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SEISMIC VULNERABILITY ASSESSMENT OF BRIDGES FOR RETROFITTING AND NEW
DESIGN
By
PEDRAM FAROKH
A thesis submitted to the
Graduate School-New Brunswick
Rutgers, The State University of New Jersey
In Partial Fulfillment of the Requirements
For the degree of
Master of Science
Graduate Program in Civil & Environmental Engineering
Written under the direction of
Husam Najm
And approved by
_____________________________________
_____________________________________
_____________________________________
New Brunswick, New Jersey
October 2017
ii
ABSTRACT OF THE THESIS
SEISMIC VULNERABILITY ASSESSMENT OF BRIDGES FOR RETROFITTING AND
NEW DESIGN
By PEDRAM FAROKH
Thesis Director:
Dr. Husam Najm
Many bridges in North Eastern region of U.S. were designed prior to the adoption of the
AASHTO LRFD Guide Specifications for seismic design and may be vulnerable to damage
during an earthquake event. This study evaluates the seismic vulnerabilities of those bridges
and the structural factors that could affect their performance during a seismic event. The
effects of load demands and age deteriorations were also studied. Aging of certain bridge
components such as bearings, columns, and bent caps can affect the capacity and demands
of these components and accordingly might affect the global behavior and capacity of a
bridge during an earthquake event. The concept of fragility curves was studied as a
potential tool for evaluating the seismic performance of new bridges, existing bridges and
retrofitted bridges for various bridge types subjected to different peak ground acceleration
levels. Fragility curves represent the probability of a structure to experience damage levels
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higher than specific damage state at different peak ground acceleration. Possible retrofit
measures for various bridge components were reviewed, and analyzed for their
effectiveness. These include superstructure restrainers, stoppers, shear keys, isolation
bearings, bent cap strengthening and column jacketing. Existing research shows that the
concept of fragility curves can be used to identify bridge vulnerability and level of damage.
They can also be used to identify performance and level of damage of various retrofit
measures. The effect of aging of certain components such as stiffening and locking of
bearings and corrosion of confining steel in columns need to be included when evaluating
bridge load demands and capacities.
Different types of concrete bridges (typical in North Eastern United States) were analyzed
using elastic response spectrum and nonlinear push-over analysis for low, medium-to-high,
and high seismicity levels. The effects of pier configuration, continuity between the
superstructure and the substructure, and the number of spans were investigated.
Analysis results showed that in the longitudinal direction, the displacement demand
increased for multi-column bents compared to single-column bents. However, the overall
D/C ratio dropped in both transverse and longitudinal directions. The results also showed
that in the longitudinal direction the benefit of having multi-column bent over single-
column bents in integral bridges is dependent on the seismicity levels.
The D/C (demand/capacity) ratio for single column bents in the longitudinal direction was
much lower for integral (monolithic) bents compared to non-integral bents. In the
transverse directions, the difference in the D/C ratio was not significant. For multi-column
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bents, the percent change by having integral bents over non-integral bents was dependent
on the seismicity levels. For high seismicity zones, the benefits of having Integral bents
becomes more significant.
This investigation presents guidance on incorporating the effects of aging and retrofitting in
the finite element modeling of bridges subjected to various levels of earthquake ground
motions.
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ACKNOWLEDGEMENTS
I would first like to thank my thesis advisor Dr. Husam Najm at Rutgers University. It has
been an honor working with him and I am so grateful for his patience and his support
throughout my graduate study. He has been always supportive and available for questions.
I would also like to thank my committee members Dr. Perumalsamy Balaguru and Dr. Jie
Gong for their support and review comments on my thesis.
I would like to thank the Center for Advanced Infrastructure and Transportation (CAIT) at
Rutgers University, in particular, CAIT Director Dr. Ali Maher for providing access to
resources such as Long Term Bridge Performance Portal (LTBP) and his support throughout
my study. I would also like to thank Dr. Nenad Gucunski, the chairman of Civil and
Environmental Engineering Department for his support during the course of my study.
Finally, I want to express my very profound gratitude to my parents for providing me with
unfailing support and continuous encouragement throughout my years of study. This
accomplishment would not have been possible without them.
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TABLE OF CONTENTS
ABSTRACT OF THE THESIS......................................................................................................................... II
ACKNOWLEDGEMENTS ............................................................................................................................ V
INTRODUCTION ........................................................................................................................................ 1
1.1. PROBLEM STATEMENT ......................................................................................................................... 1
1.2. REVIEW OF FHWA PROVISIONS FOR SEISMIC RETROFIT ....................................................................... 3
1.2.1. FHWA Performance Levels and Earthquake Retrofit Levels .......................................................... 7
1.2.2. Bridge Importance and Anticipated Service Life ........................................................................... 8
1.2.3. Performance-Based Seismic Retrofit Categories......................................................................... 10
1.2.4. Retrofit Design Approach for Lower and Upper Levels ............................................................... 12
1.2.5. FHWA Methods of Evaluation ..................................................................................................... 13
1.2.6. Retrofit Strategies, Approaches, and Measures ......................................................................... 15
1.2.7. Seismic Retrofit Category (SRC) Versus Seismic Design Category (SDC) ..................................... 18
1.3. REVIEW OF SEISMIC FRAGILITY CURVES ............................................................................................. 19
1.3.1. APPLICATION OF FRAGILITY CURVES .......................................................................................... 21
1.3.2. EVALUATING AND SELECTING RETROFIT MEASURES ................................................................. 22
1.3.3. COST –BENEFIT ANALYSIS ........................................................................................................... 23
1.3.3.1. Example ............................................................................................................................................... 23
1.3.4. APPLICATION OF FRAGILITY CURVES FOR SEISMIC VULNERABILITY of VARIOUS BRIDGE TYPES 25
1.3.5. CASE STUDY: USE OF LTBP PORTAL for IDENTIFYING SEISMICALLY VULNEABLE BRIDGES IN NEW
JERSEY ................................................................................................................................................... 26
EFFECT OF AGING ON SEISMIC CAPACITY AND FRAGILITY OF BRIDGES .................................................. 28
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2.1. CORROSION DETERIORATION OF RC MEMBERS ................................................................................. 32
2.2. ELASTOMERIC BEARING PADS THERMAL OXIDATION ........................................................................ 36
2.3. STEEL FIXED AND EXPANSION BEARINGS ........................................................................................... 37
2.3. FIXED AND EXPANSION ELASTOMERIC BEARINGS .............................................................................. 39
2.4. BENTS AND RC COLUMNS ................................................................................................................... 40
