlyimprove the per:formance.

fur important issue when retrofining beam-to-column joints, and for that matter, any deficientstructure, is to consider the effect of strengthening part of a structure on the remaining membersof the structure. Since the original steel reinforcement layout was designed for certain momentsand shear forces, the parts of a structure just outside of the retrofit no* right become the weakIink in the system. Since, during an earthquake, forces are generated through motion" the weakestlinks of a structure will experience displacements that will cause forces beyond yielding. If such

Figurc 7: Retrofit of ddicient reinforced concrtte columngbeams and joints with grouted steel tubes

133

newly created weak links are not detailed for a ductile response brittle failures might occur,

rezulting in full or partial collapse of a structure. It is thus prudent to incorporate weak li¡tks ordeliberate plastic hinge locationsu/ithin the retrofit scheme. Aneffective means of achieving thisis to cut gaps into the steel shells,preferably more than one, to¿Nsure a ductile energy dissipatinghinge location. The remainingstrips of steel shell were shown toprovide adequate confinernent tothe concrete to prevent spallingand loss of strength. Experimen-tal rezults show that excellentductile behaviour ca¡r be achievedby repairing weak joint areas and

incorporating plastic hinge zones

in the retrofit (Fig.8).

¡¡O

ÊzåzoÞzt¡¡

=oo=ff 'zo

Àft.o

so-o.l

Figure 8: Eystereticbehaviour of retrofitted reinfo¡eed conctttcsect¡on (Eoffschild et eL, 1993)

SUMIì{ARY

Concrete-filled nrbes have been shown to be an efficient construction method for several

applications, but primarily as columns in buildings and bridges. Although this method has been

used successfully in China and Japan for many decades, its introduction in North America has

been very slow. The major reason for the reluctance of designers to use concrete-filled steel n¡bes

can primarily be ascribed to the lack of expertise and familiarity in the construction industry and

wittr designerq regarding both member behaviour and connection methods. The lack ofknowledge about the topic and its absence in typical Universþ curricula also play ari importarit

role in the lack of its application.

A¡rother reason for the difference in popularity of concrete-filled tubes is the relative cost oflabour and materials in various parts of the world. In North America it might be more cost-

effective to increase the wall thickness of hollow steel sections instead of engaging in another step

and ñlling the tubes with concrete. In some countries steel is a relatively expensive commodity,

whereas concrete and labour are cheap and readily available, which makes concrete-filled tubes a

prefened choice.

In summary, it remains the desig¡er's decision, whether to use concrete-filled hollow steel

sections or whether unfilled sections would be as efficient. Most important is a good

understanding of the behaviour of concrete and steel as these two materials interact to resist

forces in a combined manner. Not only the elastic behaviour is of importance, but frequently the

response of members and structural systems under actions that result in excursions beyond the

proportional limit, requires designers to consider factors such as ductility and cyclic response.

{.qt o 0sJOlt{TROTAnOil c[radl

134

REFERENCES

Bergmanr¡ R. 1990. Composite Columns, IABSE Short Course, Composite Steel-ConcreteConstruction and Eurocode 4, Brussels, 39-68.

Boehme, J. 1988. Behaviour of Circular Steel Tubes Filled with High Strengrth ConcreteSubjected to Bendin-e, Bachelor Thesis, Department of Civil Engineering University of Toronto.

Boehme, J. 1989. Strength of Thin-Walled Circular Steel Tubes Filled with Higùr StrengthConcrete, M.A.Sc. Thesis, Dept. of Civil Engineering, University of Toronto, l70pp.

Breit, M. and Roih K. 1981. Momentenfreier Anschluß an Betongeftllte Hohlprofilstätzen -Experimentelle Untersuchung. Project 52 der StudieEisen und Stahl, Düsseldorf.

Ca¡¡adian Standards Association,. 1994. Limit States Design of Steel Structures, NationalStandard of Canad4 CAII/CSA-SI6.l-94, Rexdale, Ontario.

Chen, W.F. and Atsuta, T. lg76.Theory of Beam-Columns Votume l: In-Plane Behavior andDgstg4 McGraw-Hill, New York, pp.413417.

Dunberry, E., Leblanc, D. and Redwood, R.G. 19S7. Cross-section Strength of Concrete-FilledHSS columns at Simple Beam connections, can. J. civ. Eng., vol 14, pp.4o}4l7

Eurocode 4. 1990. Design of Composite Structures, Technical Paper R65, Annex d SimplifiedCalculation Method for Resistance ofDouble-symmetrical Composite Cross-Sections inCombined Compression and Bending", Bochum, Germany.

Furlong R.W. 1968. esign of Steel-Encased Concrete Beam-Colunrns. Journal of the Strucn¡ralDivision. ASCE, 94(ST I ), Proc. Paper 57 61, 267 -281 .

Gardner, N.J., and Jacobsen, E.R. 1967. Structural Behaviour of Concrete Filled Steel Tubes,ACI Journal, Proc., &(7),404413.

Hoffschild, T.E., Prioq H.G.L., Cherry, S. 1993. Retrofitting Reinforced Concrete Joints withGrouted Steel Tubes, Proc. Tom Paula]¡ S]'mp., Univ. Southern Calif, La Joll4 Sept. 1993 ,403-431.

Johnson, R.P. 1975. Composite Structures of Steel and Concrete Vol. l:Beams. Columns.Frames and Applications in Building, Constrado Nomograph, Crosby Lockwood Staples,Granada Publishing L¡d., London.

135

Knowles, RB. and Parh R. 1969. Strength of Concrete Filled Steel Tubular Columns, Journal ofthe Structural Division, ASCE, 95(STl2), Proc. Paper 6936,2565-2587-

Knowles, RB. and Parh R 1970. Ardal Load Design for Concrete Filled Steel Tubes, Journal ofthe Structural Division, ASCE, 96(5T10), Proc. Paper 7597,2125-2153.

Lu, Y.Q and Kennedy, D.J.L. 1994. The Flen¡ral Behaviour of Concrete-Filled Hollow Structural

Sections, Can. J. Civ. Eng., 2l(l), I I l-130.

Mclellan, AB. 1989. Behaviour of Beam Connections for Hollow Circ¡lar Steel Tube Columns

Filled with }figh Strength Concretg B.ASc. Thesis, Dept. ofCivil Engineering Universþ ofToronto.

Neog, P.K., et al. 1969. Concrete-Filled Tubular Steel Columr¡s Under Eccentric Loading; The

Structural Engineer. 47 (5), I 87- I 95.

Priorl H.G.L., Boehme, J. 1994. Thin-Walled Steel Tubes Filled q,ith HiSb Strength Concrete,

Can. J. of Civ. Eng.. V.21,pp.207-218

Prioq H.G.L., Baraka, M. 1995. Grouted Steel Tubes as Seismic Retrofit for Beam to Column

Joints, Proc. 7th Can. Conf. on Earthquake Eng., Montreal, Que., Jun. 1995, 871-878.

Priort H.G.L., Mclellar¡ A"B. 1994. Through-Bolt Connections for Concrete-Filled HollowStructural Steel Sections, Proc. Strucn¡ral Stability Research Council Annual Tectrnical Meetine.

fune 1994,239-250.

Raridall, V. and Foot, K. 1989. tügh-Strength Concrete for Pacific First Centeç Concrete '

International. pp. 14-16.

Ranga¡U B.V. and loycg M.1992. Strength of Eccentrically Loaded Slender Steel Tubular

Columns with High-strength Concretg ACI Structural lournal. V. 89, No. 6, 676{,81.

Roih K. and Bergmann, R. 1984, Composite Columns - Design Examples for Construction" 2d

US-Japan Sem. Compos. Struct., Seattle, July.

Tomii, M., et at. 1977. Experimental Studies on Concrete filled Steel Tubula¡ Stub-Columns

under ConcentricDvnamic Loads, ASCE, 718-741.

Tidy, M.S. 1998. Hollow Circular SteelTube Columns Filled with High Strength Concrete.

Bachelor Thesis, Department of Civil Engineering University of Toronto.

Virdi, K.S., and Dowling, P.J. 1980. Bond Strength in Concrete Filled Steel Tubes, IABSE

Periodica, Internat. Assoc. for Bridge and Structural Engineering, 125-139.

r36

FT.JNDAMENTAL CRITERIA FOR WELDING TTJBULAR STEEL

R. M. Bent"

ABSTRACT

Although weldabitity has no universally accepted definition, the term is commonly used to

describe the relative ease with which a steel may be joined. Physical factors such as base metal

chemistry, preheat, and filler metal must be selected with care. similarly, design aspects such as

electrode consumption, joint configuration, weld type, and general accessibility must receive

equal consideration. To áchieve the above criteria with HSS, the designer must ensure that the

joining members satisfy the geometric parameters (relative dimensions and wall thickn¿ss) to

äpti*iä¡oint efficiency ano lú¿ capacity. The design will thus feature accessible joints welded

wittr simple frllets, a competitive edge that car¡not be easily surpassed by the _concept

of minimum

weig¡t. îwo points, however, ruti be appreciated: (1) not all structures lend themselves to HSS,

and (2\, an arbitrary substituiion of one itSS member for another seldom succeeds, even if the

substitute has an equivalent load carrying capacity. The stntctuml engineer's choice of member

size and joint orientation will predetermine both the quality and economy of the final weldment'

INTRODUCTION

General Propertiesln car¡ada, the most commonly used HSS conforms to csA G10.21-350W, class H' The 350"

indicates a yield strengrtr or¡io Mpa (50 Ksi), while the flt"indicates good weldability (carbon

equivalent mean of 0l¿0, with a ma:rimum of 0.44). Cf* H tubular steel is made by: (1) a

såmless or continuous welding process and hot formed to final shape, or (2), a seamless or

automatic welding process ptøuðing a continuous weld, and cold formed to final shape, then

subsequentty stress relieved at 850;F., cooling in air. Tables l and 2 give the chemical

.o*poìition and physical properties of "350'W" and several other grades'

The majority of welding on HSS structures is done with the following three processes:

o shielded MetaiArc Welding (sMAlV) -- a conventional manual process with covered

electrodes; the weld metal is protected by gases and flux produced as the rod melts.

Although the rate of weld deposition is somewhat low and varies considerably with each

welder, the overall versatility over a wide range of applications and the relative ease of

set up maintain the popularity of "stick welding''

. Flux Cored Arc lvelding (FCAW) -- another semi-automatic process that uses a hollow

continuous wire f,rlled with flux and other chemicals; the weld is protected either by an

extemally applied gas (commonly COr) or a self-shielding gas generated as the electrode

" Senior Welding Engineer, Welding Institute of Canada, Oakville, Ontario

137

melß. The deposition rate of FCAW is about riple that of SMA\ry, generating a high heat input.On thicker walled HSS, this process is extremely effective.

llSS-Chcrnic¡l r.gu¡r.rn ln¡

Ctrartlcal rrqul'lmcnts - hcaf aneryst€ tpêrën0

f l l îræ ffi E tffi ry Ð G t t¡''. t. æ É D ÞE Fñaa nEÈ G E r ffi rt rErt. r ñ ætø rrlE t'ct D l' m h Eæ r dr o. l¡ m. tr 9ñrtrr æ ¡¡t Ð Eñ Fr. E E ffi ..ñt lrúrn D c DÞæaæfi¡Ctædo¡Offi @ rÐ É,qrDyñffi ot gËl¡l firæEcffid

'EffiüÐ E EÞO{O,åÉlat!brEG.SËr.

Gas Metat Arc Wslrling (GMAIV) - a semi-automatic process similar to FCAW, witha continuous solid wireelectode. The weld is protectedby externally applied inert gases

such as lñVo Argon or Helium,or, mixh¡res such ¡rs Argongl%oI Oxygen 5Vo. The depositionrates are almost ¿ts high ÍrsFCAW; however, p,roductivity issensitive to changes in operatingparameters such as wire feedspeed, amperage, etc. Thequality is excellent, but theprocess is demanding on thewelder. In the soon to bereleased CSA Standard tV59-1996, Welded SrcelCorutntcdon, GWAÌW willbecome a prequalified process.

Prequalified welding procedures and joints translate into substantial savings because noqualification welds, subsequent æsting, nor PQR's are required (Figure la) - however, a writtenWPS is mandatory. Not only are the savings to the fabricator substantial, but there is now oneless thing to worry about. The joints are detailed in Sec.l0 of CSA W59 (Figure lb).

138

sl¡nr¡nl I o*t"

rrr"t. I Mî P ílar S nrarlGre¡nrclümgSr I .lùn.lüS.ù fu¡¡t ¡¡Ð Ct¡

sÉ{¡¡.U.e¡-MBt I 3X¡WI ssowI sætt'I 35{¡WTI gsounI 350r"'I ¡sor¡"¡ASlì¡ A50O : Gr.AI cr.gI o.c

,fSOî : -

0.26o.ao.ao.2.o.20.æoã

030,l¿o0.50,r.50oson.i500.80rt.500.æ,t500.73r.350.75n.35

o.o.0.040.oa0g¡o.Gto.G!0.txl

0.050.0!t0.050.04O.O¡¡o.oaO.O¡l

O.¡lO'lt.r.0.¡lOmar.O.¿lO'rl.r.0.t5,O.¡¡O0.15r0..l(l0.r5¡0.¡100.15r0.¡l{,

o.toì,o0.f0nìar0.10 rnar.0.t0mü.0.lO mâr.

0ã)o.6{)020/0.60

t:t_t_t-b-go m¡¡b-mnrt

,7Oma¡.TOnnr

oâ0:6oâo2a

r35 ma¡.

o.oa0.0a0.0.30.oa

0.050.0t¡0¡5o.o5

0¿0flit.'¡0.20dñ.60¿0ntì.60¿Onh.'â

Table 1: HSS Chemical Requirements

æ221z¿z12121

8t

æ2;'a

zÐ'l26gDæt¡31743l¡"3.ga28

¡lt;bæE!raSoDæ€rç¡¡rE¡¡IYluúæralslæ ffi E æffi É d m@ ll

'!ørrdlt ñæF E

lstsææ E!ÐEE.æü F mc.@rÇ@¡ãìg.t6lSÞ.tÉ¡lErrS¡Ery

Table 2: HSS Mechanical Properties

Procedure Test

Two methodsof support

Written weldingprocedure:. WPS. WPDS

Prequalified jointsplus proceduralrequirements

Figure la: Prequalified Procedurcs

Thus, with three proven welding processes usingprequalified procedures on the highly weldable"350W" HSS base material, the fabricators begin thejob under conditions that not only offer flexibilify butalso an opportunity to reduce capital costs. Now, ifthe stnuctu¡e has utilized the special design guidelinesfor HSS member selection, the prospects will alsobode well for:

o high joint efficiencyo high qualifyo high production

JOINTS AND WELDS

General ClhservationsFillet ar¡d/or groove welds(usually without a backing bar)are commonly used in HSSfabrication. Either weld qpecan easily develop the fuXcapacity of the HSS wall. Forexample, the two fillet weldsin Figure 2 just match themaximum load of the member

- this balanced design sets anupper boundary on the weldsize. A misconception held bymany designers is that "a 100%weld' must be a CJPG weld,when in realiry a parr of simplefillet welds will likely suffice.

Figure 2 : Balanced Design

. 6 ñm mrn. lot -V '

Fïgure lb: Prequalified Joint

139

Fillet lVeldsEase of welding and minimum joint preparation and fitup requirements make the fillet weld thefirst choice, Fillet welds are used almost exclusively in web-to-chord truss connections. Theyare frequently used in T-joint configurations of Vierendeel truss€s (Figure 3). Researchen (l)have established that unreinforced equal widttr HSS connections can in some instances achieve fullmoment transfer. For an unstiffened connection both strength and flexural rigidity decrease as

bo I to increases nd \ I Qdecreases.

Connections with bt=boand a low bo / to agproachfull rigidity, but all otherunstiffened connectionsshall be classed as semi-rigrd Ø. Joints withunequal chord widths maybe reinforced to improveperformance: severalmethods have been

evaluated (3), with the flatplaæ fillet welded to thechord being especiallyfavoured.

It is generally moreeconomical to substitute acombination groove andreinforcing fillet if therequired fi.llet size l2.7mm('h"), as shown inFigure 4.

Groove lVeldsGroove welds a¡e classified as either complete penetration or partial penetration. CSA W59 has

strict criteria of what constitutes a CJPG weld (Figure 5) and a PJPG weld (Figure 6). A grooveweld welded from one side only must be done by a welder with a valid 'T" ticket. The procedureis not prequalified. Deails for prequalified groove joints in circular tubular steel may be foundin A}ryS Dl.l Structural Wglrling Code, Section 10 (prior to 1966 rærganization of the code).ln general, the material preparation and fitup is often time-consuming, making groove welds veryexpensive.

Figure 3: Vierendeel Tn¡ss lÞtails

140

1/2 in.(13 mm)

Groove welds are sometimes used in place of filret werds in thefollowing circumstances :

To achieve the required weld throat when ,n. ¡orn, geometryprecludes using a fillet (Fîgure Ð.

To reduce weld weight. For example, the weight ofdeposited weld mehl on a T-joint having a wall thicknessof l2.7mm would require a 1" f,rllet at l.9Z lb/ft.However, a l2.5mm groove weld with a 12.5mmreinforcing fillet would use only half the weld metal(similar to Figure 4).

To make bun joint splices between two HSS members,preferably with a backing bar (Figur€ E). Splicesutilizing flange plates should usê a groove/fillet,especially for highly stressed tension members (chord oftruss). See Fïgure 9: the tube has a groove reinforcedwith a fillet, providing extra strength and a bener overalljoint contour.

wekled lrom on€ s¡d€ wlth steel bactcing

2.

bæk gcugÚìg to go¡JndrnoE|l trorî oúìor sEa

cÉ¡îÞþtim of w?ld fÎfitsosrd s¡da

WELD AREA = 0.25 in2(l6O mmzl

WELD AREA = 0.13 in2(8O mm:¡

Figure 4: Reduced Arca

1/2 in.ll3 mm¡

wetct on ftsl (preDarod) ide

Fþre 5: CJPG Welds

141

t,

prepafed 10lacilitate tusion¡nto vert¡cal wallard develop largerthroal

ptaneofnoat / íiernber bu¡ld up

0 = 9f (PJPG)

Figurc 7: Contour Radfu¡sed Cotrer

Flgr¡re 9: Reinforrced Groove lVeld

ngr¡re 6: PJPTG lVelds

a) penefaüon tessüan compteÞ

b) welcted from one sklewithoutsteet bactdng

c) welcled from boü sideswiüout bad<gouging

Racking RarsBacking bars are generally not required.They are difficult to fit and do not add

strength. Two exceptions would be:

1. Butt joint qplices, Íts alreadynoted.

2. When both the web and chordhave the same width, especially ifthe gap is large at the radiusedcorner of the chord. @gure 10)

\ Omin\

Figure t: Butt Splice With Backing

142

Preferences

From the preceding discussion, the welds in orderof preference are:

. Fillet welds

o Partial penetration groove welds (PJPG)

o Complete penetration groove welds(CJPG), with backing

o Special PJPG weld made from one sidewithout backing, in accordar¡ce withAppendix L, CSA \ry59, which defines itas CJPG weld under static loading

@gure 11).

3e s t< gfrOpcn Side

whent= *"tn= |l?n" whant)r/.":

6æ < a< 90"Acute S¡do

9 ' t/tclo 7/ç

Iïgure 10: IVidth Mismatch

Fìgure 1l: HSS CJPG Weld, Appendix L of CSA W59, Static I¡¿¡ling OnIy

'I

i

143

OPTIMTJM JOINT CONTTGI,JRATION

Gap.IointsThe Gap "Ioitt¡ shown in Figure 12, connecting the truss chord and web members, illustrates an

optimum fabrication. The gap joint here requires only a single cut, a single pass fillet weldaround the web, no gr,ove preparation or backing bar, no gusset plate, with easy fitup and ampleaccess. Note that the webs are thin- walled, marginally less wide than the chord, two essentialfactors. Compare it to the conventional joint configuration in Figure 12a.

\ileight of Iìeposited lVeld MetalBesides of easy fabrication conditions, theoptimum joint minimizes the amount ofdeposited weld metal. There a¡e at least threefactors that can influence this objective.l. Angle between web and member.2. Thickness of HSS wall being welded.3. Method of design used to size the

welds

.Ioint AngleFillet welds vary from 60o to 120'; PJPGwelds are used elsewhere. The ratio betweenweld size and throat size varies with the weldangle, as shown in Fïgure 13 (SectÍon 3 ofthe CISC Ilandbook of Steel Construction).For the same resistance, larger welds are

required for obtuse angles than for acuteangles. On the same page, CISC Figure 3-f 1

has a Table that shours the minimum 90" weldleg size for the given ranges of wall thickness

getween 60-90", which is useful for comparing the throat sizes of skewed fillets. Note: Heelwelds at joint angles less than 30' do not contribute to the load sharing.

IISS lVall ThicknessIn CISC Table 342 the minimum weld size is æt according to the wall thickness (Figure 13a).This can result in a weld leg that is significantly oversized, having a capacity considerably greaterthan the web member being joined. However, the Code also specifies that the weld næd notexceed the thickness of the thinner part being joined. This criteria is obviously an advantage forwelding thin walled HSS.

ì

tE-iìllE.l r+I\i

t

Itigure 12: Optim¡¡m Ç¡p Joint

144

hernative to OPtimrrm HSS Joint

COMPETING DFSIGN CONCEPTS

prequalified Weld Si'e ,-.^--.^t:c^)¡nn¡mr fnr cizinsFor 350 Mpa tubes with gap joints, the Irw l,*,s apreqwwconcept for sizing a finet werd that

matches the capacity of t¡'. *.u; set the ,n*ã qøto r'i'ito the web thickness' The Canadian

Codes, using the minimum leg sizr^ ¿.æ,'inø uy various thickness mnges, would result in ar¡

equivalent throat value of L46t'

Calculated Weld Size - -,^-

The arternative method is to carculate the weld size needed to carry the actuar road- In theory,

simply divide the member load by n" nlu bnstharound the tube todetermine the required weld

resistance per unit length. The sloped sides o1the web member should be accounted for' as per

Figure 14.

ffinthattheactual,oreffective,w9ldlengthvariesaccordingtothewebangle. when t¡e cnord angle is 60o or *ãi* the effective rength óonsists of the rwo longitudinal

sides and the width along the toe, butttre trál weld shourd be considered completely ineffective'

The effective weld length can now be calculated, and the necessary weld size calculated on the

basis of aPPlied loading'

Transversely I oaded fillets

In the current øition of csA s16.1M-g4, the resistance of filtet welds varies according to the

orientation of the apptied stress. The resistance in tension transverse to the weld axis increases

with the angle, attaining a maximum .i 90' (Figure l5)' This is a 50Vo increase' There is no

145

I

change in weld strength when loaded parallel to the weld axis. The new formula allows thedesigner to poæntially reduce weld sizes for advantage weld/load orientations. The effects of thenew equation occur between 50' and 90". For T-joints, CSA values would possibly approach orsurpass the II1V values. The equation is given by:

V,=0.67Q ¡\nX,( I .00 +Q. 5Osint'50)(l)

WELDING DÉTAILS FOR HOLLOW STRUCTURAL SECTIONSEltcüavr Thro¡t:-T-5mmlo¡ 0-3Ooro44o1- 3 mm lor g -45o to 59o

e -eelseJiL* - r's'

Dctail Ad - 30o to 59o

Effectivc Thro¡t:-T>0.7075as per TaUe 4.2in W59'M89

Detail Bá = 600 to 9Oo

Effecrive Throa¡: - T = O.7O7S

1X,¡- S

ilsChordMcmberBuilt Up

Dct¡il C

0 -9Oo

Effec¡ive Throar - T

Add¡tio.ì!lPrap¡r¡tion toDcrclog lrrgnrTh¡o¡t

I\ cnoø

lVlcmþcrEuilt Up

Detail Dg-90"

Eflec¡ive Thro€t: - T = 0.707 ¡ F ¡ Sr-¡\ ¡

rso- I

Dctril E0 - 91o ro 1200

Effec¡ive Throat: - T = t for0 = l35oTctloc0= 1360to 1sOPT>t for 0= 121o to l34o

TYxDctail F

H = 121o to t50o

e 91-100 10r-106 107.1 r3 r r4-t20F o.95 0.90 0.85 0.80

Figure 13: Prequalified Joints, CSA W59

r-i

146

HSS CONNECTIONS90o Fillet Size þ Develop Wall Strength

Table g-42E480XX Fillet Welds Fy = 350 MPa

Frgure 13a: CISC Table of Minimnm Fillet Wetd Sizes

LENGTHoFw€LD=20- #

Figure f4: Iængth of lVeld Includes Effect ;f St"p"

FLARE BEVEL FLARE GROOVE WELDS

Nasty ProhlemFlare bevel groove weld¡ formed by setting an HSS member against a flat are not prequalifiedin canada' The poor tolerances on tne ,"diu, of square and rectangular sections precludes a

WallThickncss

{mm)

Filter Leg Si¡elmml

Wall in Sheer Wall in Ten!¡on

3.8r4.786.357.959.53

t 1.1312.70

6I

r01214r6r8

Il014r8202426

147

Tron¡vc¡sc Lood

Fïgurc 15: Transversely l,oaded Fillets are Stronger

direct measurement of thepenetration. Thus, there is anadded cost to qualify theprocedure, ensuring that therequired throat can be attained byusing appropriate weldingprocedures. At present, CSAW59 is working on statistical datato develop a mathematicalrelationship benveen visibledimensions in terms of the HSSradius. Not being prequalified,one must pay for procedurequalifications.