2.5. FRAGILITY CURVES BASED ON AGING ................................................................................................. 40
2.6. AGING IN RC COLUMNS ...................................................................................................................... 41
2.7. AGING OF BEARINGS .......................................................................................................................... 42
2.8. COMBINED EFFECTS OF AGING IN BEARINGS AND COLUMNS............................................................ 43
SEISMCI VULNERABILITY RISK ASSESSMENT AND COST-BENEFIT ANALYSIS ........................................... 45
3.1. LIFE CYCLE COST-BENEFIT ................................................................................................................... 45
3.2. REPLACEMENT COST .......................................................................................................................... 48
3.3. FRAGILITY CURVES OF RETROFITTED BRIDGES ................................................................................... 48
3.4. NETWORK BASED SEISMIC RISK ANALYSIS USING REDARS2 ............................................................... 49
SEISMIC RETROFIT MEASURES ............................................................................................................... 50
4.1. SUPER-STRUCTURE RETROFIT MEASURES .......................................................................................... 50
4.2. SUBSTRUCTURE RETROFIT MEASURES ............................................................................................... 56
4.3. SEISMIC RESPONSE MODIFICATION MEASURES ................................................................................. 61
4.3.1. Seismic Isolation ......................................................................................................................... 62
Isolation Elastomeric Bearings .......................................................................................................................... 63
Lead-Rubber Bearings....................................................................................................................................... 64
Concave Friction Pendulum Bearings ............................................................................................................... 65
4.3.2. Damping Devices ........................................................................................................................ 66
Shock Transmission Lock-Up Devises ............................................................................................................... 66
Viscous Damper ................................................................................................................................................ 66
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ANALYTICAL INVESTIGATION OF SEISMIC VULNERABILITY OF BRIDGES ................................................. 67
5.1. INTRODUCTION TO ANALYTICAL STUDY ............................................................................................. 67
5.2. SEISMICITY LEVELS .............................................................................................................................. 70
5.3. MATERIALS ......................................................................................................................................... 73
5.4. FINITE ELEMENT MODELING .............................................................................................................. 74
5.5. ANALYSIS ............................................................................................................................................ 75
5.6. RESULTS AND DISCUSSION ................................................................................................................. 77
5.7. SIMULATION OF AGING EFFECTS AND RETROFIT OF BRIDGE COMPONENTS ..................................... 83
CONCLUSION AND RECOMMENDATIONS ............................................................................................... 90
6.1. CONCLUSIONS ........................................................................................................................................ 90
6.2. RECOMMENDATIONS ............................................................................................................................... 91
REFERENCES ........................................................................................................................................... 93
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LIST OF FIGURES
Figure 1: Effect of Return Period on Seismic Acceleration in Northern New Jersey................. 3
Figure 2: Seismic retrofitting process (FHWA Manual, 2006) ................................................... 5
Figure 3: Steps for determining SRC (FHWA Manual, 2006) ................................................... 10
Figure 4: Detailed seismic retrofitting process (FHWA Manual, 2006) ................................... 17
Figure 5: Sample fragility curves for different damage states ................................................ 21
Figure 6: Seismic event time-line (Basoz and Kiremidjian, 1996) ........................................... 21
Figure 7: Application of fragility curves in determining effectiveness of retrofit measures
(DesRoches 2008) .................................................................................................................... 22
Figure 8: Fragility curves for the example bridge (FHWA Manual, 2006) ............................... 24
Figure 9: Comparison of bridge types' seismic vulnerability (DesRoches 2008) ..................... 25
Figure 10: Number of bridges built in NJ over time ................................................................ 26
Figure 11: Number of bridges based on material and construction type in NJ ...................... 27
Figure 12: Number of bridges based on type and year built in NJ .......................................... 27
Figure 13: Corrosion deterioration of RC Columns [adopted from Lower (2010)] ................. 29
Figure 14: Debris Accumulation and formation of Corrosion at Bearing [Lindquist (2008)] .. 29
Figure 15: Environmental exposure cases (Ghosh 2013) ........................................................ 33
Figure 16: Corrosion initiation time for a) splash and tidal zone, b) atmospheric zone (Guo,
Yuan, Lan, Guan, Li 2014) ........................................................................................................ 34
Figure 17: Section loss of anchor bolt due to corrosion deterioration (Lindquist, 2008) ....... 38
Figure 18: Ultimate lateral strength of fixed bearing over the time due to anchor bolts' area
loss (Ghosh and Padgett 2010) ................................................................................................ 41
Figure 19: Reduction in column load resisting capacity and yield curvature (Ghosh and
Padgett 2010) .......................................................................................................................... 42
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Figure 20: Ductility demand of 50 year old bridge vs. pristine bridge (Ghosh and Padgett
2010) ........................................................................................................................................ 42
Figure 21: System level time-dependent seismic fragility curves corresponding to different
damage states for the case study MSC steel girder bridge (Ghosh and Padgett 2010) .......... 43
Figure 22: Comparison of fragility curves for MSSS Concrete Bridge for complete damage
state under different exposure conditions (Ghosh and Padgett 2012) .................................. 44
Figure 23: Cost-benefit analysis procedure (ODOT 2015) ....................................................... 47
Figure 24: Concrete block & steel bracket seat extenders (DesRoches 2008) ........................ 51
Figure 25: Seat extenders: steel bracket, steel beam and extended seating frame
(DesRoches 2008) .................................................................................................................... 51
Figure 26: Catcher block installed at the location of a tall bearing (DesRoches 2008) ........... 52
Figure 27: Restrainer cables connected through the bent cap and restraining cables
connected directly to the adjacent girder (DesRoches 2008) ................................................. 52
Figure 28: Restrainer cables applications (DesRoches 2008) .................................................. 53
Figure 29: Restraining cables connecting girders to column/connecting girder to girder over
the pier (DesRoches 2008) ...................................................................................................... 54
Figure 30: Restraining cables used to restrain movement in both directions (DesRoches
2008) ........................................................................................................................................ 54
Figure 31: Concrete block shear key (DesRoches 2008) ......................................................... 55
Figure 32: Keeper bracket shear key (DesRoches 2008) ......................................................... 55
Figure 33: Bumpers/stoppers applications (DesRoches 2008) ............................................... 55
Figure 34: Bent cap retrofit: external post tensioning (DesRoches 2008) .............................. 56
Figure 35: Bent cap retrofit: concrete bolster or jacketing (DesRoches 2008) ....................... 56
Figure 36: Bent cap retrofit: steel jacketing (DesRoches 2008) .............................................. 57
Figure 37: Bent cap retrofit: external shear reinforcement and confinement (DesRoches
2008) ........................................................................................................................................ 57
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Figure 38: Partial and full height steel jacketing (DesRoches 2008) ....................................... 58
Figure 39: Typical details of steel jacketed column (DesRoches 2008)................................... 58
Figure 40: Application of concrete overlay retrofit (DesRoches 2008) ................................... 59
Figure 41: Composite wrap column retrofit (James E.Roberts 2005) ..................................... 60
Figure 42: Seismic isolation silos for increasing elastic length of short column (FIB Bulletin
39) ............................................................................................................................................ 61
Figure 43: Examples of vulnerable bearings (DesRoches 2008) .............................................. 61
Figure 44: Seismic force reduction due to the increase in the period (Buckle 2016) ............. 62
Figure 45: Elastomeric bearing (DesRoches 2008) .................................................................. 63
Figure 46: Deformed elastomeric bearing due to a lateral Load (DesRoches 2008) .............. 63
Figure 47: Lead-rubber bearing ............................................................................................... 64
Figure 48: Effect of damping on acceleration and displacement spectra when an isolation
system is used (DesRoches 2008) ............................................................................................ 64
Figure 49: Concave friction pendulum bearing (Buckle 2016) ................................................ 65
Figure 50: Shock transmission lock-Up devices ....................................................................... 66
Figure 51: Viscous damper ...................................................................................................... 66
Figure 52: Bulb Tee Girders Maximum Span length Vs. Spacing (PCI 2011) ........................... 69
Figure 53: Elevation of a Two Span Bridge .............................................................................. 69
Figure 54: Elevation of a Three Span Bridge ........................................................................... 69
Figure 55: Bent Cross-Sections a) Single-Column Bent, b) Multi-Column Bent ...................... 70
Figure 56: Column Cross-Sections and Reinforcements a) 5' Diameter Column, b) 3'
Diameter Column .................................................................................................................... 70
Figure 57: Generated Response Spectrum Curve for Low Seismicity Regions........................ 72
Figure 58: Generated Response Spectrum Curve for Medium Seismicity Regions ................ 72
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Figure 59: Generated Response Spectrum Curve for Medium to High Seismicity Regions .... 73
Figure 60: Material models for confined and unconfined concrete (CSI, 2017) ..................... 75
Figure 61: Moment Curvature Curves of a) 3' Diameter Column, b) 5' Diameter column ..... 80
Figure 62: Screen shot showing time dependent concrete model selection (CSI, 2017) ....... 84
Figure 63: Screen shot showing restrainer selection (CSI, 2017) ............................................ 86
Figure 64: Screen shot showing link element selection (CSI, 2017) ........................................ 88
Figure 65: Screen shot showing selection menu for modeling of column casing (CSI, 2017) . 89
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LIST OF TABLES
Table 1: Expected performance for each level of earthquake (FHWA Manual, 2006) 7
Table 2: Minimum performance levels for retrofitted bridges (FHWA Manual, 2006) 11
Table 3: Performance-based seismic retrofit categories (FHWA Manual, 2006) 12
Table 4: Evaluation methods for existing bridges 14
Table 5: Minimum requirements (FHWA Manual, 2006) 16
Table 6: SRC vs SDC- based on an example at a Zip-Code in NJ (Anil K. Agrawal, H. L. 2012). 18
Table 7: Characteristics of the bridge used as an example (FHWA Manual, 2006) 23
Table 8: Replacement costs for each damage state for the example bridge 24
Table 9: Mechanisms of degradation on bridge components (Ghosh, 2013) 32
Table 10. Corrosion rate for deicing salt exposure (Enright and Frangopol 1998) 35
Table 11: Retrofit measure's cost estimate (Padgett, DesRoches) 48
Table 12: Bridge Elements Dimension 68
Table 13: US Department of Veteran Affairs Seismicity Levels 71
Table 14: Seismicity Levels Used for the Study 71
Table 15: Material Properties Used for the Study 73
Table 16: Analysis Results for Single-Column Bents 77
Table 17: Analysis Results for Multi-Column Bents 78
Table 18: Comparison of Multi-Column Bents vs. Single-Column 79
Table 19: Comparison Single-Col Integral Bent Vs. Single-Col Non-Integral 82
Table 20: Comparison Multi-Col Integral Bent Vs. Multi-Col Non-Integral 82
1
CHAPTER I
INTRODUCTION
1.1. PROBLEM STATEMENT
Many bridges in U.S were designed and built before adoption of the AASHTO Guide
Specifications for LRFD Seismic Bridge Design. As an example, according to the LTBP portal,
441,705 bridges or 72% of all bridges in United States were designed prior to the adoption
of the first AASHTO seismic design requirements (1990) with average age of 46 years. In
New Jersey, there are 5,358 Bridges (80% of all the bridges in the state) that were designed
before adoption of the AASHTO seismic design requirement with average age of 45.8 years.
These bridges are vulnerable in the event of moderate to high intensity seismic event and
need to be identified and evaluated for potential risks, feasibility of repair, as well as
consequences of not addressing their vulnerabilities. The objective of this study is to
2
investigate seismic load demands on those vulnerable bridges, their capacities, as well as
their performance criteria including the effects of aging of their elements. The study will
review the various retrofit measures being used in seismic retrofit and upgrade of bridges.
The research includes a literature review of the FHWA Seismic Retrofit Manual
(2006).Retrofit measures for various bridge components such as superstructure restrainers,
stoppers, shear keys, isolation bearings, bent cap strengthening and column jacketing will
be reviewed for their effectiveness and how to incorporate them in computer models.
Previous research on fragility curves identifies bridges and retrofit measures that are more
vulnerable than others when subjected to seismic events. For certain, bridge types that are
not in the database, fragility curves can be developed to evaluate their seismic vulnerability
including the effects of aging of certain bridge components such as bearings, bent caps, and
columns.
AASHTO adapted 1000 years return period in 2007 instead of 500 years. The new return
period imposes higher demands on the structures. So in order to investigate the effect of
the change in the return period, a case study was done at a location in northern part of New
Jersey. Response spectra were developed based on 500 years and 100 years return period
for different soil types. As shown in Figure 1, 1000 years return period imposes higher
demand at the short period range comparing to 500 years return period and the difference
becomes more significant as the soil type goes toward soil type E.