CIJOSING REÙíARKS

This short paper can only touch on a few topics with respect to the welding of HSS. One shouldremember the fundamental distinction between the resistance of welded joints and the resistanceof conneaio¡ts. The connection has a resisance (as a function of the geometric parameters) whichis often less the capacity of tt¡e member. That resisance cannot be increased by adding additionalwelding because the extra weld will not be effective in transferring load through the connection.Such extra weld is wasteful, and could cause harm through the unnecessary introduction of extraheating, shrinking, and restraint.

Thus it is somewhat ironic that the design guidelines for choosing connection and joint efficienciesalso result in conditions that a¡e ideal for an optimum fabrication, both in terms of quality andcompetitiveness.

REFERENCFS

l. Cran J. A.; Gibson E.B.; Stadnycþi S. 1981, 2nd eÅ. Hollow Structural Sections,Design Manual for Connections; Stelco Inc.

2. Packer, J.A.; V/ardenier, J; Kurobane, Y; Dutta, D.; Yeomans, N. 1992. Íresign Guidetror Rectangrrlar Flollow .section (RF{.S) Joints lInder Orertominantly Static I oadin&CIDECT, Germany. ISBN 3-8249-0089-0

3. Frater,G.S.; Packer, J.A. 1990. l-tesign of Fillet Wetdments for ÉIollow Structural SectionTrusses- CIDECT REPORT No. 5AN/2-90173; ISBN 0-7727-7570-2. University ofToronto.

148

_lI'I

4-Packer,J.A.;Henderson,J'E'1992'rresignGuideforFlollowstructuralsectionConnecrions. CISC. ISBN G.8881147G6. Universal Offset Limited, Markham, Ontario.

5. Koral, R.M.; Mitd, H.; Mirza, F.A. 1982. Plate Reinforced Square Hollow Section T-

Joins of Unequal Width, Canadian Journal of Civil Engineering, Vol.9, No.2, pp. 143-

148.

6. International Institute of Welding Subcommission XV-E, Design Recommendations forHollow Stn¡ctral Joins - Predominantly Statically loaded,2nd ed., IfW DOC. XV-701-

89.7. Cran, J.A.; 1pg! Waren and Pratt Truss Connections, Weld Gap aFd OverlaP Joints

Using Rectangular Chord Memhers. Technical Bulletin 22, Stelco Inc.

8. CEN/TC l2lts1 4/WG 6 No 24; Welded Connections - Part 1: Steel 'Structures' (Finat

Draft) Part D, pp.22-29.9. CISC Handbook of Steel Construction, Fifth Edition, 1993.

10. AWS Dl.l-1994 Structural Welding Code - Steel, Section 10

11. CSA Standard 516.l-94 Limit States Design of Steel Structures

12. CSA Standard W59 rù/elded Steel Construction

149

BENDING, BOLTING AND NAILING OF EOLLO\ry STRUCTURAL SECTIONS

J. E. Henderson

ABSTRACT

Hollow structural sections (IISS) are bent by rolling or by mechanical means to createcurved sections for aesthetically pleasing structures. Smaller radii are increasingly attainablewith improved bending techniques. Bolting other structural members to HSS sections haslong been constrained due to the inaccessibility of the interior. Various blind boltingsolutions exist, but only recently have products with promising stnrctural and economicalperformance emerged. For some applications, an alternative to welding or bolting HSS maybe power-driven nailing a method that has recently been demonstrated to be practical.

BENDING HSS

Introduction

Curved HSS are used by designers to create a wide variety of original and aestheticallypleasing structures. While architects and engineers have been taking greater advantage ofthis potential, industry has been developing increased capability for curving HSS.

Hollow structural sections can be bent either cold or hot. Rolling or mechanical bending isused for cold curving while induction heating is generally preferred for hot curving.

Cold rolling square and rectangular HSS with conventional three-roll machines was studiedfor CIDECT (Comité Internatiotnl pour le Développement et I'Etude de la ConstntctionTubulaire) in 1988, and reported in the Packer and Henderson guide (Ref. l). Curvaturewas limited by wall distortion of the sections, which quickly became excessive. Howeveçwith custom rollers that better support the section, much smaller radii are presently beingrolled. WhiteFab Inc. of Birmingham, Alabama reports that they have newly patentedequipment that holds and bends the HSS by means of hydraulic grips and cylinders, aprocess they find more precise and more economical than rolling.

When cold forming a given size HSS, tighter curves are possible with increasing wallthicknesses. Some slight concave distortion of the wall that is next to the inside of the arcusually occurs, but the other three walls generally remain true. Mechanical properties arealtered by the cold work associated with rolling, so that ductility after rolling is less than

Principal. Henderson Engineering Services. Milton. Ontario, Canada. [email protected]

150

before and ultimate strengfh is higher. The stress-strain relationship below yield level is notsignificantly affected.

Induction heating is used to produce precise complex bends in large and heavy shapes aswell as in conventional structural sections. Examiles are2 to 12 inch diameter pipe withwalls up to 1.5 inches thick bent to radii from 5 io 60 inches, and lz to 66 inch diameter{ne- ¡tttr walls up to 4 inches thick bent ro radii from 40 to 3g4 inches, as quoted byNAPTech Inc. of Clearfield, Utah.

BENDING HOLLOW STRUCTURAL SECTIONS

Section Radius(m)

Process Sect¡on Radius(m)

Process

ROUND HSS d xt (mm) RECTANGULAR HSS (bent about y-y axis)60.3 x 5 0.4 Rolling 152x51x6.4 1.8 Mechanical114.3 x 6.3 0.7 Rolling 152x102x6.4 2.1 Mechanical168.3 x 10 0.9 Rolling 203x51 x6.4 3.1 Mechanical219.'l x 12.5 1.1 Rolling 203x1O2x6.4 2.4 Mechanical

254x102x9.5 2.4 MechanicalSOUAREHSS hxb xt(mm) 304x102x9.5 2.4 Mechanical

CUXCUXþ 0.6 Rolling 406x102x9.5 3.7 Mechanical76x76x6.4 1.2 Rolling 406x203x9.5 10.4 Mechanical100x100x6.3 1.1 Rolling102x102x6.4 1.8 Mechanícal RECTANGUI-AR HSS (bent about x-x axis)102x102x9.5 1.5 Rolling 102x51 x6.4 1.8 Mechanical102x102x9.5 2.8 Mechanical 52x51x6.4 1.8 Mechanical127 x 127 xg.5 1.8 Rolling 152x102x9.5 1.8 Mechanical152x152x9.5 2.0 Mechanical 203x51 x6 4 2.4 Mechanical152x152x9.5 2.1 Rolling 203x152x64 2.3 Mechanical150 x 150 x 10 1.4 Rolling 203x152x9.5 2.9 Mechanical150x150x12.5 3.0 Rolling 254x51x6.4 3.1 Mechanical152x152x12.7 2.1 Mechanical 250 x 150 x 12.5 9.0 Rolling203x203x6.4 4.9 Mechanical 250x102x9.5 3.8 Mechanical203x203x9.5 3.1 Mechanical 254 x203 xg.5 7.5 Mechanical203x203x9.5 4.9 Rolling 305x102x9.5 3.5 Mechanical200 x200 x 12.5 2.0 Rolling 305x203x9.5 4.9 Mechanical203x203x12.7 3.7 Rolling 305x203x12.7 9.2 Rolling254x25ax9.5 7.0 Mechanical 406x102x9.5 8.6 Mechanical254 x25a x 9.5 15.3 Rollino 406x203x12.7 19.9 Rolling254 x254 x 12.7 9.2 Rollino305x305xi2.7 12.2 Rollinq356x356x9.5 23.5 Mechanical

Table 1: some representative radii of curvature for cord bent HSS

151

Examples of HSS Curvatures

It is difficult for companies that bend steelto provide a complete range of minimum radii forcurving HSS because of the number of variables involved. They can however provide

examples of curvatures produced in the past and opinions as to what is likely feasible with a

particular section. Table I gives a representative listing of some recent cold forming results

that have been reported to the author.

EUCK INC. HIGH STRENGTH BLIIYD BOLTS

Introduction

Huck Internæional Inc. market a high strenglh blind bolting (HSBB) assembly withstructural performance intended to match A325 bolts. Figure I shows the unit inserted into

a holg both before and after tensioning. The tensioning operation consists of a hydraulic

gun being used to pull on the pintail while the gun $pages a collar onto the threaded bolt.

At the end of the operation a sleeve under the bolt head has deformed to prevent the head

from pulling back through the hole, and the pintail has snapped off

Figure 1: Huck HSBB (a) before tensioning (b) after tensioning

Due to geometry, a 20 mm HSBB unit (actually 21.5 mm) matches a f inch (19 mm)

diameter A325 bolt, and is used in a 22 mm hole in the HSS. This is less clearance than is

customary with 4325 bolts. The actual bolt within this HSBB is about 15 mm diameter.

Huck International reports that the 20 mm HSBB has minimum specified tensile strengÍh,

clampingforce,andshearstrength oî 192kN, l30kNandgl kNcomparedwith 178 Iò1,

125 kN and 98 kN (threads intercepted) respectively for I inch A325 bolts.

152

Huck International recently announced commercial availability of a re-designed HSBBknown as the ultra-Twist blind boii wr,icn is installeJ with a ,t"na"r¿-.Ëctric boltingwench (as used for. twis.t-off type uoitg rather tnan w¡tt¡ a hydraulic wrench. The ultraTwist is used in holes ft6 inch-íarger that the ourer ¿¡a.Lt", of the unit, which providesconventional clearances for fit-up. w¡tr:. these features, it is expected that erectors wi, findit a more attractive product than the original HSBB. i¡"-inrtull"tion sequence is shown inFigure 2' Independent tests have confirm-ed that ¡""u.ä pltension and ultimate rension ofultra-Twist %, % and r inch fasteners exceed the requirements of A325 borts.

Japan seems to be the-biggest potential market for the ultra-Twist, and Huck are expectingapproval there that will mean tire product conforms ,o r.pun', high tension bort standard.

The U-TRA.TV/¡ST btino ¡3¡ ,g

¡nstat¡ed lrom one srde oí lheslfuclure by a srnole opefatof.lhe ¡ns¡al,ation loot is hestandard eleclric shear rrrenchtooling used for rnslaltatron olTwist.olT Controt f¡-C.| tv0efasteners. The f¿stener ¡s

insenec and lne toot sng¿qs6

The bacKrde buto is fuilyformed rn the ai lo a uniformdiameter regardtess of gflp.

As the instaltaton ¡oad

increases. a spectal ¡ntern¿lwashef sheafs ailorMng thebackside bulb lo come tntoconlact with the work surfaceand lot All Clamp load to go rntothe work slructure.

Conlinued torquing of the unitdevelops the required clamp andthe lorque pintail snears of.completing the instailat¡on.

Us;ng a standard S60EZ shearwrench. ¡nslal¡alron l¡me for a3/4- faslener is agpror¡mately30 seconds.

Figure 2: tnstallation sequence for Huck ultra-Twist blind bolt

Experimentation

f#!j íå írfå::,ÍÌ::"" HSBBs both individuarv and in end prate momenr conne*ions

In tension tests of rigid-butt plate connedions, 20 mmHSBBs and 3/ inch ,\325bolts bothallowed separation-:Iht plates to begin at a load about equar to the specified pretension.Thereafter' the HSBBs blhaved -¡n

I ror. ductile manner (as the Hsng componentsdeformed) than did the A325 uolts. eli.xceeded specifieJ urri,n.r. tension strength.

Moment connections using w360x33 beams bolted through I g mm end plates to 203x203x12'7 Hss were used to tãtp*" moment-.otation behavturs of connections with 20 mmHSBBs and 3/ inch 4325 úott' iî. results were essentiaily identicar we¡ beyond thenominal plastic moment capacity of the beams. rd¡'¡y ru'nrlcal well

153

When similar moment connections using 254x254 HSS with 9.5 and ll.l mm walls were

tested, it was demonstrated that punching shear of the HSS wall around the HSBBdeformed sleeve under its bolt head becomes a consideration for ultimate strength.(Howeveç one suspects that overall deformation of the connection would govern.)

A¡rother report by the same authors (Ref. 3) includes resutts from the testing of trvoadditional 254x254xll.l HSS specimens. One had a 6 mm doubler plate welded to the HSS

face at the connectiorl and the other was filled with concrete after the connection was

complete. Both (especially the concreted one) showed increased initial stiffiress and fargreater post elastic stiffiress compared to the previous specimen of the same sÞe,

Tabuchi et al (Ref.4) also tested Huck HSBBs, both individually and in full scale moment

connections using either tees on beam flanges or end plates. Connections incorporated fourangles welded around the HSS column as shown in Fþre 3. The HSS was 300x300x16,

the angles 200x200x25 (trimmed to fit with their toes welded to the HSS and partially

together), and the beam was 45Ox200 mm. Design formulae were developed and verified.

A two storey building in Japan nno bays (15 m total) wide by six bays (3S in total) long was

one of the early structures erected using the above type of end plate moment connections.

Conclusions

Korol et al concluded that the HSBB moment connections weÍe similar to those using ^325bolts, in terms of stiffiress, moment capacit¡ and ductility

Tabuchi et al concluded that the ratio of separation load to preload of HSBBs is about 0.9;

that the strength and prying action behaviour of HSBBs is comparable to Japanese high

strengfh bolts; that the connections exhibited excellent hysteresis loops; and that moment

connections with end plates were superior to those with tees on the beam flanges.

SHS column

Figure 3: Schematic of moment connection

154

Figure 4: Hollo-BOLT fastener

HOLLO.BOLTS

IntroductionLindapter International (Bradford, U.K.) produces expansion bolts marketed under the

name Hollo-BOLT that are intended for hollow structural section blind bolting. Their

con-figuration is based on a truncated cone with interior threads to accept a high strength

bolt as shown in Figure 4. The 3-piece assembly is inserted into holes in the steelwork and

tightened with conventional tools to draw the cone into a mild steel split sleeve that flares

out to anchor the bolt within the HSS member. A collar on the split sleeve has two flats on

its edge for holding if the unit is inclined to turn during tightening.

Hollo-BOLT was introduced in mid 1995 as a successor to Lindapter's Hollo-fast Inserts

that are similar in action to the Hollo-BOLT. The main difference is that the sleeve of the

Hollo-fast Insert did not have a collar, and the sleeve with its cone was lightly hammered

into a matching hole in the HSS until the outer end of the sleeve was flush with the outer

surface of the HSS. Then the section to be connected was positioned, and the bolt installed

through a normal size hole in that member. The increased shear strength of the Hollo-BOLT

Ooth the bolt and the sleeve are in the shear plane) and the easier field installation make it a

more attractive unit than the earlier Hollo-fast Insert.

Development is continuing with various washers to ensure that Hollo-BOLT connections

are watertight, a fact that suggests the installation pretension is less than that of a

conventional high strength bolt.

Exoerimentation

A research program sponsored by CIDECT @ef. 5) at Lindapter International and BritishSteel in the U.K., and at the Universities of Trento and Genoa in Italy was undertaken in1995 to quantify the strengfh and utility of Hollo-BOLT fasteners. It continues in 1996.

Shear tests of Hollo-BOLTs have only been completed for Ml2 bolts (12 mm diameter), inmaterial from 5 to 12.5 mm thick. All results were approximately mid-way between thestrength of 4325M bolts with threads intercepted by and threads excluded from the shear

plane.

Tension tests show two types of failures. For l40xl40 HSS with walls less than 8 mm

thick, the material distorts and the bolt anchor eventually pulls through, but only afterexcessive deformation of the HSS. For thicker walls, the ultimate failure is by shearing offof the bent legs of the insert between the inside edge of the hole in the HSS and the cone ofthe Hollo-BOLT, apparently at loads larger than those specifìed for 4325M bolts.

Since testing is continuing, conclusions are not available.

15s

FLOWDRILLING HSS

Introduction

The Flowdrill method of creating holes in steel involves the use of a tungsten carbidesmooth-sided drilling bit that tapers from a point to a diameter the size of the intended hole.

Contact of the high speed rotating bit against the work generates heat to soften the metal so

that it extrudes to form a protruding "sleeve" firsed to the inside surface of the tube as thebit is forced through the wall. The hole in the wall and its "sleevd' are then threaded with arolling Flowtap tapping tool, without removal of materiat to accept a conventional highstrength bolt as shown in Figure 5. In effect, the hole and "sleevd' are a nut for the bolt.

The Flowdrill bit in cross section is actually not perfectly round, but some$'hat flattened onfour sides to produce four lobes as indicated in Figure 6, a shape that aids the elÉrusionprocess as the metal of the hole is displaced. A slight upset or boss is created on the outsidesurface of the material, but that is removed as part of the drilling operation, while the metalis still soft, by the use of a bit incorporating a milling collar.

Continuing research programs are presently investigating the use of Flowdrilling forstructural bolting of hollow sections.

Figure 5: Samples of bolts in Flowdrilled holes

17" -,tlVo

shank

collar

pol.vgon shapedstraight bod-v

polygon shaped cone

point

156

Figure 6: Flowdrill drilling bit

Experimentation

Flowdrilling for bolted HSS connections was examined in 1989 by Sherman (reported in theAppendix of Ref. 6), in I993 by Banks (Ref 7), and in 1995 by Éailerini, Bozz,o Occhi, andPiazz¿, (Ref 8 and Ref 9).

Sherman evaluated ftinch to I inch diameter A325 bolts in HSS having wall thicknessesranging from one half the bolt diameter to approximately one third the bolt diameter (that is,d/t ratios from 2.0 to about 3). In all cases, the bolt sheâr strengths exceede d o.Tztimes thespecified ultimate bolt tension, whether the bolts were just snug tight or were pretensioned.Tensile strengths exceeded specified bott tensile resistances foi ¿i bolts excep t for l( inchones, which were loose fitting (apparently as a result of a combination of metricFlowdrilling tools and imperial bolts).

Banks investigated Flowdrilling for M20 bolts (20 mm diameter) used in HSS walls from 5to 12.5 mm thick.

Threads produced by the Flowtap tool matched ISo profiles (except that the crown of eachth¡ead was somewhat incomplete) and were metallurgically sound with good toughness. Asubstantialincrease in the strength of materialaroundilowdrilled holes rJsulted frõm partialrefining of the microstructure in the th¡eade d area due to heat generated by the piocess(approaching 8000 C). Thickness of the parent metal had little effect on the length of theextruded "sleeve", which was generally I I to 13 mm long. Rather, the increased amount ofdisplaced metal from thicker material produced "sleeves"-with thicker walls.

In direct tension tests, bolts in 8, l0 and 12.5 mm thick material exceeded tensile strengfhsspecified for lvl20 .^325M bolts. Those in 6.3 mm material failed at 93 yo, and those in 5mm material failed atTl yo of the specified bolt tensile strength. Bolt shear íests in the samerange ofHSS wall thicknesses all exceeded bolt specification requirements.

Ballerini er a/ (Ref 8) closely examined the Flowdrill process ar the University of Trento inItaly by making threaded holes for Ml6, Mt8 andtitzo bolts in each of HS-S having 6, gand I0 mm walls. Material was 280 to 340 MPa yield (440 to 4g0 Mpa ultimate).

Hardness testing conducted on thread material gave values always within the rangespecified for structural nuts, confirming that beneficial hardening results from the heatgenerated by Flowdrilling. Optimum drilling parameters (using a ¿ iw power drill) were inthe range of 700 to 1600 r-p.m. for speed, ánd 0.1 to 0.15 mm/rev. for spindle feed rate,resulting in rapid hole drilling. The average length of effective thread in å, g and I0 mmmaterial was 72.4, 15.3 and 17.5 mm respecdv;ly and was only slightly sensitive to thediameter of the holes. Flatness of 6 mm *uik in 14ó mm square HSS was not affected, evenwhen Flowdrilling for M20 bolts.

157

I

,l

Water tightness trials of Flowdrill threads treated with a removable sealing product were

conducted with a 1.5 m water head (calculated to represent a thermal gradient of about 40o

C) for 30 days. This demonstrated that both water infiltration and orygen renewal inside

HSS can be prevented where Flowdrilled holes are used. No leakage was observed.

Ballerini el a/ CRef. 9) also performed a series of tests on Ml2, Ml6, Ml8 and lvl20 bolts

lwitfr strengths similar to 4325M bolts) used in HSS walls from 5 to 12.5 mm thick. They

examined thread stripping of Flowdrill holes, plus tension failures and shear failures of bolts

in Flowdrill holes.

The only Flowdrilled holes that failed by thread stripping were those where the ratio of boltdiameter to material thickness (d/t) was2.9 or greater. It is suggested that a mæ<imum value

for d/t of about 2.6 wrll ensure failure by bolt strength" not thread stripping.

Tension tests were performed using one bolt in the middle of the wall of a 140 or 150 mm

square HSS, both with the bolt in a Flowdrilled hole and with the bolt conventionally

installed including a washer and nut inside the HSS. Loading that produced wall distortion

of lYo of the HSS width (commonly accepted as the serviceabilþ limit) showed the same

results for bolts in Flowdrilled holes and conventionally installed bolts. As the wall and hole

distorted in ultimate tests, bolts in Flowdrilled holes pulled out at lower loads than did the

conventionalty installed bolts. When d/t of Flowdrilled holes exceeded 1.5, the tensile

strength ofthe bolts was developed.

Load-slip diagrams from bolt shear testing showed that Flowdrilled connections have

somewhat greater stiffiress and less ductility than do conventional connections, presumably

resulting from the threaded hole being an integral part of the tube. Ultimate strengths of the

Flowdrilled shear connections exceeded code requirements, but were 4 to 5 % less strong

than were conventional connections. The authors suggested that design resistances be

lowered by a cautious 10 o/o for Flowdrilled shear connections.

Conclusions

Sherman concluded that Flowdrilling has potential for blind bolting to HSS columns. He

pointed out thar the fabricator would need drilling equipment with suitable rotational speed,

torque and thrust, (but Flowdrilling permits bolt field installation with conventional tools).

Banks concluded that Flowdrilling produces sound threaded holes suitable for use instructural steel connections; that effective thread lengths vary from 1.8 (for thick walls) to

3.0 (for thin walls) times the original material thickness; that current design procedures can

be used for predominately shear loadings; and that deformation of the HSS (not failure ofthe Flowdrilled connection) is the limiting criterion for moment carrying face connections.

Ballerini et al concluded that Flowdrilling allows for very simple bolted connections oftubular elements with the capacity necessary for profitable use in structural steelworks.

158

NAILING HSS

fntroduction

The joining of overlapping coaxial circular HSS members by the use of power-driven nails

was investigated at the University of Toronto by Packer and Krutzl er in 1994 @ef. l0). The

method entails slipping the end of a circular tube snugly inside the end of a larger tube, then

driving special nails th¡ough the overlapping wall thicknesses from the outside. Similarly,

fixtures or secondary members such as purlins can be easily connected to HSS with nails.

Exoerimentation

Equipment used was the Hilti DX750 direct fastening system consisting of a powder-acfuated gun using purple cartridges (highest power available) to fire ENPII2-LI5 nails.

Penetration settings ranged from 3 to 3.5 (on a scale of I to 4) to ensure that the nail pointpenetrated the inside surface of the inner to two walls (up to 13 mm total thickness).

Outer tube diameters were ll4 mm (nine samples), 102 mm (17 samples), and 406 mm(eight samples), all approimately 6.4 mm thick. For the first group, inner tube thicknesses

were 6.5 mm, for the second group,6.5 mm (5 samples), 5.0 mm (6 samples), and 3.1 mm(6 samples), and for the third group 6.4 mñ.

The fit of the first two groups was characterised as "tight", since light machining was

required before they could be assembled. Fit for the third group was "loose", with a gap

varying from zero to three mm (due to slight out-of-roundness of the manually fabricatedinner tube). The number of rings of nails and number of nails in a ring were varied. Figure 7shows one combination. A connection with ten rings developed the tube capacity. Thedistance from the end of a tube to the first row of nails varied from 6.4 to 25.4 mm.

The smaller, tight-fitting specimens were loaded in axial tension, which always led to an

abrupt failure. The larger, loose-fitting specimens were loaded in axial compressior¡ also toa sudden failure.

More recent testing has been completed to examine fatigue behaviour and whether the nailstend to work loos'e under cyclic loading. Fatigue performance was actually superior to thatof a symmetrical bolted lap splice and the nails did not work loose before cracks developed.

Results

Thefailures were all by nail shear except the six specimenswith tubes having 3.1 mm wallthickness (plus a specimen having 5.0 mm wall thickness combined with ó.5 mm end

distance), which failed by bearing or shear of the tube wall.

159

Figure 7: Nailed specimen in test rig

Figure 8: The nipple-dimple effect

The connections resisted loads beyond the shear strength of the nails, about 2Ùo/o more forloose fitting specimens and 3O%o more for the tight fitting specimens, before nail shear

failures. This additional or secondary strength resulted from a "nipple-dimple" effect at the

interface between the tubes. A nail emerging from the inner face of the outer tube created a

nipple protruding from that surface that interlocked with a matching dimple created in the

outer face of the inner tube. Figure I illustrates the phenomenon.

Offsetting the nails of one ring from those in an adjacent ring or having more nail rings withfewer nails per ring (for the same total number of nails) had little effect upon the shear

mode of nail failure or the connection strengh.

Conclusions

The structuraljoining oftwo overlapping coaxial circular HSS by the use of power-drivennails was shown to be both feasible and economical.

The ultimate strength for connections that fail by nail shear can be taken as the number ofnails times the single shear strength of the nails. This consen'atively ignores the secondary

contribution from the nipple-dimple effect.

160

1.

3.

4.

The ultimate strength for connections that fail by bearing or shear of the HSS material is

conservatively given by the expression 2.4 d t n Fu when the end distance is at least 1.5 d

and the pitch oithe nîlr 1"long the HSS axis) is at least 3 d, d being the nail diameter' / the

HSS wall thickness. n the number of nails, and Futhe tensile stren-eth of the HSS material.

REFERENCES

packer, J.A.; and Henderson, J.E. 1992. Design guide for hollow structural section

connections. CISC, 201 Consumers Road, Suite 300, Willowdale, Ontario, M2J

4G8.

Korol, R.M.; Ghobarha, A.; and Mourad, S. 1993. Blind bolting W-shape beams to

HSS columns. ASCE Journal of Structural Engineering 119 (12): 3463 to 3481.

Ghobarha, A.; Mourad, S.; and Korol, R.M. 1993- Behaviour of blind bolted

moment connections for FISS columns. Proc. 5th International S]¡mposium on

Tubular Structures, eds. M.G. Coutie and G. Davies, University of Nottingham,

T"^tiifffi:änu,"ni, H; ranaka, r.; Fukuda, A.; Furumi, K.; usami, K'; and

Murayama, M. lgg4. Behaviour of SHS column to H beam moment connections

with óne side bolts. Proc. 6th International Svmposium on Tubular Structures, eds.

P. Grundy, A. Holgæ- and B. Wong, Monash University, Melbourne, Australia'

Occhi, F. 1995. Hollow section connections using (Hollo-fast) Hollo-BOLT

expansion bolting. Second Interim Report, CIDECT program 6G'16195'

Sherman, D.R. 1995. Simple framing connections to HSS columns. Proceedings.

National Steel Construction Conference, AISC, San furtonio'

Banks, G. lgg3. Flowdrilling for tubular structures. Proc. 5th International

Symposium on Tubular Structures, eds. M.G. Coutie and G. Davies, University ofNottingham, United Kingdom.

Ballerini, M.;Bozzo. E.; Occhi, F.; and piezzl,lly'r. 1995. The Flowdrill system for

the bolted connection of steel hollow sections--part I: the drilling process and the

technological aspects. Costruzioni Metalliche, No. 4, July-August, Italy-

Ballerini, M;Bozzo, E.; Occhi, F.; and pinzzv,lvl. 1995. The Flowdrill system for

the bolted connection of steel hollow sections--part II: experimental results and

design evaluations. Costruzioni Metalliche, No. 5, September-October, Italy.

Packér, J.A.; and Krutzler, R.T. 1994. Nailing of steel tubes. Proc. 6th

International Symposium on Tubular Structures, eds. P. Grundy, A. Holgate and B'rJ[ong, Monash University, Melbourne, Australia.

5.

6.

7.

8.

9.

10.

161

,i

FABRICATION AND INSPECTION PRACTICESFOR WETDED TUBUT,AR CONNECTIONS

J. W. Post*

ABSTRACT

Producing a simple welded tubular connection in steel consists of cutting and coping themembers, fitting, welding, and inspection. However, fabrication and inspection practices forsuch connections and their related costs are greatly impacted by design choices. Often,these choices are made by designers without a full appreciation of the costs that will beincurred by the fabricator or erector in producing such a connection. This paper willaddress the major choices to be made for tubular connections and their significance to thefabricator or erector.

INTTODUCTION

Tubular structures with welded connections provide architects and designers with elegantsolutions to steel framing. They range from simple highway sign supports to giganticoffshore drilling platforms and include aesthetically pleasing space frames seen inconvention centers, sports arenas, airport terminals, and atriums. Tubular members offerthe designer an efficient cross-section relative to their inherent material distribution forbeam bending or column buckling calculations. With appropriate regard for the connectiondetails presented herein, efficient and cost effective tubular structures can be achieved.

Before we consider typical fabrication and inspection issues for tubular connections, it isbest to first review several imponant design issues and how the choices designers ordetailers make can impact fabrication and inspection costs.

For this discussion, round tube or pipe will be considered as synonymous while the familyof hollow structural shapes with a square or rectangular cross-section will be collectivelyreferred to as box tubing.

DESTGN CONSIDERATIONS

Round Versus Box Tubing

Architectural considerations or availability usually govern the selection of round versus boxtubes. For larger sized members, box tubes would need to be fabricated from plate. Thisusuallv drives the costs high enough so that round tube or pipe are chosen. For small tomedium sizes of members, there is a wide varietv of thicknesses and dimensions available

* J. W. Post & Associates, Inc., Humble, Texas

I

162

for both shapes.

Where box tubes can be used in orthoginal planes they offer several unique benefits overtheir round counterparts. Box sections are easier to handle and stack. They are easily cutand mitered with band saws or abrasive saws since no complex copes or saddle cuti arerequired, which always occurs when a box or round tube intersects a round tube. If branchmembers overlap each other in a truss assembly, compound miter-cut box-tube members canbe inserted and slid sìdeways into place. With round tubes, overlapping connections preventsome diagonal members from being installed as a single piece. For those cases, stubs or"windows" (insert segments) may be required to facilitate member installation. A detaileddiscussion of stubs and windows is given in Reference 1 and box-tube assembly in Reference2. Also, box-tube members can easily accept backing material, a point that wiil be discussedfurther in the following sections.

Matched Versus Stepped Tubular Members

Matched-box connections are defined as a connection created by the intersection of rwo ormore box-tube members that have a common outside dimension and arranged as shown inFigure 1 so that the sides of the branch members are flush with the sides ãf the chord orthru member. By contrasL a stepped-box connection occurs when at least one dimensionof the branch member is smaller than the side-to-side dimension of the chord.

The significance of stepped versus matched-box connections occurs in several areas. First,following the AWS Di.1 Structural Welding Code - Steel (Ref. 3) prequalified detaits forfillet weld categories can only apply to stepped-box connecrions wherã the wídth of rhebranch member is less than or equal to 80Vo of the chord member width. This limitationensures that the side fillet welds occur on a flat face and not on the rounded corners of rhemain member.

In matched connections, careful consideration must be given to wall thickness of bothmembers. For instance, a designer selects a TS 4 x 4 x lf2" chord member to carry thedesign loads. Suppose some branch members are carrying small loads, so a TS 4 x 4 í l¡g,'is selected- The inherent problem here is the cornei rãdius or corner djmension of iirechord member. The ASTM standard for 4500 structural tubing limits rhe corner radius tothree times the wall thickness of the tube. Consequently, the thicket the wall, the greaterthe corner radius- MgtJ 4500 tubing ìs produced by coniinuous forming and weldin! stripsof steel into round tubing. After welding, it is drãwn through dies rolroduce final-sizådround tube or through additional sets of forming rotls to produce square or rectangular tube.When round tube is formed into box sections, the reiulting .oin"t radii usuaîy do notmerge tangentially with the side walls. This trait of box tubes is more noticeãble withgreater wall thícknesses.

Figure 2 depicts the significance of the corner dimension in matched-box connections forthe example cited. Notíce that the branch member musr be coped to fit the large curvature.Otherwise, a very large gap or weld root opening will o.órr ar the side

-zones. For

comparison, without consideration for structural loading, if the chord member was replaced

163

by a TS 4 x 4 x 7f8" member, the corner radius would be much smaller and the problemmitigated. Mismatching wall thicknesses lead to more difficult welding on the side zonesusing either complete joint penetration tClPlgroove welds or even partial penetration [PJP]groove weld details due to the larger corner dimension. There are however two goodalternative solutions for the example given. The most obvious solution would be to reducethe size of the branch member since it is so lightly loaded. For example, a TS 3 x 3 x 3/16might carry the same load while providing a stepped-box connection suitable for fillet, PJP,

or CJP weld details. The other solution is to cut a backing plug or ring as shown inFigure 2. This plug can serve several functions. It provides backing for welding whichmeans welder qualifîcation requirements are reduced for CJP connections. The plug alsoprovides for variation in fit-up tolerances in both the CJP and PJP cases. This is especiallyhelpful for field welds. For some erection sequences, the plugs can be shop installed on thechord members which facilitates rapid and precise positioning in the field.

Sometimes designers will select a common tube size for a truss for aesthetic reasons whereonly variations in the wall thickness occur. There is however another hidden benefit inchoosing stepped-box connections over matched-box connections when aesthetics areimportant. With matched-box connections using either CIP or PJP details, it is difñcult toproduce flat appearing welds in the side zones without a lot of costly cosmetic grinding.This problem does not exist with stepped-box connections. With stepped-box connectionsrhere is a natural ledge to support the weld beads of either fillet or groove weld details.The one drawback to stepped-connections is the inherently lower strength of the flat faceof the chord member as determined by yield line analysis. See References 4 and 5 forfurther design guidance.

Gapoed Versus Overlapped Tubular Members

A gapped connection is one in which two or more branch members intersect a commonchord member with some nominal space between the branch members as shown inFigure 1. By contrast, an overlapped connection occurs when two or more branch membersintersect each other. Gapped or overlapped connections can occur in both round or boxmembers in either matched or stepped-box connections. The significance of these variationsis that the gapped connections are always easier to fit with better access for welding and

inspection while the overlapped connections usually require compound copes or miters andprovide no flexibility as to member installation sequence. With gapped connections (usuallya 2" nominal gap) the branch member can be moved slightly about its work point to improvethe overall fit-up and root openings. This luxury does not exist with the overlappedconnections. Any slight shift to improve fit-up ïn one direction causes a worsening of thefit-up in the other direction. One significant drawback to gapped connecdons from a designstandpoint is rhat all branch member loads must p¿tss into the chord. This may requireheavier chords. Conversely, the overlapped branches may pass some or all of their loads

directly to each other without affecting the chord member size.

Knife-Edse Gussets

Some cJesigners or detailers feel that the use of shear plates or knife-edge gussets is the

164

surest solution to a tubular connection problem. Indeed, the gusset-plate approach has been

used successfully for many years. However, for aesthetic applicæions, the gusset plates

make the connection appear awkward, and cluttered as shown in Figure 3'

From a fabrication standpoinr, the gusser plate concept added to coped branch members

require exrra parrs (more material and weight), added cutting costs. for both the gusset and

associared slots, more welding (albeit less skilled panial penetration or fillet welds), and

,,,or. blasting and painting. Also, the fitting advantage of box tubes where coped members

can be slid sideways into final position is precluded.

Open ended branch members are especially unsightly for exposed applications. They also

piovide additional painting and maintenance problems'

From a design standpoint, the gusset plate approach may spare the engineer from dealing

with unfamiliar design rules but, the gusset plates usually provide high stress concenffations

or "hard spots". Thãse occur at the ends of the gusset where it attaches to the chord and

ar the end of the slor in the branch member. Such details are particularly susceptible to

cracking in fatigue as shown in Figure 4.

CIP Groove Welded Connections

With the preceding design choices made, the designer may next select the appropriate joint

rypes and joint details in accordance with the requirements of AWS D1.1. The choices are

CJf gtoou. welds, PJP groove welds, and fillet welds. The designer may further detail the

rp..i-fi. groove angles and root openings or more often, this task is left to the steel detailer

or the fabricator with a simple (but costly) note on the drawings that states, "All welds shall

be CJP unless noted otherwise." However, the choices related to weld types can have a

significant impact on costs of the completed tubular connecdon related to coping or

mitering, fitting rolerances, welder's skill level required, accessibility for welding and

inspection.

CIP groove welds are rhe joint detail category most frequentlv selectgd, lut not usually the

most-economical one. Often CJP groove welds are selected by default. That is, no detailed

consideration is given ro them. It is generally felt that CJP groove welds must be better

than PJP groove welds. In fatigue loading situations, this is true. Consequently, engineers

or designJrs choose CIP's even for cases not driven by fatigue. Granted, CIP's using the

AWS Dt.l pt.qualified details will develop the full strength capacity of the connection but,

PJP groov.'*.Ídr using E7018 or E71T-X weld metals on ASTM A-53 Grade B pipe or

ASTM 4500 tubing will also develop the full strength of the connections in most cases. The

problem here is thãt the D1.l Code may require the designe¡ tod_o_some additionalstrength

lhecks. Even on smaller projects, the costs of gearing up for CJP groove welds (e.g- 6GR

tests for welders) will likely exceed any extra engineering costs.

CJP groove welds for tubular connections, whether round or box, implies open root

conditions and requires more precision in fitting the members and requires the highest

welder skill levels to produce a qualiry weld. In order to achieve complete joint penetration

165

I

iI

I

,i

from one side without backing, the D1.1 Code specifies that the open root dimensions mustbe closely controlled and the minimum groove angles must be assured. AIso, the weldersmust be capable of this most difficult welding and demonstrate their skills by passing the6GR open root welder test. For box tubes, a special corner welding test is an additionalrequirement.

One previously mentioned benefit of box-tubes with their flat sides over pipe is their abilityto accept backing rings or plugs. With appropriate backing, the open root difficulties vanish.The welder qualification requirements drop back to the easier 3G + 4G requirements whichwere derived from test coupons welded with backing. Also, a greater variation in fit-up ca¡be tolerated without unduly affecting welding quality.

The AWS D1.1 Code requires continuous backing whenever backing is to be used.Commercially available rings are produced for most pipe sizes. Some fabricators choose toform bar stock to fit the inside of the pipe or box tubes. Howeveç any butt splices in therings or bars must be welded ltÙVa to prevent crack initiation from any unwelded butt splicein the backing ring or bar. In a few unique cases, a smaller size pipe or box tube can befound and cut into appropriate rings without the need for making the butt weld in the rings.Designers and fabricators should consider this option if possible as it is the least expensivewav to provide continuous backing. For instance, a TS 3-1/2 x 3-l/2 x'l.f 4" will fit snuglyinto a TS 4 x 4 x 7f4" member or loosely into a TS 4 x 4 x 3/16 member wirh minorgrinding to remove the ID weld flash from the 4" member. Some fabricators cut plugs witha photoelectric tracing head and machine cutting torches or NC progr¿rmmable cutringmachine. This provides one-piece backing without the need for lAÙVo butt welds in therings. These plugs may be solid or cut hollow where heat sink or radiography are aconsideration. In a previous paper (Ref.1), it was suggested that such plugs could be cut ona bias with a beveling head attachment added to a machine cutting torch to produce branchmember backing for other than the simple 90" T-connection cases. Figure 5 illustrates someof these continuous backing types.

P.IP Groove rilelded Connections

PJP groove weld details for box-tube connections can offer significant cost savings in severalareas; groove bevel preparation, fitting, welder skill levels, and inspection. In preparing abranch member to fit into a truss for instance, the miter cutting would be the same foreither the CIP or the PJP groove weld case. The next step is to prepare the necessary bevelangles to comply with the prequalified groove details. The PJP groove angles required aremuch less demanding and the differences are most notable in the heel zone where the localdihedral angle Psi ( I ) is in the range of 30" - 60". In this range the CIP details requirea full bevel preparation that is at least one-half of the local dihedral angle. In a common45" case for instance, the bevel preparation angle is 22V2" which leaves a fairly thin andsharp bevel. In the worse case of V = 30o, the bevel preparation is a 15o sliver of metalthat is very difficult to produce and is easily melted away when trying to make a qualiryroot-pass. For the PJP case on the other hand, the heel zone for any I in the range of 30"- 60' requires no bevel preparation beyond the natural groove formed by the intersectingmembers with only a miter cut. Of course, the side zone and the toe zone may require

166

some bevel preparation. but none with the very thin and pointed bevels as found in the heelzone of the CJP cases.

In the area of fit-up, whether done in the shop or the field, the PJP groove weld detailsoffer still more advantages over their CJP counterparts. As previously stated, the AWS Dl.1prequalified details require close controls on groove angle and minimum-to-mÐ(imum rootopenings in order for welders tested to a higher skill requirement to achieve complete jointpenetration from one side without backing. With the PJP's, there is a ma-nimum of 3/16"on the root opening. but the minimum is zero. This means that the steel may be broughtinto tight contact. which is the easiest case to fit-up. Further, PJP groove welded boxconnections could be fit u,ith similar backing material as discussed in the previous secrion.This would aid in fit-up and alignment tolerances, especially for tie-ins or fieìd erectionsituations. Such cases would fall outside of the prequalified limits when the root openingexceeds 3f76", but with backing, such modified details would be easy to qualify withmockups or sample joints.

Fillet-\4'elded Connections

Fillet-welded tubular connections are usually easiest to product and therefore the lowest costfrom a fabrication standpoint since the prequalified detail requirements of AWS D1.i arethe least onerous. For pipe the branch member diameter must be no more than 1/3 of thechord diameter and coping is still required, but the only beveling necessary is in rhe toezone r¡'hen V exceeds 120". For box tubes, only simple miter cuts are necessary. The filletdetails are applicable to anv stepped-box connection provided the branch member width isless than or equal to 80% of the chord member width. Prequalified details require thebranch member and the fillet weld to be kept on the flat face of the chord member. Thiscould be a problem with thicker chord members that may have a larger corner radius orcorner dimension. For heavy-wall box tubes this detail should be checked out prior tofabrication.

The prequalified fillet details are permitted down to Theta ( e ) brace intersection anglesof 30" which is identical to W when measured in the heel zone. This covers the vastmajority of structural cases. The root opening may vary from 0 to 3f 76" ma¡<imum providedthat the fillet size is increased by the amount that the root opening exceeds 7f76".

. FABRICATION PRACTICES

Cut and Cooe

When a tubular branch member frames into another tubular member, a connection iscreated. TYK-connection is the term referring to any one or combination of branchmember intersections. The branch members usually require some type of a cope or mitercut. For round members. the copes are more complex than for box members as shown inFigure 6. Also. compound copes in the case of overlapping members add to the complexitv.For box members, machine saw-cuts can be used to produce miter cuts to which torch-cut

167

andf or ground bevels can be added. Careful grinding is also required to provide smoothtransitions from one groove detail to the next that always occur at the four corners of eachbox-tube branch member. For round members, the conventional method for coping involvescreating a wrap-around template to mark the pipe and hand cutting with an oxy-fuel torch.The templates can be created using a drafting technique of circular intersection projections(Ref. 6). An individual template is required for each combination of branch memberthickness and I.D. versus main member O.D. and intersection angle. Once generated,however, these templates may be used again and again. Presently, computers can be usedto generate the coordinates for these templates and, if large enough plotters are available,the template may be computer drawn. Further guidance in developing accurate templatesand computer equations can be found in Reference (7).

Hand cut copes from wrap-around templates generally require two cuts. The templaterepresents the I.D. intersection of the branch member with the main member but" it isdrawn on the O.D. surface of the branch member. The first cut must be madeperpendicular to the pipe's surface with the torch always pointing toward the axis of thepipe. In this way the template outline is successfully transferred to the branch member'sI.D. surface, which is the true intersection with the chord at the root of the weld. A secondcut is then made with the torch tipped at varying angles to produce the required bevel forwelding. This is the difficult step in that the burner or fitter often must sense or feel theproper bevel angle without blowing away the tip of the bevel or "feather edge" at the I.D.surface. Sometimes these angles leave a very thin edge that is easily melted or gouged.Significant grinding and touch-up work is often required to produce suitable coped andbeveled surfaces appropriate for quality welding.

For manual coping, computer programs have been enhanced with the aid of local dihedralangle input (i.e. Appendix G of AWS Dl.l and Reference (7)) so that the program can alsogive the coordinates for the entry point for the bevel cut thus taking the guess work awayfrom the burner. If he errors on the tight side, the welder cannot achieve the weldpenetration required; and re-work (gouging, grinding, or remove the member and re-cutting)may be necessary. If he errors on the wide side, very large weld grooves are produced andwelding man-hours rise rapidly especially on thicker branch members.