3
Figure 1: Effect of Return Period on Seismic Acceleration in Northern New Jersey
1.2. REVIEW OF FHWA PROVISIONS FOR SEISMIC
RETROFIT
Based on the intensity of damage, bridges may need to be upgraded for potential seismic
event that they were not designed for, to minimize damage or potential collapse. This starts
by identifying those bridges that are at risk and later evaluate collapse vulnerability of those
bridges followed by making a decision about mitigating the seismic risk. Retrofit decisions
are based on several factors that includes the importance, age of the bridge, and feasibility
of the retrofitting. However, in some cases, the cost to replace the bridge or do nothing and
accept the damage may be the options that would worth considering and adopting. The San
Fernando earthquake 1971 was one of the most important earthquakes, which proved that
bridges build prior to that time were not seismically adequate to stand earthquakes, so it
4
was the starting time in attempting to perform seismic retrofit on bridges. Even though, the
mode of failure might be different based on the type of the bridge, design consideration and
also location of the bridge, most of the failures in bridges during seismic events happened
due to unseating of the spans. However, the FHWA addresses potential failures in different
components of bridges and accordingly a retrofit measure or technique can be
implemented to mitigate that type of failure. So in another word, FHWA manual can be
used to evaluate and also upgrade seismic resistance of steel and concrete bridge types. So
the use of the manual can be summarized in the following three steps. The first process is
screening and prioritizing bridges that are in need of seismic retrofitting, due to limited
resources and funds. Second, evaluating seismic resistance and capacity of those bridges
quantitatively and provide a methodology for determining effectiveness of different retrofit
measures or techniques, and lastly choosing retrofit approaches and associated techniques
of increasing resistance of those bridges. This process is shown in Figure 2.
5
Figure 2: Seismic retrofitting process (FHWA Manual, 2006)
In the past, the design for earthquake was based on single level of earthquake ground
motion event, which was the largest excepted motion during the life of the bridge. It implies
that ground motions larger than the design earthquake might also happen during the life of
the bridge but it has a low probability. This probability is expressed either as probability of
exceedance or as return periods. This method compared to using of the maximum historical
event for each area is more rational and reasonable. The standard Specification for Highway
bridges by AASHTO published in 2002, adopted this approach considering 10 percent
probability of exceedance in 50 years, which is same as 500 years return period. However,
later, AASHTO adopted 1000 years return period in 2007 which represents a probability of
6
7% exceedance in 75 years. The FHWA manual also adopted the 1000 years return period
for the upper level analysis. It has been also considered in the code that structures should
be designed to resist small to moderate earthquakes within the elastic range without having
significant damage. In addition, ground motion intensities for the design should be more
realistic, and large earthquakes should not cause collapse in the structure and the
associated damages should be easily detectable for inspection and repair.
It has been assumed earlier that a single-level earthquake for design and retrofit of bridges
is satisfactory which means that if a structure is designed or retrofitted for the design
earthquake, the performance of the structure should be satisfactory under any other levels
of ground motions. This assumption has been proved to be not true after the recent
earthquakes in California, Costa Rica, Japan, Turkey and Taiwan. Such large earthquakes in
the country can happen to be three to four times stronger than the design earthquake and
can cause failure such as collapse and instability, but their occurrence has a lower
probability of occurrence which should be explicitly considered in the design. That is why
Multi-level design should be used in designing or retrofitting of bridges instead of a single
level earthquake. So there will be different expectation in the performance of structures at
each level of intensity, such that in smaller ground motions a higher level of performance is
expected while the expected performance would be lesser for higher level earthquakes. So
performance-based design has been adopted which allows for different performance
expectations for bridges of varying importance while subjected to different levels of seismic
hazard. The expected performance for each level of earthquake is shown in the Table 1.
7
Table 1: Expected performance for each level of earthquake (FHWA Manual, 2006)
1.2.1. FHWA Performance Levels and Earthquake Retrofit Levels
FHWA uses four performance levels for specifying the performance Criteria’s which are
based on Bridge Importance and also anticipated service life (ASL) of the bridge. These
performance levels definitions are copied from the manual and presented in Table 2. The
FHWA manual has adopted two levels of earthquake for retrofitting of bridge structures for
seismic events. These earthquake levels along with their return period and probability of
exceedance are: 1) Lower Level earthquake (LL) or 100-year return period (50% probability
of exceedance in 75 years), and 2) Upper Level earthquake (UL), or 1000-year return period
(7% probability of exceedance in 75 years). The LL earthquake is applied to assess bridge
performance for small to moderate earthquakes making sure that such ground motions will
be resisted within the elastic range and there will be no significant structural damage.
Under the UL, no collapse should occur. According to the FHWA manual, bridges that fall
under Seismic Retrofit Category A (Seismic Retrofit Categories are discussed later in this
Chapter), will be exempt from upper level earthquake motion. However, bridges that satisfy
any of the following criteria, will be exempt for both upper and lower motions:
8
Bridges with anticipated service life of less than 15 years
Temporary Bridges (with ASL of 15 years or less)
Closed Bridges to traffic which do not pass over an active highway, waterway or rail
road.
1.2.2. Bridge Importance and Anticipated Service Life
FHWA manual considers two types of Importance for bridges and other than factors such as
detour lengths and traffic counts, factors such as socio-economics, societal/survival and
security/defense should be also considered. The manual classifies bridges into Essential and
standard bridges, with the Essential bridges being expected to remain functional
immediately after a seismic event and those that cross other routs which are expected to
remain open after a seismic event. It classifies other bridges as standard. Based on the
above definition, an essential bridge is a bridge which has at least one of the following
conditions:
A bridge that is required to provide secondary life safety; e.g., one that provides
access to local emergency services such as hospitals. This category also includes
those bridges that cross routes that provide secondary life safety, and bridges that
carry lifelines such as electric power and water supply pipelines.
A bridge whose loss would create a major economic impact; e.g., one that serves as
a major link in a transportation system, or one that is essential for the economic
recovery of the affected region.
A bridge that is formally defined by a local emergency plan as critical; e.g., one that
enables civil defense, fire departments, and public health agencies to respond
9
immediately to disaster situations. This category also includes those bridges that
cross routes that are defined as critical in a local emergency response plan and
those that are located on identified evacuation routes.
A bridge that serves as a critical link in the security and/or defense roadway
network.