Mechanized coping devices for pipe have been available for many years. Some machinesare linkage and cam driven, while others may follow black lines on a white drum with aphotoelectric cell. The more recent machines are computer driven. Most all of themechanized coping devices incorporate automatic torch tilting, so that the proper bevelangle is cut in one pass, not two, as with manual cutting.

Common limitations of the mechanized devices are their O.D. capacity and the limits oftorch tilting, wherein the torch cannot lay over far enough for the most shallow angles foundin the heel regions of braces with small O intersection angles. Another limitation of thelinkage and cam driven machines is that they sometimes cannot be adjusted to cut theprequalified joint details found in AWS D1.1. However, alternate details may be tested andqualified by the fabricator. The most serious limitations in dealing with the computergenerated template or computer driven machine, is the knowledge of the computer

168

programmer. Too often the programmer does not have a^good grasp of the 3-dimensional

geometry involved in tubular connections. Regardless of how the cope is produced' it is

wise to check it immediately. Make a trial fit against in mating chord or use a 3-

dimensional template or model of the main member' In this manner' the accurary of the

cope, the groove ungl.. and the branch intersection angle can be quickly checked' Be sure

to'include the required root opening in this trial fit.

Fitting TYK-Con nections

From a fabrication standpoint, the rowest cost connections are those simple TYK's without

the overlapping r.rU.ri. If possible, design the connection with a two inch nominal gap

between the toes of rhe a jacånr branch mãmbers. This greatly-simplifies fabrication and

erecrion. Diagonal members can usually be adjusted slightly about the theoretical work

point ro compensate for inaccuracies in lengtú and poiition. The overlapped branch

connections always have a compound .op-. ?nd r-equire more .careful layout and

.utting/Utveling, Jtp..iutty for length. For pipe, the sequence of member installation must

be plinned an{controlled to minimize the need for stubs or windows'

Welding Processes

The welding of tubular butt splices and TYK connections utilize the same group of welding

processes fi*iUu, to structural shops. SAW is routinely used for long seams in pipe where

a fabricator produces his own pipe. The process-is also used for tubular butt joints (girth

welds) and, with smaller diamåtår electroães and flux dams, it has been used down to six

inch diameter. SÀw has prequalified srarus for diameteÍs 24" and greater. Below 24"'

luãtin.ution resring on the smål.rt diameter to be used in production is required'

SMAW, FCAW (both self-shielded and gas-shielded), and GMAW have been used

successfully for tuny years on tubular co-nnections' The SMAW process is the old

workhorse with a íuíg.- selection of electrode types, alloys-,. sizes, and operating

characreristics. It i, urry"".onomical in original equipmãnt cost and is very portable, but the

cost of the weld metalâeposited is high õ.puréo io semi-automatic processes due to its

to* o.porition effici"n.y.'GMAW in ipray tiansfer mode is rimited to flat and horizontal

alpticarions. GMAW-í(rnott circuiting trânsfer) is good for thin materials less than three-

;ïáilr of an inch and for root passes ih.t" poor fit-uP may be present' The short arc

process does require more weldlr skill and ulira-clean bevel faces (sandblast or grind) to

rninitir" inherent cold-lap tendencies'

FCAW-G (gas-shielded) is a good all position plo^.9::.and weld metal depgsition-rates are

significantly higher than thosJ of SVIÄW- The FCAW-G process and the GMAW require

an auxiliary gas shield and a gas cup on the head of the welding gun to deliver gas to the

*.1ãing ,oü. This gas .up o.td* a visibiliry problem for the welder. It also prevents access

ro the root 'f the joi-nr witL thicker beveleá members or tight inters.ection.angles. Also,.the

gas is easily clisruibed by drafts. ancl wind, which limits its use in drafty shops or field

õonstructioñ sites without providing for suitable wind breaks'

169

FCAW.SS(self.shielded)'ontheotherhand,has.someoutstandingfe.aturesfortubularconstruction and däJ;;í'r"quir" ,h;:îtiä;-*il'"".t*ä' ali åi l" shietding is produced

at the arc by ,t.-úurning of -rom. of ir, .õr. ingr.d-ieîts but' more importantly' the

remaining u,n.,orpr.,Ji.'iäñ,"r"¡nun* ö;1;;d tîtt"g"n) that iÏ Tt displaced bv the

burning action J-;;icalry ."rbì;;'f *i,h ulrr*inî,,,'to form¡xides and nitrides'

Therefore, this prol"ss is immun¡ ,"rir but the "r""täriär

iinãr' .T.he welder has equal

or better visibilirfi;il ii'. sr,¡aw;;id", ""d the ;;;;r ;ã"r1''. r-r1ve an accessibiliry

;;"b,Fi1,:rl,-',f ..î.,ï¿:':.]",iTüXtr';n'*tiÏï$.':"':"i"'"\'ï'trJ:[::stopping to cnange :1":::;.J1", saos in fit-up. However' rnc ptur't

electrode .*,.nr,åi, to tun¿r" -igi ö.'in tiu"p Hãï.u.r, the process may be too hot

for pipe or tube *ittt tttinner wall thicknesses'

Welding Procedure 0ualification

There is a fam'y of prequal,fi"-9,rrt* details_for.fiK-connections provided in section 10

of AwS D1.1 suiiabi. ró, ur. *i,r,'irüÃü,-rcnrri ìJ"äilwsj s,lw can usuallv be

done using the uppìi.uur. rTï:ri*ä';il joint o-äirr i*r¿ in Section 2 or the code'

Even though tn. iåint details ,n", ;.ïP;;;ä;' ci'iîw:ð-ntu"t has prequalified status

and must always'be qualified by testing'

rr there are orher job-speciric requi11m"',: î:l:å;i:i.'î""tlå:ñi:, n :!î"r""T"1

;'#;å" ; i¡1i'5:: ï I n:',:'ru"'-f;¡Ëi,..,.1i.! I'ii.Ë,¡; I 1åï" ;' sroov e an gre and

angles less than :u'.:'1::'*'-';;"; even when the joint rs orr€rwtsc

grearesr groove äd,ffi;;;-t" ¿.i* Ërãn'*r,"n ,ri" iãi"tìr"otherwìse prequalified'

For the mosr common structurar steer pip." g, tubular connections, with grooves 3-0] or

srearer, un¿ *r,Tr. no other ¡"u-öããiti". îi*itutio.] ãn werding procedures exist' rt rs a

îe r ativery s i mpre marte r,o or.pur.îää ;iä.';l if¿d *" ilì;;;;*'du " specifications ror

tubular app'caiions. [t is, howet;;';;;;ïinitutt .o rin¿ quãtitieo welders'

Welder Performance Oualification

rhere are no prequarified werders'- E'u:h ::1Ï'"'*'"':ï:'"i::f#"ìff iiî',ir; î:¡'j'il;.are all p'opt'iuîãin"o and qualified by testrng'

ä,ee"i,,ri*,'::"'iå,ï*î'îif kff a;;.;:if

:i,'*[:.1'ä:'f ]iÍtllT'ni'"ilii'fiweld Progresstrore<l shape .ánr,ru.,ron in

. seuerar importanr IÌoä

" ñ'.t:"ir"i"ff:iË #r.tT

:ï:'r,ïîÏ!r'.-Lir*Uli¡"rr**iiffjäTåi::ii*ïtr.?üi"Ï:å'1trff lffi i:J"-'.iå:*::if :ç*ÏåÎÎT'l':'þäTTjf îft".'"î:,ï:nt:*:'r*'jO.,,i. *.r¿.r-.. ärlå-*ur, p"r,,rlïïtuît Ãnel1lit;îit'i *i'ith co¡¿ers theñ down to 15''

Such cases trequentty occur. ,",'.åäoì ,.lzs" nr"..î",.rr"riion for a cJp weld requtres

¿Z,'lzogrnnu.ängte in the rt -.1."'å"'

þurther' *ttätit î;;lìü o'n cJP welds in box tubes

musr ars. pass thã special 1nr1.î iru.ro.t.t, ,.r,.'äi, i.it .t'irr* it'rir ab'ity to deposit

souncr *eto meial ai<¡uncr tn. ,"iur¡uely- sharp .nr"år'ìrunsition zo.nes which are the areas

170

of highest load transfer across the weld'

l:î',J':":"ä'"î:å:,ff ö,,':iïi:'fi :iï11'äïå'îîlii:"#iil'ïJiJ:*ffi i"Î[

alternative and is usuatty.mu* "i:l"-;"r welders'o Ou'i] ot;'il tt-dãrd 3G + 4G with

backingi.u.."p*úì""íostrucrurar,'t'öä;;f ::"tJ,l'"Ji,':$lqpiåï,åiü','i

#rii*;'"fffi irup*;¡å:ïr.'iiï',ri.'ïiæ1'¿îã.,u',,ii''Acu'leAng'Ieri,Jrr",, qe".::Ilphp.räJ,:'frä:ïLïi:ioJåîf *n:lff':fsted

and used (i e '

less than 30')' then the Acutc rarró¡v ¡ '---

e*q"ri?,n"-1 lt1,.iî:,l;13J,::*f':i ;'äåiTi"î:i=li,Î¿:1""";3Ï1"í"Î¡i3ïå ;#'

high failure rate \

test by .o'p"'"i i' Inã !*p"i "n'"ä' i iil' :l-*:,nïîtffi iï:lüîi'ï:Ti|Ëi

å**::,:"m*r.:n:*r',1¡äî:lrqi'iiü',ffi ä;ö*'ã,"ti'ru'íori'Ivcompretes

a speciric qualiricri;;;ri can do ;"ï;ìil;?;' #'h"h;'; à"1rin"o' for all conditions

that might a'se during production î"1àing. ti is "rr"*i"i'ir'at

wet¿ers have some further

training and supervlslon' _ ^.,, ns. For all posirion

ff lç::ilJï:i,."ii:iËîff lil:î:îi:5i'liî-::'ff :.'illi,iiååî,""î.*iñ.á"er"

in the neet zone il'i5;; it'un oo"', For these cases' ;tï; 3G + 4G plate groove test rs

'.'.:multiù*Ëï*ï:fiis'+ nr*firy;¡g;i};''i;t.r'l¿:i,',',îï,:'T#ïiË:i:îi'n*"iiãcn

ror pipe or 6GR prus c

ü;tõöì.ï tãic¡P box-tube welding'

Fromafabricationviewpoint'itisclearlyb",l"l,::,usefilletconnecdonswherepermittedor pJp,s o, c¡p;r'with'backing *he'euér practi""uî" t"Javoid-T:nu of the difficulties

rypica'y "n.ou;,.räTirt, *"to,"e'iîJ";å;;;ã#ffä".ìr'' ocR iesting requirement'

INSPECTION PRACTICES

visual Inspiction ' rsoection method for

y"îïil,'îìT'""i:TLII:äi:'.=,'l::o:::;' jli'ï"1'åiïlo":"T'lË'¿"ïÏ'i'ä"wäi¿i'e

Inspector, U", îïr!"lly needs ,ô-t,*" "*pl'i"nt" with tubular connectrons'

The comperent inspector canêvaluate weld quarity from surface workmanship' He can

quickly ¿.,"rrnìn" .î;;;;; pi"rnå r..ptau'irv, r"Jri"g""ir irrr workmanship requirements

set bv tt. cnää';#ñ n''ncr-upllst'where specified'

171

For the cJp tuburar connecrions, more inspection effort shourd be praced on inspection of

the fit-ups prio, ,oï.ì;;;. i" ihi, *uy, th. prop.' 'oãi

optnings and groove angles can

beverified.Withoutgqo--dt9Ît'oråîiii-upi'-"ï"n-ttt"u'""*ãld"t'"illhauedifficultyoroducing crr groou";;ld, ot trre.exi;ää"""rtq^ 1.¡

u",*t to pu-t.!ll inspection effort

,ro front ano rorroJ-u-p ;ìh a good-visual inspection and perhaps re.quire. some random or

siot checking *ithîïü;; ãã Jr l"rp"ñs after *;rdilg. ttt" tit-op insoection seldom

leads to controversy because the ;;äp;ng and g'oãut ãngles are easily measured and

verified.

For comprete joint penetration gl99ue werds-of^theric"trest qualiry, it is essential that all fit-

ups be inspected.,'ñil;*p.i,¡t, iru".foJ..9nn1.19nt tnãt øir not.or cannot be tested

with uttrasoruc n,.riäãr'li-e. ttrinnli wa¡ thickness pipe or box tubes)'

Radiographic Examination

Radiography of tubular butt join-ts is practical T9j:::*mended whe11

1'surance of higher

quality i, n"..rrufr. î;ii;t:^f^t..rînieues are. ,o"in'tv ute! to cover-the entire diameter

range encounrer.á. Fo, diameters à;ú to 10", p-o'u*it shots are practicable' contact

shots are acceptabte down to r".-errîii:;i;? Ët"eht;äg** try T used on pipe3'/2"

or sma'er. Box tubes may requir" "åãitional

shols t" pi"pËrry interpiet the relatively sharp

corner radii'

Radiographyisnotpracticableforthestanda¡dTYK.connectionsinpipe.However'Somespeciar t..i,niqueï'å"y"ilï;;i;" investigate portions of matched box fiK-connectlots'

Ultrasonic Examination

urtrasonic testing methods have been developed and used successfulry for many years in the

offshore ptattorå i"å"r,ry. rn. ,u*" tec'hniques îrïrr""ule onshore' conventional

techniques are applicable to dia."*;;;;;1.-t-:T:iå" ;;;i"rs ri. and thicker' and e

srearer than 30j. special tecr,nique, är.. reeuired^;J;;-ihese limits and should be

Ët"p"ttv',.tt"ã ano -Jvaluated prior to implementatron'

Designersthatspecifyc¡pglîe.welddetailsl::-:''tubularconnectionsarelikelytoberhe same on., ,hu, bu.r-rp".ify inspection requi"il""

';;"1 1:ï will requir e 700%

ultrasonic Tesring (ur) of each #äã; ó;rig"rly, there are..critical cases where the

higher level or inípection i, *urru,îäîuì.ã.p.ndtöö;;-iir^:ltl:îd experience or the

ur technician, this inspection Ä;ñ oftËn leadi tä disputes among the rechnrcrans'

ánriu.,nts and engineers/owners'

Mockups or sample connections with known defecrs should be preoared from tubular

connection, ,"'är!ir, ìn technici"rr'irîning anq -..:";"ilp1iãt tb nit inspection of the

orrduction *urt ' iurther, visual .ä"ïir*"iion or uT indications on productión work should

i,e require¿. rt ir'i, best achievea by forming an excavarion party consisting of a craftsman

inp.irn,*-'"., jgil:äåf;';H¡iå;f i"åUIgi,:l*.*:îl;::::::Ï'i::'á?;sometimes a tol

172

ravers of metal are progressivel,v removed ro revear the uT indication' As the predicted

inãication depth i; ä'p-p-;".n._L1],,ñembers presen*nãrii u" -giy:n ln opportuniry to

observe the progrer, þiio, * ,:,"::,:g t'r't n'*i.lavlr' ö;;;;¡ iidj-c'ations which exceed

the acceptun." ,runäår¿. ur. ,n"n ,.iåä îl "ir.í"r .oiriä"tiãn' weld repairs are then

made and those tå"tHË;;i; u'i examined'

pJp,s are seldom suitable for. ulrrasonic examinatio¡. pJp's,- like allÂws D1'1 welds'

requirel007o'""ìi"-"*''"ii""'s"'f 'r'!'M1s1?l';'i**ruì3¡;*itîI-ff-Í1!ïË

iiî"i*: r' " T*:'"#,;'J"J.:'" î:î"::'l:i ff{ilî F ä o'iï' i' i'ar cas e s' as wiitr h " ?Y

wall thicknesses, ;ä';n;; nnrv *i.î'ã"q*rn.á i"ri"irøn with tubular connectron

äfrltiã"c. can be obtained' :^ r ¿^

rnspection,,",1,1,,,îä"jff .Jffi 'ïä;i,îi'",',i:ìî*ï'XÏ=ä:*î'iilgiå"i::å

Occasionally, spot cn^TrÌtró.:i","'-^'the required size ancl posse"":"T-l -"^"*it;ri^,.'ç

determine ,nu, iï,."'rîuät'î.lo ir'ff'tr'. required ,:ä'^räo*p"iitËti,'inä p"o'riuili.v.t

:-"å5''*:¡:f rru:*"iÌ*ll"':1"¡;,'iiËJiä'í'Ji'ni^î"ñi'e'iäFo*hisreason, tt "

in'pJtiä' 'îo"r¿ tt'"tli¡i-ups prior to welding

Magnetic Particle Examination

Masnetic particle testing is usefur with 50tsi,T9_-_l:gh., vi-"19 "trngtl steels that may be

susãeptibre toderayed t;,oroe.n .ru.l'iil su.ctr testin'g rîJ;ld;r ao1ã 1

minimum of forty-

:îî1.tå"j:;xxlt;Ír,¡,':*¡¡i:ih#-Ï'#,nirîå:'":':Ëî:';;îJxil'r'l:iadversely ur".t"iuv yino "{.r:.ö"Ëor

this c^å, . *tti* background paint is applied

to the werd joint. síacr< magnetic puäi.r", in u.*ut"'',;.;;;tü"'-.-1iü,:tlZîtt;itü:;il;;;"in,i,ånî",ãi.;.;11::,åå.,,#tï,Jî11f, :iü:Ë,:r,Ëf ",f*lfJ:¡;ti;li:;;,;;"tpended parricles h-"]-:.åi"l

contrast and a smoother surface' T:::lit"t"?t;iiiitîlni,"' paint provides excepilor

we, out-of_posiïion and in drafts. *öi1

particres ur" ï"ï-ãinicult to apply overhead and in

drafts.

r u b u I a r c o n n e ct t o n s' t h e Lt ? Y l

d. :::-î 1Ti i: :i:Î"ït :läand messY and usuallY onlY

::i"li:::îïilH:î';iîJhHii:'Jä;:i'i;ff:Ë"äi'u'ion' or determining the extent

used

of a known crack'

SUMMARY

properly c!esigned and consrrucred, welded tubular connections provide efficient' economical'

and aesthetica'y plea-sing ,,nrutinn.. tå';r*iî;ffing- Há;rver"there are a variety of choices

to be made ny arci,itec-ts and .ngin;1" that hãve "':;tgñiãt impact on costs of the

,112

completed connections. The key points are:

1. Choose box sections over round sections for simple trusses or space frames for easeof fabrication.

2. Choose gapped connections instead of overlapping connections wherever possible forease of installation of members as well as welding and inspection accessibility.

3. Choose stepped over matched connections for aesthetic applications to reduce theamount of cosmetic grinding.

4. Choose fillet welded connections wherever possible as the least c'ostly to fabricate.Choose PJP groove welded connections over CJP groove welded connections for lowercosts in bevel preparation, fit-up, welder skill level requirements, and inspection.

5. Choose backing in CIP or PJP groove welded connections wherever practicable toreduce welder skill level requirements and minimize rejected welds.

6. Don't over-specify inspection requirements. Rely on visual inspection of joint fit-upsand completed welds.

7. For architectural and aesthetic applications, require a workmanship sample or mock-up connection from the fabricator and erector prior to production work to set thestandard for visual acceptance.

REFERENCES

1. Post, J. W. 1989. Gaining confidence with the fabrication, welding, and inspection oftubular connections. Proc. AISC National Steel Construction Conference.

2. Post, J. W. 1990. Box-tube connections; choices of joint details and their influence oncosts. Proc. AISC National Steel Construction Conference.

3. Structural Welding Code-Steel, ANSI/AWS D1.1-94.

4. Marshall, P. W. 1992. Design of welded tubular connections. basis and use of AWSCode provisions: Elsevier, Amsterdam, The Netherlands.

5. Packer, J. A. and Henderson, J. E. 1992. Design guide for hollow structural sectionconnections: CISC, Ontario, Canada.

6. Blodgett, O. W. 19óó. Design of welded structures, James F. Lincoln Arc WeldingFoundation, Cleveland, Ohio. 5.10-9 to 5.10-14.

1. Luyties, W. H. ancj Post.J. W. l9tìfì. Local dihedral angle equations fortubular jointsand related applications Welding Journal 67 (a): 51-60.

174

Bronch Member

Toe Zone

ur

SideMoin

ZoneMenrber

Circulor Connection

0 = member intersection ongle.I

V = loc.ol. dihedrot ongle. The ongle,meosured in o plone perpendiculoito the line of the weld, betweentongents to the outside surf ocesof the tubes beinq joined of theweld. The exterioi ¿ifreOrot ongle,where one looks of o locolized-section of the connection, suchthot the intersecting surf ocesmoy be treoted os plones.

Bronch Member

Heel Zone

ïoe

!,Zone

Side ZoneMoin Member

H

nFffi:Bo

Fig u re

Box Connection

Tubulor Connection Nomencloture

Overlo pped

Motched

Stepped

!ot I

rt 4x4x1 /8"

4X4X1/2"

4X4X1 /2"

4X4X1/8"

Fig u re

l.-Connection

2. Con tin uous

Plug Style Box Ring

.-^ po]tgr.l

BurnoutsI rorn 3/4', or l,; ïiotu

,'Þ

bqcking f or box

Y-Connection

8:ffiï"íî;flil;Bevelino

Attochmént

tube applicotions

Bios-'cut pú;///,///

TY! ÑWffiM.*rb*

i

ffiffi

a.

Figure 4. a.,¡ Fadgue cracks initiaring aÎ "smooth" 19e

of weld at end of gusselt'

b.) Fari_2e crack at end åigort." added to "strengthen" a crane boom'

Figure 3. Tubular connections with unnecessary knife-edge gussets'

b.

177

ú.¿

:

Figure 5.

Figure 6. NC machine wirh plasma cutting torch and examples of simple (gapped) and

compound (overlapped) cope and bevel preparations'

Various tvpes of backingmake them "continuous"

that are easily fabricated and require no welding to

in compliance with the Code.

178

DESIGNoFHALF.THROUGHoR''PoNY''TRUSSBRIDGESusING SQUARE OR RECTA¡{dULAR rrollow srnucrunar' sEcrroNs'

S' J' Herth' P'E''

ABSTRACT

The initial part of this paper will outline some of the research, testing, and investigation which has

been done on harf-thróugn t * uridges. Tï,is research is p*natity concerned wittr two items:

1. Design of the top chord of the tn¡ss considering out-of-plane buckling problems'

2. Design strength and stiftess-requirements for the "u-frame" formed by the tr¡ss web

members a¡d the bridge floorbeams'

The second part of the paper w'r-outrine continentar Bridges' design approach to "pony" trusses'

Referencing the above-mãntione¿ r","-'t' i*dings' t't" p;;; oottttt õontinental's approach to

detemrining upp-pi"L ii-ioro* fo, ¿o'p oftn"îop "no'ã'ut

*ell as süength and stiftress design

of the "IJ-frame" members'

The finat part of the paper is a discussion of some of the connection design ra¡rrifications of a half-

through *, ,*"ã,,ã labricated with square or rectang'rar hollowitructural secúons' The

connection primarily discussedh.J' *iff Uå tU" one betweto '¡" tn¡ss web members and the floor

beams. This connåction has design 't'*$h re'uir"l"n¡ for bending moments due to lateral

supportr"qoir"-"oä"iiut tp chõrd in a "pony" truss bridge'

RESEARCH AND FINDINGS

The out-of-plane buckling probreq of the compression chord of a "pony" truss can be equated to a

corr¡mn supported by elastic restraints ";ür"-;;, paner points Theiaterat support for the top chord

is provided by the **Ã"*.rr. t'*'äJ"'lU-to-t';1não'bea¡ns and tn¡ss verticats)' The 'U-

fraÍres,, must be adeq'ately designed iorlotu strength *ã Jno.ss to provide the lateral support

needed for toP 'chord stabilitY'

lReprintedfromSPATIAL,LATflcEandTENSIONSTRUCTuRESProceedings,IASS-ASCE

lnternationalSymposiumTgg|,He|d"*'¡*"'*withtheASCEStn¡cturesCongressXII,Apnl 24-28, I 99¿, Atlanta GA'

2 chief Bridge Engineer, Continental Bridge, 8301 State Hwy 29N, Alexarrdria' MN 56308.