An importance factor for deciding on retrofitting of bridges, is the anticipated service life of
the bridge. The ASL can be related to the remaining service life of the bridge or in another
word the age of the bridge. AASHTO LRFD specification adopted 75 years as the service life
of a new bridge. However, some of the long span bridges and some of the bridges in low-
density areas might have higher service life. So FHWA manual divided bridges based on their
remaining service life into three categories as shown in Table 2. Bridges falling under ASL 1
are not economically justified for retrofitting since they are near the end of their service life.
The purpose of incorporating the service life into retrofitting process is due to the fact that
retrofitting a bridge with a short remaining life might not be economical and in addition it
would be really unlikely that the design earthquake can happen during that period. The
opposite is true for bridges with higher ASL or the bridges which are almost new. These
bridges should be retrofitted for a longer service life. It should be taken into consideration
that rehabilitation of bridges due to other factors such as aging etc. will extend the
anticipated service life of the bridge, as an example it might lift the bridge from ASL 1 to ASL
2 which means that this time the bridge should be considered for seismic evaluation. In case
of other rehabilitation being implemented on the bridge, an advantage can be taken of the
10
contractor being on the site to reduce the cost by implementing the seismic retrofit also at
the same time if needed.
1.2.3. Performance-Based Seismic Retrofit Categories
As shown in Figure 3, once the performance levels and seismic Hazard levels are
determined, Seismic Retrofit Categories (SRC) can be obtained.
Figure 3: Steps for determining SRC (FHWA Manual, 2006)
11
Table 2: Minimum performance levels for retrofitted bridges (FHWA Manual, 2006)
(SRC) are used for determining minimum level of screening, evaluation and retrofitting
required for a certain bridge. Required performance levels will be satisfied once these
12
minima are satisfied. As shown in Table 3, Seismic Retrofit categories are categorized from A
through D with the A being the default which does not need any screening, evaluation or
retrofitting and with D being the highest category which needs to be screened, evaluated
and retrofitted more rigorously.
Table 3: Performance-based seismic retrofit categories (FHWA Manual, 2006)
1.2.4. Retrofit Design Approach for Lower and Upper Levels
As mentioned earlier, a return period of 100 years should be considered for the lower level
retrofitting. The performance of the bridge should be elastic. The approach to be used for
the lower level is force-based and an assumption is made that the displacements within the
capacity of the structure are small. Screening and prioritization for the lower level is done
by comparing the earthquake load with wind and braking loads. The retrofit design process
for the upper level can be categorized into screening and prioritization, detailed evaluation,
determination of retrofit strategy, approaches and measures. The purpose of screening and
prioritization is to screen the whole inventory of the bridges and find the bridges with
seismic deficiencies and then based on the vulnerability; hazard and other non-structural
factors prioritize them for seismic retrofitting. This process is fast and conservative so the
13
bridges that fail should be screened further under detailed evaluation. There are three
methods available for screening and prioritization, with the first one being the most
conservative while the easiest, and with the last one being the less conservative while being
more complex. These methods include: 1) the Indices Method (FHWA, 1995); 2) the
Expected Damage Method; and 3) the Seismic Risk Assessment Method. The risk
assessment method uses fragility curves which will be discussed later in more details.
1.2.5. FHWA Methods of Evaluation
There are six different methods proposed in the FHWA manual for evaluation of bridges. In
general all of these methods involve demand analysis, capacity assessment and
capacity/demand ratio calculation either for each critical component or for the bridge as a
system. These methods are shown with their details in Table 4.
14
Table 4: Evaluation methods for existing bridges
15
1.2.6. Retrofit Strategies, Approaches, and Measures
A retrofit measure can be a technique or a device used to physically modify a component in
a bridge for upgrading its seismic performance. These techniques could be column
jacketing, assigning restrainers and so on. Retrofit approach is the philosophy adopted to
improve seismic performance of a bridge. Strengthening is a retrofit approach, which can be
implemented by one or more measures together to improve the performance. The overall
plan for the seismic retrofit of a bridge is called strategy which might use one or more
approaches together thus it might be a combination of several different measures.
Strategies could be partial or full replacement or even do-nothing (incase retrofitting is not
justified).Some of the retrofit approaches are listed below:
Strengthening
Displacement capacity enhancement
Force limitation
Response modification
Site remediation
Partial replacement
Damage acceptance or control
Any of these approaches can be accomplished through the use of one or more retrofit
measures. Different retrofit measures are discussed later in Chapter 4. Based on the seismic
retrofit category determined earlier, bridges are subjected to minimum of screening,
evaluation and retrofitting. These minimum requirements are presented in Table 5 for both
lower and upper level earthquakes. The complete seismic retrofitting process is shown in
16
Figure 4.
Table 5: Minimum requirements (FHWA Manual, 2006)
.
17
Figure 4: Detailed seismic retrofitting process (FHWA Manual, 2006)
18
1.2.7. Seismic Retrofit Category (SRC) Versus Seismic Design Category (SDC)
FHWA manual has guidelines for seismic retrofitting of bridges considering two earthquake
level. Bridges are expected to remain elastic during the lower level one and collapse should
be prevented when using the upper level earthquake.
Seismic Design category based on AASHTO-SGS for new bridge design is based on SD1 while
the FHWA retrofit manual uses both SDS and SD1 in determining the seismic retrofit category,
which might impose higher requirements on retrofitting of existing bridges comparing to
design of new bridges. For a Zip code 07022 due to a 1000 years earthquake spectra,
different Hazard levels were distinguished while based on AASHTO, all locations have been
considered as Seismic category A for new design as shown in Table 6 (Anil K. Agrawal, H. L.
2012).
Table 6: SRC vs SDC- based on an example at a Zip-Code in NJ (Anil K. Agrawal, H. L. 2012).
19
1.3. REVIEW OF SEISMIC FRAGILITY CURVES
The probability of a structure to face a damage equal and beyond a specific damage state at
different ground shaking levels can be represented by fragility curves. The fragility of a
structure can be defined as conditional probability of failure of that structure due to a given
seismic response parameter such as peak ground acceleration (PGA), spectral acceleration,
etc. One of the most common application and use of fragility curves is to estimate the PGA
value at which the structure’s capacity is not sufficient for the seismic response of the
structure, which results in failure. Different methods can be used to generate fragility
curves, such as empirical methods, expert opinions or by the use of analytical methods. Past
earthquake data especially those from Loma Prieta (1989), Northridge (1994) and Kobe
(1995) can be utilized to develop empirical fragility curves. Comparing to other methods,
empirical method is comparatively straightforward.
Expert opinion based fragility curves, can be developed by expert’s opinion and the
structural damages from an earthquake are estimated by experts’ collected opinion. This
method is subjective since it is based on the experts’ opinion. Damage probability matrix is
used to describe damage state for various ground motion levels based on the survey results.