179

lI

i

iem of tue top chord buckling of a half tb¡ough tn¡ss was brought to light in the- late I 800's

eries of ,,pony" tn¡ss faih¡.r. Th" first succèssfi.rl attempt to explain these failures and to-i;.,h"á

ofLdysis was done by Engesser @ef. 1, Ref. 2). Since that time a number of

,Jve investigated "pony" tnrss bridges'

lhe most extensive resea¡ch and æsting of "pony" tnrss bridges wzs done by Edward C' Holt

Ë';, ú1,ã.r. o at rhe pennsytvania state colege. with sponsorship from the coh¡mn

.,Council of Eneineering Foundadon andthe Pennsytvania State Highn"ay Deparment' Holt

h-r"d;il;r"¿fll scãe testing on "pony" tnrss bridges and wroæ a series of fol[ ¡eports

"ä;; ;* ¡"J*p"" (Ref. O gives recornmendations for design ofbridge chords with'out

acine.t-I

the DeBor:rgh Manufacturing comFanY, a manufacturer of pedesrian 'l!oly" tn:ss bridges

tily n¡bula¡ constn¡ction conducted strain gage tests on a full scale (80' long x 10' wide)

Ër;il;;;; fr"- square and rectang'lar rubi"9 ßef- Ð. Their findings indicate

more stringent require,ments for t'bular "poiy" tnrss bridges than'*'ere dictated by the Holt

I

I

kling of the top chord of the "Pony" truss has two limiting bounds:

I

./ tn" uter¿ restraint provided by the "IJ-fra¡nes" is very flexible. the chord will tend to

-uckle in one sinele half wave over its entire lengfh'

Ëä;ä;;;;;;;"vided are infinitety stifr, the chord will tend to buckle between the

nrss panel points.

ì

he of these exte'es is seldom if ever reached in practice as either would be uneconomical.

tal buckled shape used in design is somewhere between these nl'o exEemes: some nr¡mber

' la,res less than the total number of bays in the truss'I

I "u-FRAME" srrFFllESS REQUIREMENTSI

,,proach will be utilized here for the determination of top chord K-factors used in design'

[.."t ¿"".-i"es the K-factor for out-of-plane buckling of the top chord based on the

, Ë;;;iîî.:'u-frames,,. Holt's sotution for the allowable buckling load of the top

'ã "porry" tnrss assumes the following conditions:

i

The tra¡rsverse frames (u-frames) at all panel points have identical stiftess'

l

[e radii-of-gyration of all top chord members and end posts are identical'

he top-chord members a¡e all designed for the same allowable unit stress (A's and I's are

"loportional to the compressive forces)'

þe connections between the top chord and the end posts ale assumed pinned'

180

5. The end posts act as cantilever springs supporting the ends of the top chord.

6. The bridge carries a r:niformly distributed load.

The results of Holt's investigation are presented in Table 1, which gives the reciprocal of the

effeçlivç-!_e¡gth factor K aåa fimction of n (the number of panels) and of Ql/Pc where:

C

Ll

Pc

is the funished stiftess at the top of the least stiffnansvrol**". (See Figure 1)

is the panel point spacing of the tnrss

is the mærimum design chord stress multiplied by the desired factor of safety.

Note: Because of uncertainties involved in the analysis of the top chord of a "pony" trtrss, it isreasonable to require a factor of safety for overall top chord buckling greater than that used when

designing typicalcolumns; However, since each member in the continuous top chord of a "pony"

truss with parallel chords çannsf be simultaneously stessed to its critical buckling load" it isreasonable to use a safety factor of i.5 for this situation.

Various secondary effects on top chord buckling such as the lateral support given to the chord by

the diagonals, eflects of floor beam deflections due to live loads, etc. have been studied by Holt and

others. A full discussion of all aspects influencing the top chord stability of a "pony" truss bridge

is prohibited here by the tength limit of this paper. I recommend obtaining the "Guide to Stability

Dèsig¡ Criteria for Metal Stn¡ctures" (Ref. S). Much of the information on "pony" tnrss design

presented here is contained in Chapter 15 of that reference. Table 1 and Holt's assumptions are

reprinted from that source with the permission of John Wiley and Sons,Inc.

''U-FRAME'' STRENGTH REQTJIREMENTS

Strengfh requirements for the "LI-frame" members vary from source to source (research findings,

design codes, etc.). Most approaches require an additional moment capacity in the tnrss verticals,

floor beams and their connections. This moment is over and above the moment determined by

classical analysis and is calculated assurning the vertical is a ca¡rtilever, fixed at its base, which

carries a transverse force at its upper end. It is the opinion of this author that the most rational

"pony" tnrss design approach equates the required out-ofplane bending strength of the "IJ-frame"

to tUå top chord compression and to the K used for top chord design. (If K out-of-plane equals ttre

number ofbays, the chord would be designed to buckle in one long half wave. In this case, no out-

of-plane bending stengfh would be required in the "tJ-frames" for lateral support of the top chord).

The strength requirements suggested by Holt (Ref. 6) are:

l. The end post is a cantilever which carries, in addition to its æial load, a transverse force of0.3 percent (.003) of its æcial load at iæ upper end a¡rd

181

TABLE 1 - I/I( FOR VARIOUS VALI.JES OF CWCAITTD n

UK1.0000.9800.9600.9500.9400.9200.9000.8500.8000.7500.7000.6s00.6000.5500.5000.4500.4000.3500.300

4

3.686

3.352

2.96t

2.448

2.035

t.750

1.232

0.121

6

3.6t63.2843.000

2.7542.6432.5932.4602.3132.1471.955

1.7391.6391.517

13621.1580.8860.5300.187

I

3.6602.942.6652.595

10

3.7r42.8062.542

2.3032.t462.0451.7941.6291.501

1.359t.2361.133t.0070.8470.7t40.5550.3520.t70

t2

3.7542.7872.456

2.2522.0941.951

t.7091.480

1.344t.2001.087

0.9850.8600.7s00.6240.4540.3230.203

t4

3.7852.771

2.454

2.2542.t011.968

1.681

1.4s6t.2731.111

0.9880.878

0.7680.668

0.5370.4280.2920.183

16

3.8092.7742.479

2.2822.tzt1.981

t.694t.465t.2621.0880.9400.808

0.7080.6000.5000.3830.2800.187

2.2632.0131.889

1.7501.595

t.421.338t.2lrt.0470.8290.6270.4340.249

{

c= Eh2 [h/3I" + b/2r6]

FIGURE I-PONY TRUSS ''IJ-FRAME''

182

2. The moment at the lower end of each vertical may be approximated satisfactorily by

applyng atansverse force at its upper end equal to 0.2 percent (.002) of the average of the

ærial loads in the two adjacent top chord members'

While never going less than Holt's suggested requirements, Continental Bridge has adopteg 9:foltowing gurã" ünes based on the more conservative "German Buckling Specifications" @IM4lI4) which are now out of Print:

1. For the interior "IJ-frames" use l/100K times the average compressive force in the two

adjacent top chord members as the force applied at the top ofthe tn:ss verticals. NOTE: We

have chosen to limit K for uniformly loaded pony truss bridges of nrbular construction to a

mædmum value of 2.5. This gives a minimum out-of-plane force of 0.004 (l/100K) times

the top chord compressive force. This minimr¡m is in close agreement with the 1989 strain

gage testing of tubular "pony" tnrss bridges done by DeBourgh Manufacnuing (Ref. 7)

which for¡nd for the stnrcû¡¡e tested that an average of 0.0027 times the top chord axial load

was transmitted as a lateral load to the center vertical member.

2. For end frames, the same appiies except K is omitted (0.01 agrees with the recommendations

of the "Guide to Stability Design Criteria for Metal Structues").

DESIGN APPROACH

The economical design of a "pony" truss bridge using hollow structural sections is an iterative

process. There exists an almost infinite nr:¡nber of solutions to the design of the top chord and its

iateral bracing system (J-frames). The best top chord tubular section for a "pony" truss is

rectangular with a wide horizonal face. This section has a good radius-of-g¡nation for out-of-plane

buckting. Directly opposed to this in regards to economics wiil be the requirements of this face for

"ooo.rtioo strengfh ã"rigp (simpte tubular connections are more economical when the chord face

is na¡row and thich ha.dng a low width to thickness ratio). While the most economical design for

large heavily loaded stnicnues may be to size the truss members for srength and stifÊress

,"qrrir"¡¡"ot , then design connections as required, most stn¡ctures least cost altemative will be

¿etermine¿ by considering steel cost verses the cost of the tubular connections-

Following is the design approach adopted by Conùnental Bridge for uniformly loaded simple span

bridges t¡iføi"g simfle *"1¿"¿ tubular truss connections (tubular members a¡e miter cut and welded

aireãtly ro the ø"" ãf tn. framed to member). These bridges will have their floor beams welded

directly to the truss verticals (See Figure 1).

l. Detemrine the tn¡ss configuration required based on span, deflection limits, aesthetic

considerations, etc.

2. Analyzethe bridge structure for all applied loads'

183

J.

4.

5.

Using a K factor of approximately 1.5 for out-of-plane buckling (1.3 to 2.0 is typically an

oooð-i. range for tubula¡ stuctures) and 1.0 for in-plane buckling, detennine a tr¡be size

required for the top chord based on the design loads'

Design the tnrss web members and floor beasrs for thei¡ design loads, including the ow-of-

planã bending moment required for top chord stability. Keep in mind that the vertical's'dimension

perpendicular tó the chord face, must be equal to or less tban the u/idth of the

chord face.

Calculate the spring constant (C) firnished by the "IJ-frame" having the least ffinsverse

stirr',ess (See figrrre 1). L/trl.t'l çICalculate the value ClÆc. t h\

Enær table I with n (the nr:mber of bays in the truss) and CIÆc and find the correct lff valve /for a comFression-chordpanel, interpolating ¿u; necessary L !..

\/

Determine the actual K value and: \

- If the calculated K is less than the K value initially assuned, check the "U-frame" for

the new out-of-plane bending moments based on the lower K value; however, it may

be possible to ieduce the size or thickness of the top chord based on a lower I(Ur

value.

(or)

- If K calculated is greater tha¡r the K initiatly assu¡ned in sizing the top chord you

mr¡st either:

a. Check the top chord for a higher KVr value and if necessary, increase its si2e,

b. Increase the stiftress of the "IJ-frame" members to achieve a lower K value'

or

c. Some combination of a and b above'

g. Check tubular connections as outlined in the next portion of this paper-

10. Iterate steps 4 through 9 to final solution'

Bear in mind that while the "pony" truss considerations and the connection design criteria are kept

sepamte here for si-fu"ity, ti. ".ono*ic design of a "pony" tn¡ss fabricated from tubula¡ members

will consider both ,,Û-frame" requirements and tubular connection efficiencies simultaneously'

7.

8.

184

CON¡IECTION DESIGN

As stated above, the economical design oftubula¡ strucflrres is highly dependent upon connectiondesign. The most cost effective design is usually some middle ground between the least weightalte¡native a¡rd the least fabrication cost alternative.

Ifyou are doing tubular connection design, I would highty recom:nend obtaining the "Design Guidefor Hollow Stn¡ctr¡ral Section Connections" (Ref. 9) published by the Ca¡radia¡r Institute of SteelConstuction. This g¡ide is a¡r excellent source of curent design infonnation on hollow stn¡ct¡ralsection connections. Portions of this guide are reprinted here with permission.

The connections of primary importance in a tubular "pony" truss are:

l. The main load carrying (vertical) tn¡ss connections at each nodal joint where the truss webmembers attach to the chord members.

2. The joints between the floor beams a¡d ttre tnrss verticals.

The design approach for tn¡ss nodal joints is well documented in the above-referenced design guide.In the United States, the same design approach found in this gurde has al5e been adopted by theAnerican Welding Society (Ref. 10)- Either of these sources may be used in checking tuss jointcapacity.

While a full discussion oftubularjoint design is limited here by the length of this paper, I wor¡ld liketo make the following poinæ:

1. The vertical members in a tubular Pratt type "pony" truss, becar¡se of economics and "IJ-frame" considerations, are typically very nearly or are the same \ñ¡idth as the chord members.

2. The design capacities which have been developed based on full scale testing oftubularjointshave a somewhat limited "range of validity".

Based on these two points, I have found that once "Ll-fizme" requirements and validity limits aremet the actr¡al mein ûusis connection resistance provided is in many instances greater than thatrequired for actual member loads; therefore, during the iterative design process, you typically needonly consider connection parameters, staying within the appropriate "range of validity" for theconnection you intend to use. You can then make final connection capacify checks after all membershave been selected. NOTE: If staying outside the "range of validþ" established for tubula¡connections, the designer is on his own. While connections outside the validþ range obviouslyhave some capacity, I do not recommend their use. If using cormections outside the appropriate"range of validity", the designer needs a very good understanding of the possible faih¡re modes in

185

8.1]I

a tubular connection (i.e. punching shear, chord shear in gap joints, chord face plastification, etc.)

and how these factors influence connection capacity'

The second connection of importance, which is primarily controlled by "U-frame" considerations,

is the one be¡¡reen the tn¡ss verticals and the floor beams. Along with the end shear reaction of the

floor beam, this connection must be capable of resisting the out-of-plane bending moment induced

in the tn¡ss verticals (See previors discr.rssion on shength requirements of the "U-frame"). NOTE:

Secondary stresses due tó floor ber- deflections are typically quite small in a uniformiy loaded

bridge and in most cases can be neglected.

Simple n¡bular connections have a certain amotmt of flexibility due to deformation of the tube face-

ln a "pony,, tnlss, the floor beam to vertical connection is assumed to be rigid in order to provide

hterj *ppott to the top chord. Because of these facts, p (the width ratio be¡r'een the floor be'm

and vertiõal) should be approximately equal to one for this connection'

After 5izìng the ,U-ûame" members and detemrining design loads, the connection must be checked

for its ,"qoir"a capasity: Tpical tubular floor beam members are deep narrow sections (TS 8x3's,

TS lOú';, eæ.) with aielatively high bending sængth about their stong axis. These efficient beam

sections are r:sually outside the "range of valid.ity" cr:rrently established forplain T-type connections

with in-plane benåing moments (See "Design Guide for Hollow Stn¡cn:ral Section Connections",

Chaptei6 (Ref. 9). It is still usually more cost effective to use these efficient beam sections and

design appropriate connections for their r¡se'

In designing tube-to-tube floor beam connections which are outside the established "range ofvalidity; for T-type hrbula¡ moment connections, one may conservativeiy treat the floor beam zìs you

would a wide flange beam framing into a nrbular colurrn. The vertical faces (webs) of the tube are

assumed to carry the shea¡ load in the floor beam to the tn¡ss vertical tbrough the side w'elds- The

end moment in the floor bea¡n (out-of-plane bending moment in the tn¡ss verticals), as in the case

of a w-shape bearn, can be resolved into two equal and opposite flange forces- These forces a¡e

applied at the top and bottom horizontal tube faces of the floor beam. The top and bottom tube faces

can then be equated to a plate welded transversely to a hollow stn¡crural section. The "flange"

capacities of the tubular floor beam (or w-shaped floor beam) can then be checked using existing

aesign rules for transverse plates welded to hollow stnrctr:ral sections (See Table 2 copied from the

"DeJign Guide for Hollow Stnrctural Section Connections" (Ref' 9))'

Weld design for both main truss joints and floor beam connections shall be P..,

th. applicable de¡ifr

code. Bear in mind that in tubular connections such as these, tra¡rsfer of load across the weld is

highty non-r¡niform. Welds must be large enough to enable adequate load redistribution to take

ptã."'*itt i' the joint, preventing a progreisive failure of the weld and insuring ductile behavior of

the joint.

186

FACTORED CONNICTION RTSISTANCTCONNTCTION TYPI

ß = 1.0 Bosis: CHORD SIDE WALL FAILURETronsverse Plote

b1-l r

H_

where B

I r'r,NI= Fyo to (21 , + loto)

0.85SDlr - t/y Bosis: PUNCHINGSHEAR

Nî= fr& Czt, + 2b"p)

ALL Þ Bosis: EFFECTIVE WIDTH

FU N CTION S

N i coNNECTIoN RESISTANCE,Fy o SPECIFIED MINIMUM YIELD

f V, SPECIFIED MINIMUM YIELD

AS AN AXIAL FORCESTRENGTH OF TUBE

STRENGTH OF PLATE

= bo

2lo

, tor, br but ( b'

bo/lo '

b- : 10,, lYo lo p., but ( b've bo/to Fy' t1

RANGI OF VALIDITY: bo/to ( 30

TABLE 2

FACTORED RESISTANCI OF PLATE TO RECTANGULAR HHS CONNECTIONS

llLvrr sTATES oR uLTIMATE LoAD FoRMAT)

187

i{ïIi

REFERENCES

1.

,)

).

4.

Holt, E.c. 1956. The Analysis aftS Ð-e$gn ot o"'å^" -'='T-- Rep' No' 3'

of Bridge chords *rinïffiø Bracing' column Res' ço

American Wetding SocietY 1994'

ChaPter 10'

Engesser,F' 1893'

Vol.II, Berlin'

Hott"E-C. 1951'Hott"E.C.-1951' 'nc'ReP'No' 1'

äË¡*¿* ao** without Lateral Bracu

il:. ;;Buskr i'f "r q"ry r#-# Xs:'

Stablitv or Brideç chords without

HoIL Þ.u- tr,"'"---i-- I.s. Cor¡nc. RepNo.Lateral Bracing, Col

StabilitY

StabititY

Stability

4th ED.,

5.

6.

7.

8.

9.

10.

i

I

truc¡sa]jgeldig

188

CASE STT]DIES OF RECENT TUBULAR STRUCTURES

C.M. AIIen*

ABSTRACT

Tb¡ee quite different projects are presented, in which hollow stuctural steel tubes are used as the

principie structural framing. The National Aviation Museum of Canada featr.¡res an all welded

space-frame roof and exterior wall stn¡cture comprised of circula¡ steel tubes. The Toronto

SgOotn" is a retactable roof stadium in which the roof structue is comprised of square steel tube

..i t *r.r with a combination of welded and field bolted corurections. The ttrird case study is a

series of steel square tube tn¡ss access towers used in the constrr¡ction of the Hibemia oil platform,

off the east coast of Newfoundland, Canada" Each project presented to the design tearn unique

challenges in the design of steel tube structues, providing lessons for its'futr¡re use and illustrating

certain areas where additional research could be beneficial leading to improvement in cr:¡rent

standards and design practices for steel tube sfi¡ctu¡es.

CASE STT]DY 1

NATIONAL AVIATION MUSEI.iM OF CANÄDA

Building Description

The National Aviation Museum was deveioped by Public Worla C-.anaÅa to store and display

Canada's aeronautical collection representing Canada's involvement in aviation and qpace

technology in the 20th century. The museum, located at Rockliffe Aþort in Onawa was

completed n lg}7. The a¡chitectural fooprint of the aircraft display hall is tiangular shaped to

suit the orientation of the north-south taciway and tl¡e east-west n¡nway. The single storey

triangular buildi''g is divided into nvo e.qual right angle triærgles by means of an exp"ttsion joint'

Structural Framing

The ñ¡nctional and architectr.ual considerations, with the requirement for a wide oPen space suitable

for the display of large aircraft combined with the desire for a light weight yet economical exposed

roof structure, dictated the stn¡ctural planning for tbe museum-

The building fooprint is an isosceles right angle triangle with a short side of l6lm in lengfh and a

clear height of 13.2m from the floor to the underside of the roof s¡n¡cture. It is divided into two

eq¡al smaller tiangles by means of an expansion joint located at right angles to the hypotenuse ofthe larger panel, as shown in Figure 1.

The stn¡cnral framing resulted from considerations of function, architectu¡al expression, lightness

in appearance and economics. The selected stn¡ctural system \¡vas a space frame with circular steel

tube members and all welded joints.

* Adjeleian Allen Rubeli Limited, 75 Albert Steet, Suite 1005, Ottaw4 Ontario, Canada KIP 5E7.

189

NORTHWING

x

Figure 1 AviationMuserm-Keyplan

The roof framing is comprised of a double layer off.set gnd in which each grid is directionallysimilar with the lower chords located below *â io benveen two upper chords and with the upperand lower nodes connected by diagonal members (Figure 2). The grid spacing horizontally arrdvertically is 3'30m' The offset grid system was selected due to the tcre¿sea $iffiress and lateralstability it provided together with its overall pleasing appearance

Three interior columns for each wing ofthe museum, spaced at46.2m,provided an economic roofspan while permitting the movement oÏthe largest aircran in the co[åtion an¡avhere within themuseum' Each of the interior columns is shaped as an inverted pyramid, 9.9m higb with a pinnedbearing at the vertex of the pyramid" The contorned sbape oro" i"t ioi.;;,*", provides severaladvantages as follows:

o Acting as a shea¡ head' the inverted pyramidal column reduces the clear span of the space roofedd'o The load transfer from the roof to the column supports is smooth and more gadual" The configuration allows for a stable stn¡cture for lateral loads ûom *i"d

'íJ earthquakes.

The bearing at the vertex of the inverted pyramid tansmits vertical and horizontal loads to aconcrete pedestal in the shape of an upright pyramid 3.3m high. The overall config¡x;;;rhilvertex ofthe pyramid located ¿f rhis height provides the muúl¡m

"f*, ,pu.":* "

the wing levelof an Argrrs aircraft, the largest aeroplane inthe mr¡seum collection Gig¡¡re ¡)."-

GRAVITY COLUMNWIND COLUMN

¡

190

Figure 2 Aviation Mr¡ser¡m - Bottom chord Pla¡r ofNorth wing

Figure 3 Aviæion Mr¡ser¡m - Interior Column

In addition to the th¡ee interior columns, a nrunber of smaller columns are provided along the

perimeter of theuuil¿ing to support th9 roof edge and þ¡itrting cladding and to provide additionat

stability against rateral îoads.'i portion of thise are road uraring while the balance a¡e wind

columns *itU u sliding vertical connection at the roof (Figr¡e 4)'

Secondary Framing

The roof stn¡cture is a metal deck supported on steel T sections attached by verticat supports to the

top chord nodes and at the mid-points of the top chord members nrnning parallel to the principle

diagonal of buildi.g fooprint. The top chord members at rigbt angles to the principle diagonal are

not zubjected to secondary bending from roof pwlins'

BR^C'DI;nlvl ?

191

GIRT

GIRT

GIRi

GIRTI

Figure4 AviationMuserm-E:rteriorColumn

The overall stabillty of the framing, with its inærior pyramid columns and exterior tiangularcol 'mns, provided overall resistance to the lateral forces åu" t" wind and earrüq'ate. This primarysystem was augmented with a two brace finmes in each of the ¡ro comers of the base of thetriangle of the overall building fooprint, T ot¿.r to improve its torsional stiffiress for both windand eartlrquake induced forces, and thus reduce lateral deflections at the vertices of the ûiangle aswell as atthe expansionjoint

Loading

The overall dead load of the ,tn "t*l steel roof qpace frame, including members and joints butexcluding secondary framing and columns, was .sipa (ll lbvsq.ft.). s*perimposed dead loadstogether with the roof space fiame load resulæd in an allowa¡rce or i.e¡ lpa. The design snow loadwas l'73 kPa Due to the height of the building and light roof shrcture, lateral forces due to windgovenred so ttrat earthquake loads were not a considerafion. For most members, the design wascontolled by gavity loads.

Space Frame Members

Theroof space frame foreach ofthetwo smallertiangles is comprised of about 5000 memben and1300 nodes' The members are all circular steel tubesfoth a yielá shength of ¡só tñ".[; ;*;in size from l0lmm to l68mm. Column support members were also circula¡ hollow steel tubes but

I

t

¡

!

D

I¡

SIDE ELEVATION

192

with a mærimr¡m size range up to 324.mn. All steel tubes a¡e cold formed and stress relieved

(Class þ.

Space Frame Joints

There were a number of considerations which led to the selection of the eventual joint configuration

and connection method ¡"¡uding:

o Requirement specified by the owner @ublic V/orks Canada) for all-welded consur¡ction due to

the aesthetically superior appeamnce'

" õ*"**l p"rør*-rr r"q.rir"*"rrt that the joints be sEonger than the members framing into the

joint to ensure member faih:re prior to joint failure'o Custom design joint that experiencedlocal str¡ctr¡¡al steel companies could fabricate and erect

without relying on single sotuce prcÞrietary space frame suppliers.

The joint detail selected after careful consideration of many configurations is shown in Figrue 5'

Each chord member, consisting of a ror¡nd tube, is capped with a circular mild steel plate of

diameter equat to tt"t of the n¡Ui. Tiris cap plate i5 v¡slded to the tube using a square groove weld

with a cylindrical backing ring inserted iff; the end of the h¡be. A rectangular tongue plate is then

welded to the cap plate-witli a double V groove weld. The joint itself consists of a specially

fabricated star-sirapea plate. The tongu. !ut" of each of the chords meeting at the joint is

connected to the upper tutfu"" of the star plgte by apair of fillet welds'

PLÂÎÉ FOR OIAGONÂL

fuge coNNÉclloN

Lr.t:sJ"orrorl

."0.N

PLAfE FOR DIAGONAL

TUB€ CONHEC¡¡ON

S¡AR PIAlE

¡ON6UE PLAIE

SECTION ]

uil,¿tH{=-_

SECTION 2BOTTOM CHORD NODE

PLAN VIEW

Figure 5 Aviation Mr¡setrm - Joint Detail

The diagonal members are also capped with ror:nd metal Plates r.rsing square groove welds' The

tongue plates to be welded to the caps are, in this case, not rectangular but are specially shaped to

193

fit the horiæntal starplate of the joint The tongue plates ofthe forn chords meeting æ the jointform a cross. Each of these plates are welded to the star plate by a pair of fillet welds. In additionthey are connected together by a weld at theirjunction.

Testing was carried out on full size joints in a protot¡pe segment of the space tuss in which theoverall dimensions of the member lengths were scaled dor¡¡n to l/3 to allow testing of the tn¡ss intest rigs of apractical size (Ref.l).

The test program confirmed tbat the joint detail was adequaæ and that failure in a joint would notbe expected to precede faih¡re in a member.

Fahrlcation anq Erectior

The fabrication of the all-welded qpace frame was carried out in a series of steps at differentlocations, as follows:

o Fabrication of indiviú¡al tubes with end closne plates and tongue plates at the workshop of theprime steel fabricaûors

o Fabrication of larger-size tn¡ss elements with a length of 19.8m that cor¡ld be træsporæd bytruck Each tn¡ss element was tiangular in shape with one top chord and one bottom õnor¿ anãa temporary connecting angle replacing tlre other bouom chord.

" Assembly oftr¡ss elements on site into 13 large lifr zub.structures." Lifting of lifr zub'stn¡ctures with mobile cnrres, one by one, connecting substuctures together

by welding of connections to adjoining rub..süuctures already ..."æ¿ The subsmótrnesdirectly zupported by the interior coh¡mns were suspended higher than 1þsir firal positions whilethe coh¡mns were being erecte4 they were then lowered down and connected io tl" pyramidcoh¡mns.

Testing and fnspection

The original specifications called forthe visual inspection of 100%of welds, nondestn¡ctive testingfor all welds be¡veen the cap plates and tubes, and atl welds between a cap plate and a tongue plæãwhenever any of the two plates w¿ts over 30mm in thickness. Of all re,maining welds, 2iyn wererequired to be subjected to ultasonic testing. As the work progressed, thJweld çratity wasobsen¡ed to be uniform with a very low rejection rate. As a resuf the requir€,ments for tãstini wererevised, with testing frequency of the critical welds reduced from 100% io Z}%and the less criticalwelds from71%oto l}Yo. The welded joints of the interior coft.rmns and the exterior columns weretested using magnetic particle testing. Ulûasonic calipers were t¡sed to measure the thickness oftube members which could not be inspected by mechanical means due to the closure plates.

?II

I!

tI

I

it

It

t

t

t

t

t

I194

,ìI

i.t

General Comments

Although the overall aPpearance was in keeping with the original expectations, and the overall roofdead load related to the large sp"ns was stn¡ctr¡rally efficient, the requirement áf an all-welde.d jointwas an issue with regard to additional costs and time delays. The need for extensive weta testingand the practical considerations of winter constnrction can cause une4pected costs and increasedconstn¡ction time. Another factor is the accuracy required in the fabriåtion process to ensure themininizing of the internal stess effects of force fitting of the va¡ious elements or Iifr substn¡ctr¡¡esduing thei¡ assembling and connecfing

CASE STUDY 2TIIE TORONTO SI(YDOME RETRACTABLE ROOF

Project Description

Tlie Toronto SþDome is the world's first major league multi-purpose stadium with a fullyretactable rigid steel roof (Ref. 2,3). -Ihe SþDome converts Êorrra j¡,ooo

seat football stad.iumto a 51,000 baseball stadium by means of a rotating lower seating stand system. The principlefeatr:re of SþDome is the roof stn¡cture which can open or close creating an open air stadium forgood weather conditions and a closed roof dome stadium for bad weathL cond.itions and d'ringwinter.

Roof Description

The overall roof shape is dome-like-in appearance, approximately circular in its base plan, coveringa stadium which is essentially circular in plan. The roof consists of for¡r separate panels numberedconsecutively I to 4 from south to north with the roof in the closed position (Figr¡e 6). In its base

Fþre 6 Toronto SþDome - Roof plan - closed

plan, the panels a¡e delineated by dividing acircle into four parts with three parallel lines atthe mid point and two quarter points. The twomiddle panels a¡e in the fomr of barrel var¡ltswhile the two panels at each end are in theform of quarter domes.

Panel 4 is a fixed roof panel, located at thenorttr end of the stadium and is the lowestpanel in the sequence of nested panels in theopen position. This panel is shaped in theform of a quarter dome with a circular base inplan and an arch at the front or leading edge.The panel is supported on the concrete sub-structure by mears of sliding bearings.

II @1 I

tI

ttI @1 t

II

\p_----t

195

ffi--'!í!;;.¡,''

ìi,I

J

rI

l.

panel 1 is a quarter dome located at the south end of the stadiuq in the closed position This panel

is similar in sbape to Panel 4, but is larger in size with its base located at a higher slsvatie¡ than

Panel 4 to allow it to nest over Panel4 in the open position. Panel I is zupported on s'teel boges(t1cks) constn¡cted with steel wheels uihich intum are supported on a circular steel tack system.

Panel 1 moves on this circular táck system, rotating 180 degrees in its opening or closing cycle.

Panels 2 and3 a¡e each parabolic arch panel segments supported on the east and west sides of the

stadium with steel bogies containing steel uùeels on sets ofpæallel steel tacks runing in a north-

south direction. Both panels move in a north-sor¡th direction on these parallel tracls, In the open

position, Panels 2 and 3 nest ôver Panel 1. Panel 2 is larger in size than Panel 3 and iæ srpport

elevation is at a higher elevation in order to allow Panel 3 to nest below Panel 2.

The roof mechanism is operated by a computer progrrim and a reúmdant control syste,m ensrning a

safe and dependable operation. The roof opens or closes in 20 minutes in wind qpeeds of up to 65

lqr/hor¡r.

Roof Geomefry\

The geometry of the foru roof panels is complex. Each of the four panels has cr¡rrafure in ¡vodirectionso each are diferent in size, and each arch component in each panel is diffe'rent except for

symmetical aspects about the longitudinal æris.

The geometric complexity was resolved by developing simple mathematical expressions which

defines the cr¡n¡atr¡re in two directions (Ref. a) Gigr¡re Ð. Tü/ith this mathe'matical model in placç,

all of the roof geometry, could be automatically generated for use in the static and dpamicanalysis, CAD drafting and model studies, and forreleaseto the steel fabricator.

Roof Framing - General Conditions

The four roof panels are constructed of stn¡ctural steel arch trusses comprised of hollow structr¡ral

steel tubes. The núes are, for the most par! squa.re, varying in size from 254mm square to 304mm

squre for chord members and 202mn square typically for verticals and diagonals. In isolated

portions of the roof, rectangular tubes and plated tr¡bes were used. All steel tubes are Class H(56rtr relieved) with a yield stength of 350 MPa With the exception of the two leading arches ofPanel 1, all arch tn¡sses have a consistcnt cetrtre to cente of chords tnrss depth of 4.2m.

The roof arch tn¡sses a¡e connected to the boges or other supPorts by means of pin connections.

The pin connections allow for distortions in the roof geometry dùe to thermal effects and

differential movements of the steel roof and supporting concrete structt¡re withor¡t generating

significant membçr forces within the roof str¡cture.

t196

-BL2=-3.¿?3110 x

qi6:È - P^r.tL tÀv

sPRtñ6 P0tNiELEV - P^NEL 2

Figure 7 Toronto SþDome - North-South Secúon

A key aspect of the design concept for the roof is ensuring strucrural integnty should single

elements fail. The test for strucûral integnty u/as to check the stucnre for stability with allstnrctural members removed within a vertical cylinder of 4.5m radirx with the centre of this

"cylinder" located on any one panel point including a support. The design check for integrity was

based on one half the 1/100 year design live loads with the live load factor reduced from 1.5 to 1.1

and the dead ioad factor reduced from I .25 to 1.05.

All steel tube framing members were cleaned to SPIO followed by a prime coat of inorganic zincpaint.

Roof Panels 2 and 3

Both Panels 2 and 3 consist of eight parabolic arch trusses spaced at 7.0m excePt for a 5.0m spacing

of the first two arches of the south end of Panel 3, dictated by snow dtifting conditions. The arch

tnrsses consist of double tube chords with single tube verticals and diagonals using conventional

double tube chord tt¡ss technology. These arch tn¡sses are interconnected by transverse tru.sses

consisting of single tube chords, verticals and diagonals. The transverse tnrsses support standard

wide flange purlins which in turn support a 75mm deep acoustic steel deck. The diagonals of the

¡.ansverse tn¡sses a¡e oriented in altemate directions from tnrss to b:t¡ss so that they cornect to the

main a¡ch tn¡ss chords at joints which do not have connections of the diagonals of the main arch

üïsses. This technique effectively minimized congestion of members framing into any one arch

truss joint. Top and bottom chord bracing, consisting of single tubes, completes the framing ofthese panels.

197

i

JRoof Panels I and 4

lEach q'arter dome is framed with foru arch tn¡sses at the reading edge of each panel spaced at

l;;;;il;ì;ã:õ; The a¡ch trusses supporr a series of¡b trusses, radiating in a circular pattern

ûom the circular base in a direction to**ás the centre point of the circular Ot T:-lî::r,::äri ff äii äl*.u by circutar t"*t no.irontat iro¡ections from the north-south centerline

geometry. These noãota projections in tr¡m establish the geometry of the leading arch tnrsses.

A circular arrangement of transr¡erse tn$ses support the roof prulins' ao! snan between the rib

trusses. Top and bottom chord diagonal bracin! of the rib tr¡sses compretes the quarter dome

fr"-i"g. ttre steel deck and roofing ãetails a¡e similar to Panels 2and3'

A plan view of the roof framing is shown in Figure 8'

Figure 8 Toronto SþDome - Roof Framing Plan

Roof Loading Conditions

The roof panels were arnlyzedfor the following load condiúons:

- Dead load

- Snow loads as determined by wind tunnel tests by R.W.D-I- of Guelph, onta¡io @ef' 5)'

- v/ind loads as determined from wind tunnel tests performed by RW'D'I'

- Seismic effects based on an earthquake level of 8% of gravity

198

I

I

- Dynamic effects with a sudden application of brakes

- Loads imposed bY thermal effects

- Loads imposed by deformation of the concrete supports or rail location tolerances

- fsarls imposed on Panels 2 and3 due to skew effects under motion

- User loads suspended ûom designated points

A I in 100 year return period is provided for in the design of all live loads. The design is based on

limit sates design with an importance factor of 1.15 applied to all live load effects

Joint Details

A combination of shop welding and field bolting is r¡sed for all connections of the roof stn¡cture.

Truss secrions of approximately 15m in length were fabricated in the shop with welded

connections, primarüy fiUet welds, and with stiffener plates where required. After delivery by truck

to the site, the truss sections were assembled by bolted connections into tn¡ss assemblies of one or

two segments in \¡¡idth and n¡¡o or tb¡ee truss segments in length These truss assemblies were then

hoisted into the air and connected to previously erecæd assemblies by means of a bolted

connection, with 4325 galvanized joints.

Two basic types of bolt connection details were used as follows:

' bolted tube end cap plates wittr bolts in tensiono slip critical con¡ections with end tab plates connected in double shear by bols.

Testing of Roof Joints and Steel Tubes

A progra:n of testing of samples of the different types of roof tn¡ss joints, constructed at l/2 scale,

was ca¡ried out at the University of Toronto (Ref. 6). The testing included dynamic testing of the

joins as well as static tesß to failu¡e.

The dynamic testing included 5,000 cycles of low load levels, followed by 200 cycles of higher

load, follo'*ed again by 5,000 cycles of lower load. After dynamic testing, each sample was

inspecred for fatigue cracks using a dry magnetic particle technique. No evidence of fatigue

cracking was found.

Steel tubes for tbe roof tnsses are manufactu¡ed by cold forming and welding of the longinrdinal

joint Lack of fi¡sion problems along the joint led to a testing program at the University of Toronto

io evaluate the effects on the compression capacity of long columns with different degrees of lack

of fusion (Ref. 6). ln addition, tests on the compression capacity of tube columns, plated with steel

plates with lowei yields tban the tubes, were carried out 1o evah¡ate the effect of the two material

t¡'*gthr and the effect of the build up of intemal stress due to the welding process for tubes which

are originally stress relieved (Class þ. Steel plating of certain tubes was required in order to

.o.p.o*t. for steel tubes for Panels I and 4 being supplied to the project an average of 7.8% less

in average walt thickness (and mass) than specified.

199

Í

i

Erection of SlryDome Roof

The nesting geomftv of the roof panels-inthe open position rilas utilized to facilitate the rooferection by using lower roof panels as_shoring pËrorrÀ for subsequent erecte¿ qpper panels.Paneldbeingthelowestpanelwasthefirsttou".o*to"æ¿ Tbreetemporarytowersinlinewiththe leading edge of the panel and locat{ at the þlf point and the two quarter points, providedtemporary support for the two leading a¡ch tt¡sses. Panel I was construc,ûJ dir"rrly over panel 4i¡ ¿ 5imil¿¡ fashion with the extension of the temporay towers though panel 4 to support theIeading edge of Panel l. Panels 3 and 2 wqethenêrecte¿ nqpectively-*ia tu"

'se of æmporarysqpports offPanel l. As each series of arch trusses for Panels 3 and 2 were completed they wererolled north on their boge system to allowthe t*sJ;àruon tr¡sses to be eæcted.

General Comments

stuctural steel trúes were selected for the sþDome roof due to their superior efrciency inst¡pporting the large compression loads ofthe uoú tnor"r ortn i*irt"r.n*Lo tn, overall clqanappea¡ance

A nr¡mber of issues became apparent in the design and constrt¡ction of the SþDome roof whichcould have an effect on futr¡re hollow steel tube dJvelopment and use and are presented as follows:o As a direct result of the experience at sþDome and other projects, the canadian code on steelDesign and construction (cAlI3-s16.1-M) (Ref. Ð, now requires the tube weighrs to be wirhin -3'5To or +l0o/o of the published values. other jurisdictions orll-p*Jt-,,ru* man'factr¡red with

a +l0Yo wall thickness tolerance.o As a'result of the experience at sþDome, it is recommended that any tubes, uåich a¡emaur¡factr¡red under a cold formed and automatic fr¡sed weld process, should be continuouslymonitored by ultasonic testing as part of the manufacturing process

CASE STUDY3IIIBER¡IA PRIII{ARY ACCESS SYSTEM

Project Descrintion

The Hibemia Project is currently under consEr¡ction æ Bull Arrn, Trinity Bay, Newfoundland"canada The project is comprised of concrete base str¡cture supporting a steel fra¡ned oil drillingplaform.

The Hibernia Gravity Base stn¡cture (GBS) is consür¡cted, for the most parL as a floating, mooredstructure in Trinity Bay' when completed" it witl be towed out to its final resting position in theAtlantic Ocean offthe coast ofNewfoundland-

a

a

t

¡

¡

:

i

¡I

I

t

t

t

200

In order to constn¡ct the GBS, a series of access str¡ctures, called the Hibemia himary AccessSystem, a¡e used to proride a link ûom barges moored adjacent to the floating GBS to the interiorconcrete structure of the GBS. The Access System is used primarily for personnel access duringthe constn¡ction period and consists of a series of towers, bridges and mechanical liftingmechanisms.

During the moored. floæing phase of the constn¡ction ofthe GBS, the GBS progressively increasesin overall depth in the se4 as the height and mass of the concrete structure incteases. The primaryAccess System is a modular type of steel tube framed stnrcture, which also increases in height asthe construction of the GBS progresses.

The structu¡al design of the Primary Access System is unique due to a number of factors related tothe Neu¡for¡ndland offshore constnrction site locatior¡ the fiurctional requirements for theconstn¡ction of the GBS, constn¡ction staging requirements and specific design criteria set out bythe project contract requirements.

Primar:v Access System Description

Figue 9 provides pian aad elevation views of the final confignration of the Primary Access Systemdesþ. Preliminar)' versions of the design included towers fixed to the exterior concrete wall of theGBS, which required analysis for wave/current effects. Both the East and West Access Systemsindicated in Figure 9 are simila¡ in design, with variations resulting from differences in the supportdetails at the barge deck levels ar¡d the GBS concrete stn¡ctures.

Each of the East and \!'est Access Systems consist of eight components described as follows, insequence, from the outer service barges to the interior GBS sur¡cn-re:

' A 20m high tower fi-rçed to the service barge deck and bulkhead strt¡ctue (Towers T9, Tl0).o A gangrvay 8m long ünking Towers T9, T10 to the Main Bridge.' A Main Bridge supported at the perimeter GBS ice wall and the interior main tower assemblies.o A Main Inner Tower u'ith a mædmr:¡n heigbt of approximately 80m, tied back to in¡rer concrete

wall str¡cnres at inte¡¡als of 6.4m, (Towers T11, Tl2).o A sliding 'miniyoke' assembly, a steel fiame structu¡e which allows repositioning of the Main

Bridge at the Inner Tower support, and provides support for the Main Bridge at the Main innertower.

" A Support frame assembly at the base of each of the main towers which provides an accessplatform and a base support framework for the towers at the GBS concrete base sbuctu¡e.

The main inner towers Tl I a¡rd Tl2 are consbr¡cted using modutar units 6.4m in height, fieldbolted in place, as the concrete construction progrcss in height. Tie-baclcs at 6-4m intervals'between Towers TIl-Tl2 and inner concrete walls provide lateral suppor! although for someconsûn¡ctio¡ 5raSeS the upper tower units are free standing. The Main Bridge is supported on amoving bearing assemblrv at the perimeter GBS ice wall atlowing rotation and translation of thebridge support as the ice wall increases in height dr:ring construction.

201

tdiilt;!t

I

PIærView

1. Tower T9 -

2. TowerTl0î Tower Tl I-¿.

fower ttZ5. Garig}vay-6. Main eriaql?. Service Bnoge

S. GBSs MitriYolkio. suPPortFraoe

SectionView

Prirnas'Access System - Ptan ærd Section

Figure 9

zoz

The miniyoke is guided on a rail system fixed to the exte¡ior Tl l-T12 columns. A removable pinmechanism allows the miniyoke and in tr¡m the Main Bridge support at the Inner Tower, to bepositioned at increments of 6.4m along the exterior force of the to\¡/ers. As the exterior MainBridge support at the GBS ice wall is raised duing slipforming procedures, the interior MainBridge support is raised by meam of the miniyoke to minimize the horizontal slope of the MainBridge. Tlpe V/T was specified for all primar¡'load carrying members. Hollow stn¡ctural tr¡bularmembers varying in size from l50x150mm to 350x350mm were used typically throughout. Heavyrolled'WT sections were ræed for transfer girders, at the Tower TlI,Tl2 support frames. The totalmass ofthe entire Access System is approximately 900 tonnes.

Desisn Considerations

The detailed s¡rucnral design of the Primary Access System required consideration of numerouscombinations of design loading and geometry variables, which resulted in demanding computermodelling requirements. Additionally, careful assessment ofmember a¡rd con¡rection design detaiiswere required in order to optimize HSS connection design. The project conmct specifications,ñ¡nctional requirements and the assessment of va¡ious stn¡ctural configurations required designconsideration of up to 60 load cases, a¡rd 250 load combinations, for numerous stnrcn¡¡al modelswith up to approximately 1700 members ar¡d 900 joinæ. A fi¡rther compiication was therequirement to satisff the requirements of th¡ee different codes, the National Building Code ofCarørcr-' the CSA Offshore Structu¡es Code and tbe Project Specific Code for load values, factorsand combinations. These requirements required the development of in-house software programs tomaniFulate input and output data into formats which could readily be used for design puqposes, as

the demands of this project exceeded the capacity of vendor-pr¡rchased softrvare to fi.:nction withinpractical time requirements for design pìÌrposes.

Operational requirements had a major impact on design loads and conditions. The operationalrequirements included definition of live load for persorurel, equipmen! piping fluids, constructionelevator loads and the basis for the derivation of environmental loads due to \¡/ind" ice build up, andthermal effects. Dead load requirements were outlined for piping, elevator self weight, andconstuction related plaforms and supports. Other operational requirements included assessment oftilt effects of the CeS auring the constn¡ction phase on the Access System stnrcflres, bridgemovement effects due to slipforming operations, spacing and frequency of miniyoke pin positions,personnel exit/egress requirements, and shut-down requirements for environmental effects.

Environmental effects derived fiom studies of local site conditions, were specified in contractdocu¡nents. These included wind velocities based on 1:10 and l:100 year return periods at 10m and50m elevations, ground snow load and mæcimum/minimum temperatue values, a requirement forice build up thickness, and wave/current effects from which barge motion cha¡acteristics werederived and specified. Directional effects of wind were modelled using eight wind loadorientations, based on increments of 45o, over a 360o wind directional distribution.