Numerical simulations of bridges’ structural response due to ground motions for a certain
type of bridge can be also used to generate fragility curves. There are different numerical
methods such as Elastic Spectral Analysis, Non-linear Static Analysis and Non-linear Time
History Analysis. Under Non-linear Static Analysis, Mander and Basoz used a capacity-
spectrum approach for developing fragility curves. Empirically derived fragility curves based
20
on bridge damages of 1989 Loma Prieta and 1994 Northridge earthquakes were used to
validate their rapid analysis procedure and later FHWA seismic retrofit manual also adopted
a similar procedure for developing fragility curves. Analytical methods are the most
common and diverse method of developing fragility curves especially in areas where there
are not sufficient seismic damage data available such as in the Eastern part of the country.
The first step in use and development of fragility curves is the clear definition of failure for
the concerned bridges. Based on the analysis goals, the failure definition might vary. For
instance, failure can be defined as loss of function, strength, integrity, etc. According to
FHWA seismic retrofit manual (2006), there are five damage states as presented below:
• DS1 = no damage (pre-yield)
• DS2 = slight damage
• DS3 = moderate damage
• DS4 = extensive damage
• DS5 = collapse
A sample fragility curves for different damage states is shown in Figure 5.
21
Figure 5: Sample fragility curves for different damage states
1.3.1. APPLICATION OF FRAGILITY CURVES
Fragility curves can be used for assessing seismic performance of bridges. Randomness of
shaking intensity and the return period, make earthquakes of the most unpredictable
natural hazards. On the other hand, bridges are of the most important elements in highway
systems, while previous earthquakes such as 1971 San Fernando, 1994 Northridge etc.
proved that bridges are also one of the most vulnerable components of highway systems
and they need special attentions during the whole seismic event time-line as presented in
Figure 6.
Figure 6: Seismic event time-line (Basoz and Kiremidjian, 1996)
22
The first item on the seismic event time-line is the Risk assessment which is interrelated to
the rest of the events. Fragility curves are efficient tools for seismic risk assessment and
retrofit screening as mentioned in FHWA under EXPECTED DAMAGE METHOD. Fragility
curves can be further used in natural hazard risk assessment software such as HAZUS and
REDARDS2 for performing network analysis for regional seismic risk assessment.
1.3.2. EVALUATING AND SELECTING RETROFIT MEASURES
Fragility curves can also be used in evaluating seismic retrofit measures and selecting the
most viable measure based on median value improvement and performance objective.
Figure 7 shows fragility curves developed for different retrofit measures on Multi-Span
Simply Supported (MSSS) steel bridge for slight damage state and shows how effective each
retrofit measure can be in reducing the failure probability. (DesRoches 2008).
Figure 7: Application of fragility curves in determining effectiveness of retrofit measures
(DesRoches 2008)
23
1.3.3. COST –BENEFIT ANALYSIS
Fragility curves can also be used in the cost benefit analysis for determining the best retrofit
measure or whether to perform retrofit or not by comparing fragility curves based on
current un-retrofitted structure and the fragility curves developed for retrofitted bridge. In
general the loss in both cases can be determined by calculating replacement Cost Ratio
(RCP) and multiplying it by the estimated replacement cost. This process will be further
explained later.
1.3.3.1. Example
A simple span pre-stressed concrete bridge having characteristics shown in Table 7, has
been used to estimate losses at a given spectral acceleration (0.224 g).
Table 7: Characteristics of the bridge used as an example (FHWA Manual, 2006)
The developed fragility curves for the mentioned bridge is shown in Figure 8 for all the
damage states, and a vertical line has been constructed at the spectral acceleration under
consideration ( 0.224 g) to intersect all the four curves in order to read the probability
associated with each of them.
24
Figure 8: Fragility curves for the example bridge (FHWA Manual, 2006)
The total Replacement cost Ratio (RCPT) is calculated as shown in Table 8. So the Loss under
that spectral acceleration will be equal to replacement cost (estimated as $616,000 based
on the deck area and unit construction price available in the market) multiplied by the RCPT
which leaves the loss value of $138,046.
Table 8: Replacement costs for each damage state for the example bridge
25
1.3.4. APPLICATION OF FRAGILITY CURVES FOR SEISMIC VULNERABILITY of VARIOUS BRIDGE TYPES
Based on the studies available in literature, seismic vulnerability of bridge types based on
fragility curves are shown in Figure 9 (DesRoches 2008). It shows that Multi Span
Continuous Steel brides (MSC) and Multi-Span Simply Supported Steel bridges (MSSS) are
among the most vulnerable bridges and the MSSS Concrete-Box bridges are among the least
vulnerable bridges during a seismic event (DesRoches 2008). Fragility curves database in the
literature can be used to evaluate the seismic vulnerability of many bridges in the North
Eastern region of the country. For other bridges, fragility curves can be developed for those
bridge categories that are not available in the literature.
Figure 9: Comparison of bridge types' seismic vulnerability (DesRoches 2008)
26
1.3.5. CASE STUDY: USE OF LTBP PORTAL for IDENTIFYING SEISMICALLY VULNEABLE BRIDGES IN NEW JERSEY
The LTBP portal includes a comprehensive database of quantitative data on bridge
performance in the United States. It can be used to analyze and apply the data gathered to,
facilitate improved life-cycle cost and predictive models, better understanding of bridge
deterioration, and more effective maintenance and repair strategies. In this study the LTBP
portal was use to provide data on seismic vulnerability of bridges in New Jersey as an
example. It includes information on bridge age, year designed and upgraded, bridge type
and location. As discussed earlier many of the bridges in NJ have been constructed before
implementation of seismic design code as it is shown in Figure 10. The LTBP Portal can
provide data on the number of these bridges that were designed before implementation of
these design guidelines.
Figure 10: Number of bridges built in NJ over time
Figure 11 also shows that how many of each type of bridges are there in NJ. A combination
of the bridge type and the year it was built is shown in Figure 12. This Figure is a better
0
500
1000
1500
2000
2500
3000
Number of Bridges vs Year Built
27
representation to monitor the vulnerability of bridges in New Jersey. The LTBP can also
provide information on the location and traffic data on these bridges that can help in
preliminary categorization of these bridges as ‘standard’ or ‘critical’. It also can provide
information on the potential impact of aging on those bridges based on their location and
climatic conditions.