243

t,'

l.t-

Constru ction Stage Reo uirements

most t¡pical,sEtrctures,.d*iog constr¡ction the Pr-imav Access System was designed for a.nr¡mber of snuct'rar configurations with fi¡ll operationar rive lãads æpriá ã eaph of the stages ofconsEuction' Five construction stage models selected fiom approxiäately 50 constuction stageswere developed as critical design cases for the Access system sqpport frame/maintower/miniyoke/main bridge assembly. The HSS steel ûämed to** were assembled in thefabricatot's shop into 2-storey higb súusüuctures using welaed connections The subsür¡ctr¡reswere assembled on site using botted connectio¡s

TVelded Connections

During recent decadcs, desig of welded HSS connections na1 F€n developed to the plese,nt stagewhere well defined formulations a¡e available for most of the connffins and load t¡pesencounte'red in practice. continuor¡s intemational research has regularþ-;te'raø the knowledge,butr¡ntilrecentlyresultswereoftennotwidelyavailableb"yoo¿*¿*í.publications. Toremedythis sitr¡ation' the canadian lnstitræ

"{T3l b"r'htr.d";-;ruished a *-ï*"i"*¡u, design guideby Packer and Henderson (lggz),(Ref. 8), which pr"r"otá the mosr helpftl informæion availableon welded and bolted HSS connectio* io-rpractiìing structural engineers. This book was usedextensively for designing the connections ofthis p-j"ã.

For the most par! the Hibernia Ptt l.y Access system contains HSS connections withconventional T' Y' )(, K or N configurations, with ¿omtant a¡rial loads and negligible bendingmoments' welds were sized by considering the effective weld lengths i¿ent¡nø in chapter eighq(Ref' 8), or for T, Y or x connections, usin! informæionfrom more recent research by packer andCassidy (1995),(Ref. 9).

other stucfi¡res ofthe system have T and Y connections \ rith substantial in-plane and out-oÊplanebending moments in addition to axial loads. These connections required -JÃuuo..r" desþ inthat the axial connection resistancg the in-pJane bending moment connection resistance and theout-oÊplane bending moment connection resistance alt had to be evaluated and compared with thereÐective forces, and then combined for total connection resistance. ---

Bolted Connections

The vertical legs of the tower sub-shr¡ctures \üere connected with the use of bolted butt platesplices' where possible, bolts were placed along oory *o parallel sides in order to r¡seformulations in chapter 2 of Ref. 8. There is an obvlous "oa

r*ã"rig¡r;d;ä bofted btrtt plateçlices where bolts are placed along the four sides of the connectorplate.

-I

I

t-

I

i

rit

iL

rL

t--

L

rL

trt

t

204

SUMIVTARY AND CONCLI]DING REMARKS

All th¡ee Case Studies presented in this paper are quite different in scale a¡rd fi¡nction but with acorrmon ingredient, namely, exposed hollow' stmcn:ral steel tube stnrctures. The combinedfeatu¡es of economy and pleasing appearance \Ä'as a major factor in the selection of HSS membersfor these projects.

Al*¡stgh there has been considerable resea¡ch and design aid development in recent years, for ther¡se of HSS members, additional development is required in bolted connections, quality assuftÐcein cold formed tubes, and code tolerances in the manufacn:ring process. As a result of experiencegained on these and other structures, it would appear that the most pftìctical and cost effectivemeans of joint connection of HSS members is a combination of shop welding and field bolting.HSS members continues to be the stn¡ctr¡ral steel type of choice for exposed spacial sûuctures.

REFERENCES

l. Adjeleiar¡ J.; Allen, C.M.; Huma¡, J.L.; and McRostie, G. 1986. National Aviation Muser:rn,Ottawa Canadian Joumal of Civil Engineering. Vol i3. Number 6. pages 722to732.

2. Nlera C.M. 1992. Toronto SþDome Roof Stn¡cture; Engineering Challenge. Innovative La¡geSpan Stn¡ctures. IASS-CSCE International Congress. Vol. I pages 63 to7l.

3. Allen, C.M.; and Duchesne, D.J. 1989. Toronto SþDome Retractable Roof Stadium - The RoofConcept. ASCE 7th Strucn¡¡al Conference. San Francisco. USA.

4. AIIen, C.M.; Duchesne, D.J.; and Humar, J.L. 1988. Application of Computer Aided Design inthe Ontario Domed Stadium Project. Canadian Joumal of Civil Engineering. Vol. 15 pages14to23.

5. In¡riq P.A.; and Gamble, S.L. 1988. Predicting Snow Loading on the To¡onto SþDome.Proceedings of the Engineering Foundation Conference, Santa Barba¡a, CA.

6. Allen, C.M.; and Packer, J.A. 1989. Stn¡ctu¡al Testing of RHS Joints and Members for theToronto SþDome Roof. International Symposium on Tubular Stn¡ctr¡¡es. Lappeenranta,Finland.

7. General Requirements for Rolled or Welded Stn¡cn¡¡al Quality Steel. CAN/CSA-G40.20-92. ANational Standa¡d of Car¡ada

8. Packer, J.A.; and Henderson, J.E. lgg2. Design Guide for Hollow Stn¡crural SectionConnections. Canadian Institute of Steel Constuctior¡ V/illowdale, Ontario.

9. Packer, J.A.; aird Cassidy, C.E. 1995. Effective V/eld Lengths for HSS T, Y and X Connections.Journal of Sructural Engineering. A¡rerican Society of Civil Engineers. Vol. 121.

205

WELDING OF STRUCTTJRAL ALT]MINT]M TUBING

By R. Bonneau*

aBsqacr

Atuminum tubing is used in large volumes in overhead structures supporting roadway and nafficsigns. The light weight of aluminum allows prefabricatiol of large sub-assemblies that can be

reãdily transporæd and quickly erecæd. The very good atuospheric corrosion resistance ofaluminum minimi2s the mainænance costs of the structures.

This paper describes the significant differences between steel and aluminum in reference to code

requiiements and welding fabrication. hactical aspects of avoiding difficulties when welding

aluminr¡m fibulil components are outlined. The conte¡t is a reflection of observations made inthe course of adminisfrating the CSA welding certifîcation standa¡ds.

INTRODUCTION

Overhead sign structures bpically consist of a rigid box truss, square in cross-section and

supported at each end by tapered tubular aluminum frames as shown in figure 1.

The structure may consist of one or more truss sections fabricated of 6061-T6 alloy. When

multþle tn¡ss sectionr¡ are used they are joined by means of cast ryrought aluminum flenges of356.0 alloy. These flanges are welded to the chords with inside and outside fillet welds and

bolæd together. The truss sections fastened to the supporting end frames fabricaæd of 6063-T6

alloy comprise the complete structure.

The main advantage of using aluminum is its light weight which allow long span with a lightstructure..

Desigr Reouirement

The overhead structures are designed according to AASHTO Standa¡d Specifications forstructural supports for highway signs, luminaires a¡d Eaffic signals.

The sign structure and the end frames must withstand a wind load of a 100 milelhour (160

km/hour), or a wind pressure of 55 pounds per square foot on the sign panels plus 45 pounds

per square foot on all the overhead sign structure without excessive deflection, vibration and

without permanent deformation, fracture or structural failure.

I

n

II

ll

I!

I

t'

fr'l

:

*R. Bonneau is with the C¡nadian Welding Bureau

206

No

9'III

I

i

:

;

J

Fig. I - Overheacl Box Truss Sign Structure

!LEvAltoNSCÂLE l:75

Fabúcation

For fabricarion of each truss section the GMAW process is used for joining braces ro the main

chords with fillet welds. see figure 2. Braces are cut and Eimmed for proper fit'

Theendframessupporttheendsofthelowerchordsofthesignsupportingsntcturesonplaforms a *ni.Hrlî, chords are fasþned by means sf stainlæs steel U bola' The end framæ

consist of two tapered columns joirr.à togrth., by mea-ns of filler welded braces. The columns

¿¡'g 5samlsss extruded tubes taperrã *¿-.o*"cied with fillet welds to a shoe base made of a

casting 356.0 alloY.

Thedimensioninmmofeachitemvarywiththetypeofstructu¡e:

Columns:Øzo3taperedtol52x6wailor2l|taperedtoJI3x6wallQslumns Brace: Tube 48 O'p' * 5 wall or ilbe 89' O'D' x 5 wall

Chords: Tube 89 O'p' x 5 watl or tube 127 O'D' x 5 wall or

,'- tube 152 O'D' x 5 wall'

Vertical diagonal õ^ rframes: Pipe 48 O'D' x 5 wall or 89 O'D' x 5 wall

Inside diagonalframes:

Horizontal diagonal

frames:

Pipe 42O.D' x 4 wall or 48 O'D' x 5 wail or 60 O'D' x 6 wall

Pipe 42O.D. x 4 wall or 48 O'D' x 5 wait or 60 O'D' x 6 wall

VERTiCALIIAGG|\¡ALS FRAII:S

:-rORtZON lALDtAGON.a,_S iRAV:S

It-(----

lERÏCÁLDIAGCNA'-S ÍRAM:S

It\¡StDEDIAGCNAiS FRÀVES

-ORIZONÏAL]:AGONALS 'RAMES

Figure 2 - Schematic Arrangement of Box Truss

208

CIIARACTERISTICS OF ALI]MINUM THAT MAKE IT DIFFERENT TO T1ELD

TÉdN STEEL

Preparation for Welding

Cutting and Edge PreParation

The cutting and edge prepararion of aluminum include atl the usua-l methods used for steel,

excepr flamã cuning,-¿ue tô tne aluminum oxide skin that has a melting point much higher than

the aluminum that it covers.

The success of mechanical cuning methods is related to high cutter speeds and suit¿ble rake and

clearance anglæ, to avoid loading up of cuner and the possibility of cutter jaroming or catching.

Aluminum Oxide

Aluminum oxide instantaneously forms on aluminum surfaces exposed to air. This oxide is thin,

transparent and has a melting remperature about three times higher than aluminum' The

thickness of the oxide film inciease rapidly at the beginning and then is self controlling.. An

excessively thick oxide film can cause welding diffieulties and affect weld quality as fusion may

not occur. Excessive oxide on the surfacs to be welded must be removed by mechanical or

chemical methods of cleaning prior to fit up..

Mechanical methods inciudes wire brushing with uncontaminated søinless steel wire brushes,

scraping, filing, pl¡ning and grinding after ttre surfaces have been cleaned of oil and grease.

Chemical metfrods includes causric soda solution and proprietary products. They are useful for

batch sls¿ning. The interval between cleaning of the su¡faces to be welded and welding must

be as short as possible, usually within 6 hou¡s.

Oils, Greases, Other Hydrocarbon and Loose Partides

Oil and grease films and loose particles on the edges to be welded wilt cause porosiry in the

weids.

Solvent degreasing.applied by qpraying, dÞping or wiping are used, prior to fit up. Non-residue

leaving solvent must be used.

Water

'water on surfaces o be welded may result f¡om outdoor exposure or from condensation caused

by temperature changes. The surfaces must be dry before welding.

Water stain must be removed with disk grinder, a po\¡/er-driven stainless steel wire brush or

other abrasive or machining method or by chemical methods.

209

I

I

TVeldabilitv

Eigh Eeat Conductivity of Aluminqnt

Aluminum conducts heat away from a weld a¡ea atarate 3-5 times as fast as that when ¡r¡elrling

steel. Welding currents and welding speeds must be higher and stringer beads are generally

used.

Eigh Coeffrcieut of Thermal Í'.xl¡ansion

Aluminum expalds about twice as much of steel for a given increase in æmperature. Stress

induced by the contraction during solidification may cause excessive weld joint distortion or

cracking unless proper welding procedures and filler metals a¡e used.

trìlter Metal

High srength alloys such as 6061 or 6063 a¡e welded with filler metal of different composition

than the base meAl to prevent hot cracking. Hot cracking occurs during solidification when the

metal is passing between the liçidus and solidus temperatures under contraction strains. The

standa¡dgrecommend filler metals having enough silicon or magnesium such as 4043 or 5356

to produce a crack resistant composition in the weld-

Preheating

Preheating of aluminum is not generally required. Whren welding thick aluminum sections,

preheating is sometimcs used to avoid cold-start defects to achieve heat balance between

ãissimilarthicknesses, or to remove moisnue from the metal surface in the welds joint area.

If preheating is necessary, the application of heat should be for as short a time as possible 15

minutes marimum and a base metal temperature of 120"C should not be exceeded as the

propreties and metallurgy of aluminum alloys are almost always affected adversely by elevated

temperatures.

No Colour Change During Heating

Unlike steel, during heating aluminum shows no colotu sþange during heating. The welder has

to look carefully for a liquid wet appearance of the area being heaæd to know that the metal has

begin to melt.

tttttttirL

ttt

ttrL

L

|..IL

iL

t210

**"'ffi'#o€'ff#R{'fi"fi BäüH'^ffiffiil'**Avoid Craters in the Tl'-eld

Most weld craters contain cracks; both tr'2nsverse and longitudinal types are usually present'

These cracks may extend into the weld bead or into the parent metal under service conditions.

Crater cracks can be repaired by gouging out the unsound metal a-nd rewelding'

crarers can be avoided by proper manipulation of the torch and/or filier metal in manual

welding.Thetechniquesforterminatingaweldincludes:

- accelerating arc travel speed just before releasing the gun trigger;

- reversing the direction óf travel for a dista¡ct suffi.itot to create a smooth transition

_ providing suitable build-up and dressing the crater area flush with the weld surface by

mechanical means

Stop/Start

When welding braces to the chords of truss section or colum¡s of end frames' the stop/start

during welding should be made on the side rather than in the toe and heel a¡ea of the joint-

Incomplete Fusion

Incomplete fusion occurs when an aluminum oxide film present on the surfaces and is not

completely ,."*ourã either by cleaning prior to wetdinq or by the scouring acdon of the arc'

unrike steer, rhe oxide film ii insolublã io tn. weld pool and is high melting point prevents ir

from being melted bY the arc-

Other sources of incompleæ fi,lsion are inadequate joint spacing or edge preparation and too long

a welding arc.

Incomplete Penetration

In fillet welds, incomplete penerration resulß when the filler metal tends to bridge accross the

toe of the joint and not peneÍate into the root'

In groove welds, incompleæ penetation occurs when the weld bead does not petretrate the full-

thickness of the p."nt ,ort"l when welding is done from one side or where the weld beads do

not inter-penetrate when welding is done from both sidæ of the joint'

This defect is usually caused by insuftrcient welding current; arc ravel speed too high; too long

an arc; inadequate edge penetration.

211

Overlapping

This defect is caused by a welding current too high and improper welding æchnique.

Undercut

causes include welding crnrent too high, arc Eavel speed too low or improper torch angle

Porosity

Hydrogen is the most common source of porosity inalumr¡um *9le.. $ldrogen is introduced

io tn.ïud pool from water vapour, grease and oil, surface oxide in the weld zone or either

from residuailubricants tlat conåin hydrocarbons or from hydraæd oxile films on thl surface

;irhr;;Hi"g *itr. wnro these conaminants enter the arc they are broken down and hydrogen

is liberated. In the molten state, aluminum absorbs 19 times more hydrogen thal it can sustain

after solidification. Depending on the rate of solidification, the hydrogen released in the weld

may become entrapped causin! porosity in the weld. Fast solidification rates result in greaær

porosity than do slow rates.

Improper Fillet Welds

over grinding of fillet welds or a too concÍrve surface can cause a reduction of the effective

throat thickness and cracking of welds in service'

Control of \trelding Yariables

Main variablqs which need to be controlled are:

. correct welding arc (stabte, with sufficient energy, proper lenght)

. correct electrical power sor¡rceo matching of welding consumables with base metals

o care of welding consumables. design of welded connections. clea¡liness a¡d protection of jointo manipulation or confiol of welding electrodes

To properly connol these variables the following is required:

. welding procedures for continurty and consistenry during the welding operation

o skilled wllders for the process and position used

e Qualified supervisor reqponsible foiensuring that welders, welding operators and tack

welders weld in accordance with approved procedures

o Qualif,red engineer responsible for welding design and welding procedures and practice

FL

E

t,

E

E

E

tttttttIE

t

tt

t212

ALI.iMINI.IM \4IELDING STANDARDS

In Canada, it is a contractual requirernent of provincial uansportation departments that welding

shali be ca¡ried ou, uv companies cenified by ttre canadian welding Bureau to the requirements

of CSA Welding St¿ndards-

The Code and standards associated with rhe design and fabrication of aluminum are given in

figure 3.

Fig. 3 - Aluminum welding: codes and standa¡ds Involved

CUSTOMER'S SPECIFICATIONS

AASHTO STANDARD SPECIFICATIONS FOR

STRUCTIIRAL ST'PPORTS FOR HIGH\ilAY SIGNS'

LIMINAIRES AND TRá'FFIC SIGNALS OR

EQUWALENT FOR DESIGN

csf,.w47.2CERTTFICAT]COMPANIESWELDING O

:ON OFFOR FT]SIONF ALI.'MINT'M

csA \il59.2WELDED ACONSTRUC

LTTMINT]MTION

ANSUAWS 45.10BARE ALT]MINT]M ANDALUMINT]M ALLOY

WELDING ELECTRODESAND RODS

ALI'MINT'M ASSOCIATIONSSPECMCATION FOR BASEMETAL ALLOYS

csA w178.1CERTIFICATION OF WELDINGINSPECTION ORGA}'TIZATIONSCSA W178.2 CERTIFICATIONOF \ryELDING INSPECTORS

213

rdards

^lsr"od"' dW4T.z"Certifrcation of Comp¡niss for Fr¡sion Welding of Aluminr¡m"

jsundard specifies the requirem:nts for documeûraúon and verification of a basic quariry

"Ët;";-íor welding' Ii includæ requiremens for:

, Welding PersonneíQualification : ::l:-t:.:,. Supervisoro Welder

o 'Wetding Engineering Standards

. Welding Procedure Spectt-tcauons

. Welding Procedure Data Sheee

)

I *"..rrary Welding and Auxiliary Equipment

Use of Bæe Metal and Filler Alloys conforming to standards

Third party verification and audits

A Standar d,W5g'2 "Welded Atuminum Construction"

ì

^,1 standa¡d specifies the requirement for:

Design ãf Wtt¿e¿ Conne-ctions

) w;lãi"e ConsuÀaules, Workmanship *¿ Jsshnique

I À"."p-r":*e Criteria for Welded joints

, Weiding lnsPection

,l - c^- D^-a Àt,rninrryn and Ahminum Alloy*sUnws standard a5.10 "specifrcations for Bare ah¡minum and AI

.\ding Electrodes and Rods"'

;,L rono* dsw4j.zand w59.2 require that welding rods and elecuodes be certified by the

,radian \Melding Bureau as conformid; rh. requirãments or eNsvAws standa¡d A5'10'

I

jAstandardW1TS.l..CertifrcationoflVeldinglnspectionorganizations''l

I sundar¿ specifies the requirements for:

d;;;;*'ã"".r quatificaiion: : ìJ,no".i,i:'i ' Teit EquiPment oPeratorI

- llding insPection Procedures

bessary testing equiPment

hird partY verifica¡ion and audis

I

Welding Procedures

214

CSA Standard IV178.2 "Certification of TVelding rnspectors"

This standa¡d specifies the requirement for qualification as inspection superviso¡ or inspector.

CONCLUSION

As in the welding of steel, there are many variables to control when welding aluminum. properbase meai preparation before welding and fit up are key elements.

Welding standards provide information on qualification of welding personnel, procedures andtechniques, welding equipment, consumables, acceptaace criteria and inspection to assu¡e thataluminum weldments will meet the service requirements.

ACKNOWI,EDGEMENTS

I would like to rhank ¡¡s À4inistere des ftensports du euebec to have share their experience asuser of the structures and to Lampadaires Feralux Inc. for their cooperation.

REFERENCES

Welding of Aiuminum. Alcan C¡na¿¿ Products Limited, Sixth Edi¡ion, 19g4.

, The Aiuminum Association, 'Wæhington

D.C., 1977.

Canadian Standard Association, CSA W47.2-M1987, Certification of Companies forFusion Welding of Aluminum , lgï7.

Canadian Standard Association, CSA W59.2-M1991, Welded Aluminum Construction,r99t.

Welding Aluminum with the inert gas processes, Australian V/elding Resea¡chAssociation and the Austalian V/elding Instirure, AS/RA-AWI Technical Note 2, 19g5.

1.

2.

3.

4.

5.

215

it1i

!