Figure 11: Number of bridges based on material and construction type in NJ
Figure 12: Number of bridges based on type and year built in NJ
0
500
1000
1500
2000
2500
3000
3500
NJ Bridge Types
0 500 1000 1500 2000 2500 3000 3500
0-Other
1-Concrete
2-Concrete continuous
3-Steel
4-Steel continuous
5-Prestressed concrete *
6-Prestressed concrete continuous *
7-Wood or timber
8-Masonry
9-Aluminum, Wrought Iron or Cast Iron
NJ Bridge Types vs Year Built
1792-1814 1815-1837 1838-1860 1861-1883 1884-1906
1907-1929 1930-1952 1953-1975 1976-1998 1999-2021
28
CHAPTER II
EFFECT OF AGING ON SEISMIC CAPACITY AND
FRAGILITY OF BRIDGES
Transportation network across the country consisted from different components, of which,
bridges are considered key components of it. At the same time, all these bridges are
extensively deteriorating due to different factors such as natural hazards, heightened traffic
loads and adverse environmental conditions. As an example, most of the bridges in the
state of New Jersey were constructed before implementation of seismic design code (1990)
which means that these bridges are not designed according to seismic standards. The
average age of New Jersey Bridges is 39.2 years, so these bridges are more vulnerable to
earthquake not only due the lack of adequate seismic detailing and design, but also due to
the aging deterioration they have faced over time. As mentioned earlier, fragility curves
have been used to quantify the probability of a bridge under given seismic intensity to reach
29
or exceed a certain damage level. Unfortunately, most of the attention so far has been given
to pristine bridges or retrofitted bridges. However, in many cases the concept of fragility has
not been considered in conjunction with the aging phenomena. This is due to significant
number of simulation needed to include this aspect. Repeated seismic events can also cause
reduction in the strength of the bridge and making it weaker to stand future seismic events.
The research done by Elnashai et al. showed that ductility demand of a structure under
multiple seismic events is often several time higher that the ductility demand on the same
bridge under single seismic event. Structural aging and degradation can be in the form of
corrosion of steel reinforcements, concrete spalling in RC members, accumulation of debris
causing “freezing” of steel bearings, loss of steel area in anchor bolts at the bearings,
stiffening of elastomeric bearing pads, loss of section in steel piles, soil scour, thermal
oxidation, soil erosion etc.
Figure 13: Corrosion deterioration of RC Columns [adopted from Lower (2010)]
Figure 14: Debris Accumulation and formation of Corrosion at Bearing [Lindquist (2008)]
30
There have been few studies considering the deterioration effect of different bridge
components on the bridges fragility. As an example Akiyama et al. (2011) developed fragility
of deteriorating RC columns in marine exposure by determining the displacement ductility
capacity of the members based on the corroded longitudinal rebar buckling. There is a need
to further consider the deterioration effects not only on one element of the bridge but also
on the whole system. One of the highly vulnerable component of a bridge is its bearings.
These bearings are affected by different aging and deterioration mechanisms. 1-One of
these mechanisms is strength loss of bearings due to corrosion of anchor bolts. This type of
corrosion can happen either due deicing salt application when chloride laden water leaks
through the bridge joints (Silano and Brinckerhoff 1993) or due to bridge being in marine
exposure and exposed to atmospheric chlorides. 2- Translational and rotational movements
of steel bearings might be restricted due to accumulation of rust products which causes
freezing or locking (Mander et al. 1996). 3- In concrete bridges, thermal oxidation and aging
might cause stiffening of elastomeric bearing pads (Itoh et. Al 2006). Most of the research
done on the effect of bridge elements corrosion on its seismic fragility and capacity is based
on the atmospheric exposure, however the research done by Stewart and Rosowsky
showed that chloride induced corrosion from deicing salts can considerably cause higher
degradation comparing to chloride in marine environments (Stewart and Rosowsky 1998).
The application of the deicing salts is more serious in areas across the country with cold
environment which are characterized by moderate to heavy snowfalls according to
Broomfield (Broomfield 1997) which include North Eastern States as well. Therefore, there
is definitely a need to address this type of degradation on bridge aging fragility. Bridge
31
deterioration depends mainly on construction type, material, environmental exposure and
age of the bridges. A research done by Nielson (2005) showed that most seismically fragile
bridges are Multi-span Continuous (MSC) steel girder bridges and Multi-span simply
supported (MSSS) concrete girder bridges. As an example, the state of New Jersey has 452
MSC bridges with an average age of 27.4 years out of which 9 bridges have scour critical
condition, and has 1,101 MSSS concrete girder bridges with an average age of 48.8 years out
of which 83 bridges have scour critical condition. According to a research done by Ghosh
and Padgett (2010, 2012) these bridges are more vulnerable to seismic events due to aging
and deterioration. MSC steel girder bridges are continuous over the bents having high type
steel fixed bearings and have expansion bearings over the abutments. MSSS concrete girder
bridges mainly have elastomeric expansion and fixed bearings at the supports. Table 9
shows structural components of a bridge affected by mechanisms of degradation.
32
Table 9: Mechanisms of degradation on bridge components (Ghosh, 2013)
2.1. CORROSION DETERIORATION OF RC MEMBERS
Corrosion of reinforced concrete members mainly results in loss of cross sectional area of
steel reinforcement embedded inside the members. In addition to this phenomena, it might
cause secondary effects such as cracking and spalling of concrete cover. This type of
deterioration affect all the critical concrete components of the bridge such as deck,
longitudinal and transvers reinforcement in columns, dowel bars in elastomeric bearings
33
and anchor bolts in steel bearings. Environmental exposure can be categorized into 1-
deicing salt exposure 2- marine splash zone exposure and 3-marine atmospheric exposure
with the deicing salt exposure being the most severe and marine atmospheric being less
severe as shown in Figure 15.
A research done by (Thoft-Christensen et al. 1996) proposed an equation for finding the
corrosion initiation time for deicing salt exposure. An equation for corrosion initiation time
of reinforced concrete members in marine zone under the exposure of chlorides was
proposed by the research done by (Bertolini et al. 2004; Choe et al.2009). The presence of
atmospheric oxygen simultaneously with sea water chlorides under marine exposure can
cause the most severe deterioration under this exposure (Broomfield 1997). So structures
located in splash and tidal zones face higher level of deterioration comparing to structures
fully submerged under water where lack of atmospheric oxygen presents and structures in
atmospheric zones away from the sea.