' ,E-BASED E)CERT SYSTEMS IN THE TI]TT]REhn cTwLENGE oF KNoIvLEDG

ôr rlm DESIGN oF TuBULAR srRucruRrs

G Davies*, W Tizani* and K Yusufr

þ, pup", examines the potential of Knowledge Based Expert Systems in the design of rubular

dtn¡ctu¡es, drawing on experiences gaintã ¿*åg development of a system aimed at supporüng

{esig¡ers in pro¿ucinË, *ã""*g *d d;;ïbdartnrss designs' Th" ryryT *'orks bv guidine

þers towards desig'recisions that recognise the consequenc", io, cost of fabrication, and this is

llusrated bY a design case snrdY'

KEYWORDS

hnteerated Desigo and Constn¡ction, Knowledge Based Expert Systems' Decision Supporq

E.oão*i" Appraisal, Cost Modelling'

ì

i 1¡TRoDucrIoN

ì

lw" H.r" in a rapidiy changing w9rld, where the rate of change is faster tha¡ at an'v dme in human

,history. The relative ."onã*v of electonic communication and the great strides-made in the field

iof infomlation technology appearto bl the dominant factors which ir" ati'rri"g the other maniford

lchanees taking place in ari area¡-orrif" i";r"di"; th"se in the constnrction industry-- we a¡e familia¡

,with word processing for Specifications, i.re^r-o9"r11111and Bills of Quantities' with compuær

rAided Drafting of drawings and their electronic transmission beween interested parties, and

lComputer Aided Maoufa"ãrr" in the *u"hin" shop' $/e a¡e well versed in stn¡ctural analysis by

computer and a¡e also getting ullto th" il;;il"ttyi"g outthe design itself at a terrrinal' We are

iabte to view our designs fromdiffer.n unìr", *ãíiewpoints at the:ou:l:-*:Ï"t' and before

f too long we will b. ;;1.1o*ak through]st*"t*es and view or:r dreams using Virnral Realiqv'

Alr these have the advantage of ease orpro¿u"alon and of initiating change witb rapid response'

ABSTRACT

\!'hatever our present position and current perspectiv-e o1 ,h: role of information tecbnology, it is

something we cannot iossibly iqro¡e pr ri" n,*e. professional structural engineers wilr arw-ays

require a sound..¿"åi*¿irrg ãrr¡. bJ;;;;Ptes of their professior¡ and-a-firm grasp of the

relevance of ,t .r"n,,Jã"*i tã ,tn "t,,a

rur..r, ró *,u, ttrer.i¡ont{r¡':-:T:lt:uv manage the

design and constn¡ction of engileerins oi9.,;Ë. H;*tt'¡', ÏP ü" rapid expansion of information

in the knowredge field it is widely ,".ogrú;"ã ,hu, a surfeit of indiscriminare information can easily

swamp the engineer. ðË-r", *ií i, n""¿ã¿ is that ttre informæioi be rrad'y available in a distilled

form which utro u¿"i'"s th"""' on how it is or is not to be used'

@eering,UniversiryofNottingham.UniversityPark,NoningharrNG72RD'UnitedKingdom

216

Traditionally this knou,ledge tempered by'experienced l,isdol has been in the hands of the Expert,

*,ho is much more rhan à speciatist, uul" ã ,L*on about the appropriateness of the use of the

inforrration in generar a¡rd for particurar r*.r. e, fe*,er and r"w"ip"opte with this required expert

knowledge a¡e readily u"".rribl. fo, "o*Jäiion

oithin sm'l toïe¿ium size organisations' t'e

appropriate ioro,,outi'o'i;ã;i; be readily available in some other way'

It is unreasonable to assume that a smlctural engineer *U:.T"V,b required to practi-ce across a wide

range of *.*..,r-;;'r-'l ãro P],: i;;"diui. ,t"¡l i" d"pth * *v p*'i"*- aspect'

'with

tubura¡ srn¡crures,,.ã f. assumed rh";;;*tg"ers *]ll ta* "

good grasp of member design

procedures, but ma¡, u" *ø*'ia¡ u'ith th" ;pecrJ i":Jt't:î ï*ciäe¿ *itu tuittute joint details

ro satisfy both strength and economy. And what olt¡. young inexperienced engineer also? There

r*-ould be considerabíe aduantag. in ft"ting so'n" ;i*tUigent';s'ppon alongside at the design stage'

The paper describes *"h ;;;;ro".t *ui'"i *iìl ,ir""r,êously attow the investigaror to examrne

several different lattice girder layouts, ó*;;;th advice oo ho* to modifi the joint details i¡

terms of sfren$h a¡d relative costs'

Nethercot and Tizani ßef. 1) have recently summarised ¡om3 of the advan*:'h:lll"e been made

in applyng ior"rrriào i""rrrroroçu **d,1, ¿"r*"tion industry. They pay particular attentron

to tbe themes of using Krou,ledge-u^.u Ë*ã;]tr,.* rr"h"tãurt *¿-i"tegræion of the design

a¡d consmlction process, *ithin the-a¡ea of iteel ton't't'"tioo h"y aso o"tlitt" advances made

r¡nder the themes or 't ir*¿i*tion' and "";";;'i* and explore ttt" titt"ty ways industry u'ill be

effected and operate in the future' The main points are:

. lncorporation of intelligence' as a supplement within design and decision support tools'

. lntegrarion of the design pro..rr, *nåurr.na and constructionled design'

. visualisation of the consmlction proárrrr, 3D design and modelling tool5 for production'

. Effecrive communication betwe."ä"""oJ.iti"r"äittror"i"g rut a-¿r for the electronic

.ornrruoltiã" oii"ro""ation of data and information'

This paper describes how the frst rwo aims have been rearised in the design of rattice girders formed

from ci¡cula¡ Hollorv Secúons (cHS)- rr'* 'y'** workl r1 such a way as to allow the designer to

remain in conror and to respond "eerd;¿ry io,¡, advice offered. Fina'y some comments are

made of what is realisticall-v possible over the next few years' a¡rá w¡at effects this will have on the

sbape of the indust4''

BACKGROUND

The trad.itionar approach to the production of steel stn¡ctures in the united Kingdom and many other

counrries is that the srructtua outline ÃJmember designî carried outby the Engineering

Consultant on behalf of the Client. wt"'"lont'u"t' "" u*-d"ôon ttre basis of competitive tenders

the consultant and even the main *n*"a, *ill not have b";;;;t "oncerned

with the details of

the stn¡cnral connections. at tbat rog., rniliriog i.g.ty-teftto tfre to the ingenuity of the fabricator

uùen finally chosen. Thus at the early t"gt th$t'ióti f'* u"w fi6e information on how to form

the connection or u-hat it is likely ro ,oî' ri" t"iL¿o*o of 'h""o'o

involved in the fabrication

217

process is commercially sensitive and often jealously ggarded by the fabricator' It is therefore

difficurt forthe designeito estimate what econãmies can be made in the fabrication process. often

b¡r the time the fabricator is involved in checking the stength.ld cos¡ "t,-*jj:Ï" the member

sections have already been ordered on the basis o¡'member only'requirements and the joints can

then only be made strong enough by expensive stiffening -*d *elding' To avert this danger a

fabrication led approach"tras ue-en á"u"ñp"d, where the designer is enabled to carry out full

economic appraisals taking into accormt not;dy the member but also joint fabrication requirements

at the initial design stage.

SuchanapproachisdependentonthemorerecentlydevelopedtechnologiesofKnowledgeBaseExperr Sysrems (KBESi and Object Orie¡t¿ted Programming (OOP)' IVhile conventional progams

such as those written i' fo*u' or Basic are used to carry out numerical calculæions or retieve

information, KBES computer programs ar" oesigned to manip-ulate knowledge as well as dzta' These

more flexible systems may be r:sed to representîum* "*p"ri"ote (knowledqÐ P a particular field

and also to provide advice on how to use the information (by applytng logical deduction and

induction procedures) as paft of a reasoning Process gull:d-'T lnference Engine'' In contrast to

convenrional prograrnming OOP assists pt"il;.rs by linking together those parts which form

consistent ,etationst ips ãi i.¡ world óU¡"i"' fn" con""ntiooal flow diagram is replaced by

hierarchy entities caUå¿ classes or objectstorn:nunicating via message passing'

The work described in this paper was carried out atNottingham universir."-* and sponsored by the

Engineering and Physical Sìiãnces Resea¡ch Council, as a protot]?e investigation for all forms of

steel construction. Tubula¡ lattice girders afe Pafticularly interesting as tne cost of the joints make

an important contribution to the total cost of ihe girder, and a minimum weight solution based on

the members can be quite misleading. Getting the joints wrong can inflict a large cost penalty' The

project has been r.#;;; ógs ñ¿. wiãtr trreir reduced nr:nber of orientarion options, in order

to concentrate on getting the programming right'

The system also operates in such away ttrat it can be useful to the fabricator in managing the

throughput of several different jobs in the production shops in an economical and efñcient manner'

FABRICATION.LED DESIGN PROCE SS

In order to overcome the shorrcomings of the traditional design approach to the.production of steel

stuctures, a fabrication-led design Process is advocated' ln this ptåt"tt tbe designer is empowered

to assess the practicalitv of various design options and lhe relative economics of various altemative

design details, ,*'ith a v'iew to alleviating theìecessity for expensive fabricarion operations such as

sriffening during fabrication. For the aeiign of tubular stn¡ctures the fabrication-led approach could

involve the following sequence: ,oo"r*ut analysis and design, joint capacir¡- checks' design

critique, and cost appraisal, r*jth modifi.*ior6 being made in response to results at each stage'

A prototype Inregrated Design system (IDS), aimed at supporting designers in carrying out

fabrication-led designs for tubular tn¡sses rtL u".n developed ßef' 2). The IDS is modula¡ in nature

and consists of links to standa¡d *¿ysis and design packages' a joint and a member capacity

218

checking module, a knowledge based expert system for design critiques, and a cost model used forcost appraisal, as shown schematically in Figure l.

Figure l: The lntegrated Design System

The Joint Capacity Module checks the joint capacities of the tubula¡ truss. Allowing the designerto identify inadequate joints, and e4plore other less expensive methods of remediai action, such as

increasing the chord thickness, or reducing the gap between braces, before the member design isfinalised. These checks are typically not ca¡ried out in traditional desigr¡ a¡d can prevent the needfor stiffening, which is prohibitively expensive, and is required often as a result of the memberdesign prescribing sections u'hich aithough capable of transmiuing the required forces, a¡e urableto provide adequate suength at the joints. The module applies the appropriate formulae fordetermining the joint capacity, which were obtained from the IIW (Ref. 3) and the CIDECT designguide (Ref- 4). The forrruiae are applied to the joint, based on the joint type, and the designerinformed of the adequacy of the joint. The designer can request a detailed report higblighting thestn¡ctural efüciency of the joint, identiffing the anticþated mode of faih¡re. Through links to anadvice knowledge base, the designer can be further provided with advice on how to improve the jointcapaeity if required- The designer is able to check individuai joints or carr), out a global chech thatinspects all the joints in a given truss. The integration provided by the IDS. mean-s that the user doesnot need to input any additional information for these checks to be executed. \

The Member Capacity Module examines the stn¡ctural adequacy of member sections, in responseto a¡y modifications made to a joint or Euss, and ensu¡es that modifications made do not result ina¡ unsafe stn¡ctue. For instance, in reducing the gap between braces at a joint, in order to improvethe joint capacity, a moment due to the resulting eccenticity is set up in the chord. The membercapacity module checks that the cho¡d is capable of resisting the moment, alerting the designer ifthe chord is inadequate.

Economic Appraisal Module

;_____sz_1\

219

ì

I

The KBES, comprises an Inference Engine, Design, Fabrication, and Joint Capacity KnowledgeBases- The inference engine consults the knowledge bases to evaluate and cámsrent on designdetails, and provide advice on remedial actions. The design knowtedge base mainly contain rules ofthumb that represent good practise. These include n¡les thæ check if the tnss is too deep or shallow,identifu arangements that involve high or low brace angles, and advise on the steel grades used foithe design. The rules in the fabrication knowledge base, examine the layout of the tn¡ss and the fitup of its members, identifting features that may adversely affect fabricàtion cost. For example thepresence of overlap joints, the occurrence of a member that overlaps others at both ends, or theidentification of a particular joint t1pe, as an expensive fabricatiãn detail. The joint capacityknowledge base provides advice on how to improve a joint's capacity. Its rules exa¡ine the modeof failnre and generate advice on remedial action. For instance, computation and advice on limitingvalues in response to a validity violation during a joint capacity check. The knowledge baseimarripulated by the inference engine, fulfill an advisory role, drawing the attention of the ãesignerto adverse details, and recommending suitable modifications.

The Cost Model estimates the cost of fabrication providing indicative costs for alternative designoptions- It consists of an object hierarchy, that is made up of objects that represent the variousentities and relationships within the fabrication process. These include stuctural entities, such astn:ss and joing fabricatiòn operations, like weld and cug and fabrication machinery such as profilingmachines (Ref' 5)- The model estimates cost by representing the cost of carrying out the basiõfabrication operations(cu! assemble, weld etc.). These ¿¡re computed using a combinæion ofmachinealgorithms and rules of thr:mb. The cost of fabrication is estimated in minutes, to which a cost rate(which can be modified by the user) is applied. Material costs a¡e computed by calculating the totallength and tonnage of each section, to which a rate is applied from the .,-"rrt price list. The costmodel supports the existence of a number of tnrsses, enabling comparison ienryeen d.ifferentschemes. It also supports the costing of an entire truss or individual joints, facilitating global andlocal compa¡isons- The cost model is tuned to give relative costs, the aim being to support thecomparison of various alternatives with difterent fabrication and material content.

Supported by the IDS, the designer thus has all the tools required to underrake a fabrication-leddesign of tubular trusses integrated into one envi¡onment. using the IDS the following designsequence could occur: The design process commences with the designer selecting a number ofsbr¡ctural solutions for the tn:ss, typically based on past experience. The designer then analyses andselects adequate member sizes for one or a number of these solutions. Theãesigned truss is thensubjected to joint capacity checks, the outcome of which may inform recommended modificationsto the joint and member details. Hal'ing obtained a sbr¡ch¡¡ally adequate rnrss, the designer requestsa critique of the design. This highlights design details that should be avoided. or may represenrimprovements from either a design or fabrication point of view. The designer might make furthermodifications in response to the cofirments made, or note them for later investigaiion.

A summary'of the cost of manufacnring the truss can then be obtained, the summary identifying thetotal time required to fabricate the truss (in minutes), the total weight of the r*riin tonnes), andthe total surface area (in square meü'es). The fabrication time and surface area are converted intomonetary values by application of a rate w'hich can be modified by the designer, the material costbeing based on the current British Steel price list. A number of feasible solutions for the truss can

220

thus be assessed based on fabrication and material content-

The designer can further inspect the cost associated with individual joint detailing, embarking on a.ï'hat-ifiscenario. by modiffing details and requesting cost assessments. These scenarios may be

as a result of comments made dr¡ring the critique of the design, or may be required to judge the

sensitiviry* of the cost of a detail to changes in the joint geometry or welding specification. The

designer óould also assess different splicing options, The design process as facilitated by the IDS is

illustrated using the follou'ing design case study.

DESIGN CASE STUDY

A scheme for a flat roof truss solutior¡ involving K type bracing arrangement, is shown in Figure

Za.Tbetrusses a¡e to be placed at 6m centes and has a 36m span divided into l0 panels. Nodal

loading is computed ar 32.4IOrl a¡rd the analysis and design results a¡e shown in Figure 2b- Members

-uy b" placed into groups where the same CHS section is used as decided by the designer. ln this

*r" ttuãy based on the guidel.ines in the CIDECT design guide, a single section is selected for the

chords and the number of selected brace sections is restricted to ¡¡¡o' All member sections are

527 1JZH (Grade 43D) material grade.

(a) General Arrangement

cHsf 68.3X5.0

cHs60.3x3.2

cHs88.9X4.0

(b) Part Structure

Figure 2: Details of Case Study Lattice Girder

cHs139.7X6.3

ø r*ì

221

The tn¡ss was then imported into the IDs environment, via the lirks to the analysis/design package,

and designated T5, Figure 3. On importutioo of tn¡ss dat4 tDS ascertains the joint types and

computes the joint geoãetries (i.e. o''gies, ;;Py";¿tl"Ps) assuming zero eccentricity' It also attaches

a defaurt werding specification with ãro-"rrtt"t welds to the bracls at the joints, this specification

can be modified bY the user'

I

Figure 3: Truss T5

An examination of the joint capacities reveals failure of the chord plastification criteria in the top

chord joinæ J2, J4,¡0, ig, ¡f O,^*a tfr"1r¡f-*"ttical equivalenS' ihe advice presented a nr:mber

of options involving .i,*åi"g itt" .n:tq-*Ãss, reducirg the gap between braces' increasing the

brace diameter and stiffening. Typicalry;;ütit 'og", tttt øu¡tutor would have no choice but to

stiffen the joints, and ,rr" "rr.I,

or-¡ri, limitation "*-b" readily explored by the dgsiener in the IDs'

The IDS can automatically apply u "*itffi Oift*t to u¡oioì' selecting and placing a suitable

chord section at ttre joint to impart ua"q*L'rtiffrress, as has been done for joint J6, utilizing a can

made from cHS 16g3;6.3. The fabrication cost of the stiffened joint J6 is shown in Figure 4a' An

alternative to local stiffening is to change the entire. toi chord from CHS168'3x5'0 to

CHS168.3x6.3. The .r".iortr,î, simplifrcatiín i' upp-tnt in tne revised cost estimate for joint J6

Figure 4b, showing a 3[Voreduction i" ;rt "attiuu,"¿ ,o suitable savings in cutting, assembly'

werding erc. due to trre Jmpiified detaü. ruri"g into consideration the facr that local 51ffisning at

ttre ten inadequate j"irl *u¿ be avoided b;;;g:? *ctiolr¡1 ttre top chord that is onlv 20% more

expensive than the;;;;l d"rigleg section, *¿ tnt economical advantage of the fabricationled

approach becomes q;i;;pú.ã,. rur*¡"r rä"itur "o"ld T.-Tade

due to the fact that delavs that

would have occurred while the fabricator and designer commrmicated to decide stiffener details have

been avoided, as well as the high premium^*"iãr"¿ with buying the relatively small quantities of

cHSl6g.3x6.3 required to form the stiffeners. The prices g.n"*La by the cost model are relative'

their main purpose being for comparison oiJr.*,u,i't e details' Howevãr they show close correlation

with cr¡¡rent Pricing in the UK'

AdesigncritiqueofT5advisedthattheuseofhighgradesteelbeusedforthechords'whileafabrication critique higtriighted the presence of oveilap joints at J3 and its slmmetrical win' The

222

cost of J3 is shown in Figure 5a The joint *'as then simplified by gapping, uith the IDS computingthe new geometry. and checking the joint capaciqv. and capacity of the chord due to the inducedeccentricity. The reduced cost of the simplified derail is shown in Figure 5b, the IDS guiding thedesigner towa¡ds economical details.

(a) rù/ith local stiffening

Figure 4: Alternative cost of Joint J6

Figure 5: Altemative cost ofJoint J3

Before costing the tnrss, flange plate support connections were specified at the support joints (Jl,J?,J?2,J23). having a 16mm thicloress and 6 grade 8.8Ìvl22diameter bolts. The trss *us spliced,the chosen configuration having splices atl2mlengths with nvo braces ransported loose. The IDScurrently supports the placement of bolted ring-flange splices or butt *rldr, the former wereselected and computed as 20mm thick plates with I No.M22 d.iameter bolts. Having obtained astructurally adequate truss. with all the fabrication details specified, the cost of T5 was then

ß BD! stlfÍlllú lXSló1.3tr4.3 llT¡ata ?r.p¡r¡t16 ! ¡¡ ¡16.

Cüttl¿9 to lr¡gtÞ : 32 d,É.tioflllDg : t j,.t.Drll¡t.g : a d6.¡¡sdly : ¡2 d¡r.L¡dlng . J2 d.É-!i6il!9 i tS .1.t.lnspÈctlo : ã d¡t.

(b) Without local sriffening

r.c.rE.r¡oñ Eost: :a¿.! Ers ¡tùi5 cúsitti of:¡r+l¡t. ?r.Dù¡tion ! ¡g ¡i6.û¡ttin9 þ l.Btñ : .tó úË.Proti¡ing : 13 ¡i6.Þi¡¡i.g : a riÉ.ttFÈ¡9 : ¡¡ ¡is,I?lÉ1n9 3 2, d,B.DGrl¡g : tt ¡tE-¡D3tlction 3 lS dE.

(a) Overlapped Joint (b) Gap Joint

223

#,'ri:',rf;htri'#*1"-'' lldt¡¡¡Ð . udr. l-orilli¡r9 | i¡¡ ¡n. Iñrdrg . SrS.ir- i:L¡ó¡¡B . ¡¡ .¡r. l'fg'!- ' ¡i'i-' I:

äåiî:LiifgËHìiÏ ff;ilr,',"ä,5,:{iì i.

ffi:ä:!iËffii*:Ei,rr,Ëä: ffiËñä'? rs* (e'trt)

þute4 and is shown in Figure 6'

I Figrre 6: cost of rrus T5

.l.*, stage the designer could carry.out,fi'ther investigæions, for ins'nce determining the effects

oiusing a higher **i:'"r,*ü; ,n" i,r,*¿rî"îäJ."ï. ifuLi.u,åî*orîi"g sire welded butt join¡

-p*;;,*;e"f":î;ï[#"iru;"-'Ëç:*i*iHf::Fji#Ë:Jäffi :þs schemes, tor tns

rbs irrt"grut", ir,,o oäî;;i;,;;"J¡r tnt iå"f' "qt'iita

to i"""rop an economic¿l and practical

::Ï;r;;or"r"n .u been only subjected to the conl;ntionar process of design' with all

the associat.¿ "r,*elî;;ö;;* .* i1.?äi"utio" '9q''. #äJisu"u 1

design would be

sisnificantly higher. Tht, ñ :{i h.Y., îiãi*t¿ r"u'i""ti* ii-t' u'" '¡t: tut wasted in

relolving the conflict, *ã *o,rrd irave u,å. ;t;lös r,., **ï* rrtrut.t-To: at tbe desieD

$age, ens'red that " #äi; r"r"ri"" t"* u"*ä"-r*.*¿ ä¿ ru¡rication rie*?oints has been

rt . -.aca.srsd demonsmtes the scare of benefits th1 can be derived from the use of the

rhe designcase.present:ij;i:;1it:;r$l design firnctions;;;il4'1:li"*ent' roint

'iDS. Since tt'e designei is able to execute itiji:i 1nï#ï:'""ift*î"t options can be easil-v

Lm:"'.'m:#hï:"*:,ä';+:m'"tf:+iï;J:ffi:.'n:ffi îq'q*rï'

enã¡"i,,o,tr,"n*¿ro-,;;i:!;sr.-"i{îiijf :hrtr#,f ;=Jii|'LH!f iH:îåïfJ#jffi'*üã; risk of industrial disputes-$ a

no, orrlv alerts the o"lîää'T" *""'oo'" lï:#iö:'äå;;;:1:,.:*Í,f;i,i..Ír-i,:î;;; artematives, """ijå"'*ã

uv- ioa'u'ive costs' B-v

lùe designer to T":i*-:i,^,;;"';ki"; rabrication-ted design"9'^lïr1T::iJ.fi"^from thesesupporring trre desigîä;ï;J"n oi"* däÄ;jnu *"*"''* ñs ru"'i'tts the production

r$åî:#:r#::äîärîäiii#ä;iffi ii""'*.r.ffi ïffi .'i.:îioää'."*

;;ñ;nefitbeingtotheciient' - rr^.iæ withall

achieved.

224

STTMMARY AND CONCLUSION

The development of the IDS has successfi.rlly tested the feasibility of using information technology,en-eineering knowledge and cost data, to develop a tool that ca¡r practically aid execution offabrication-led desi-ens, by allowing the simultaneous consideration of design and its effects onfabrication, in effect integrating the design process. The IDS fi,rther demonstrates the potential ofincorporating "intelligence" as a supplement \ /ithin the design and decision support tools, as shownby the embedded knowledge based system,that enables the critique and advice functions, providingaccess to "expert" knowledge. With minimal additional effort at the design stage, the IDS preventsesoteric solutions and ensu¡es the production of workable a¡rd competitive designs from both thestn¡ctural and fabrication l'iewpoints. This not only heþ to reduce costs but also saves time thatmight be saved in resolving potential disputes. Futr¡re work will extend this approach to includeother constn¡ction stages such as tansportation and erection and will broaden the structr¡¡al formsconsidered.

The paper has indicated one area, using circula¡ holloç'sections, where integration of design andfabrícation in a cost effective way is nou'possible. Such procedrues will however only be generallyaccepted, as education and training to meet this potential is encouraged and developed across thewhole steel constr¡ction field. Such developments will make progress, if the potential is recognisedby all the players in the ¿¡ren4 including computer software companies. What of the shape of thebusiness in the fi¡n:¡e? That will be for you to decide, as you leave a¡ld reflect on the usefi.rlness ofwhat you have hea¡d in meeting the pressures of deadlines for both designers and fabricators.

REFERENCES

Nethercot, D. A.; Tizail U/. M. K. 1996.IT in Constuction: Adva¡rces and Potential.Submitted to the l5th IABSE Congress. Copenhagen.

Tizari, W. M. K.; Davies, G., McCarthy, T.J., Nethercog D.4., and Smith, N. J., 1993.A knowledge-based approach to constn¡ction-led stn¡ctural design, in Informationtechnologies for constmction. cir,il engineering. and transport. Powell, J. A. and Day, R.@ditors), Brunel University with SERC, UK, pp.30l-309.

IIW 1989. Design Recommendatíons for Hollow Section Joints - predominantly staticallyloaded. International Insritute of V/elding- Doc XV-70-89. UK.

Wardenier, J.; Kruobane, Y., Packer, J. 4., Dutta D., andYeomans, N., l99l- Design guidefor circular hollow'section (CHS) joints under predominantly static loading. publ. CIDECT.Verlag TUV Rhineland German]¡. 1991, pp 46-51.

TizanL,W. M. K.: Davies, G., McCarthy,T.J.;Nethercot, D.4., a¡rd Smith, N. J., 1994Cost modelling for the economic appraisal of tubula¡ tn¡sses. in Topping B. H. V.A¡tificial and oriented approaches for strucnral engineering. publ. Civil-Comp Press, 1994,pp. 59-67.

t.

2.

J.

4.

5.

22s

Documents

International Conference in Tubular Structures-1996