Figure 15: Environmental exposure cases (Ghosh 2013)
A study by Anxin Guo, Wei Yuan, Chengming Lan, Xinchun Guan, Hui Li revealed initiation
corrosion time for splash and tidal zone as well as atmospheric zone based on the cover
34
depth as shown in Figure 16. For concrete cover of 40, 50, and 60 mm, corrosion initiation
times of 5.8,8.2 and 11.9 years for the splash and tidal zone and corrosion initiation time of
21.8, 25.5 and 30.9 years for the atmospheric zone were obtained respectively.
Figure 16: Corrosion initiation time for a) splash and tidal zone, b) atmospheric zone (Guo,
Yuan, Lan, Guan, Li 2014)
The time dependent cross sectional area loss of reinforcements after initiation of corrosion
can be calculated by using original rebar diameter and the rate of metal loss due to
corrosion (Thoft-Christensen et al. 1996; Enright and Frangopol 1998). Due to lake of data
especially for deicing salt exposure condition, in many cases the corrosion rate is considered
to be a constant rate over the service life of the structure. The constant corrosion rates for
deicing salt exposure based on the work done by Enright and Frangopol 1998 are shown in
Table 10.
35
Table 10. Corrosion rate for deicing salt exposure (Enright and Frangopol 1998)
So the area of steel can be calculated as shown in the following equation (Thoft-Christensen
et al. 1996; Enright and Frangopol 1998):
(1) where, n is the number of reinforcement bars, Di is the initial diameter of steel
reinforcement, t is the elapsed time in years after corrosion initiation, r corr ( t) is the rate of
corrosion, and D (t) is the reinforcement diameter t years after corrosion initiation, which
can be represented as: (Ghosh, 2013).
D(t)= Di - r corr ( t)
(2)
Corrosion of steel reinforcement in columns, during an earthquake might cause loss of bond
strength leading to potential buckling of longitudinal reinforcement following concrete
36
cover spalling. A research done by (Fang et. Al. 2004; Aquino and Hawkins 2007) showed
that loss of bond strength in unconfined reinforced concrete members is significant while a
study done by (Fang et al.2004) showed that the effect of bond strength loss in members
confined transversely is negligible. In investigating the effect of bond strength loss and
reinforcement buckling, Ghosh and Padgett (2012) performed a study which showed that
even considering explicit incorporation of rebar buckling phenomena in analytical models
shifts the fragility of the bridge by less than 1%. The corrosion of steel girders also does not
really affect the seismic fragility of the bridge structures.
2.2. ELASTOMERIC BEARING PADS THERMAL OXIDATION
This type bearing which is mainly used In slab type and concrete girder bridges, is consisted
of two parts, elastomeric rubber pad and steel dowels which are both vulnerable to
deterioration and aging. Due to thermal oxidation, the bearing pads face increase in shear
stiffness while the steel dowels loss cross sectional area due to corrosion. A research done
by Itoh et. Al (2006) showed that the rubber’s shear modulus is not constant during the time
and is vastly affected by degradation mechanism such as thermal oxidation. Initial shear
stiffness of elastomeric pads can be calculated based on the work done by (Kelly 1997; Choi
2002) as shown in the equation 3.
(3)
37
Where Apad is the area of the pad, G is the rubber‘s shear modulus and tpad is the thickness
of the bearing. So the shear stiffness of the rubber due to aging can be modeled considering
the age of the rubber by the equation proposed by Itoh and Gu (2009). The Cs term in the
following equation is the strain energy temperature dependent coefficient
(4)
2.3. STEEL FIXED AND EXPANSION BEARINGS
Two primary degradation concerns that affect the performance of the steel bearings are
corrosion of anchor bolts both in fixed and expansion bearings in addition to accumulation
of corrosion debris which causes freezing or locking of the bearings. Keeper plates in
expansion bearings are also subjected to corrosion. Corrosion of anchor bolts might become
so serious that might form a “weak link” (Mander et al. 1996) during an earthquake in
transferring the forces from the superstructure to the substructure which might results in a
shift in performance of the bridge. Anchor bolt corrosion results in reduction in ultimate
lateral strength of bearings and this phenomena is more critical in areas where deicing salts
are being used. Cross sectional area loss of anchor bolts also affects the bond strength of
embedded bolts.
38
Figure 17: Section loss of anchor bolt due to corrosion deterioration (Lindquist, 2008)
Cross sectional area loss can be calculated based on the environmental exposure as
mentioned earlier. For the fixed steel bearings, the ultimate lateral strength of the bearing
can be obtained from equilibrium of forces by knowing the fact that the strength changes
over time due to degradation and corrosion of steel components of the bearing. In this
regards, equations have been proposed in the literature to calculate the ultimate lateral
strength of the fixed bearings such as the one proposed by Mander et al. 1996. In case of
expansion or rocker bearings, the coefficient of rocking friction is the most important thing
in defining the ultimate lateral strength of these types of bearings since mainly the
longitudinal motion is rocking. Mander et al. (1996) suggested a range for this coefficient
which starts from 0.04 for rocker bearings with clean well-worn conditions and goes up to
0.12 for heavily corroded bearings by considering the locking effect. In the transverse
direction, the performance of expansion bearings is mainly based on sliding frictional
component. Once the sole plate-rocker frictional resistance exceeded by the lateral
frictional force, it will cause the sol plate to slide over the rocker and the excess movement
due to this phenomena can be prevented by keeper plates. However, the keeper plate itself
might bend significantly and failure might happen due to the fillet weld tearing in case of
39
excessive horizontal load (Mander et al., 1996).There might be another mode of failure in
case of expansion bearings which is the shear failure of anchor bolts due corrosion
deterioration which reduces the shear capacity of the anchor bolts to transfer the lateral
forces to the substructure when the keeper plate is being struck by the rocker.
2.3. FIXED AND EXPANSION ELASTOMERIC BEARINGS
Both components of expansion bearings are subject to degradation. As mentioned earlier,
the elastomeric rubber pad which transfers the lateral forces by the mean of frictional
forces can undergo increase in shear stiffness due to thermal oxidation. At the same time
the steel dowels which resist the horizontal loads through a beam type action (Taylor 1969)
are subjected to corrosion and section loss. The time dependent yield shear strength and
ultimate shear strength can be calculated by the following equations proposed by (Hwang et
al. 2001; Ghosh and Padgett 2012) in which Ad(t) is the cross sectional area of dowels based
on time:
(5)
(6)
40
2.4. BENTS AND RC COLUMNS
While the cross sectional area loss of longitudinal reinforcement can be modeled and
anticipated by the procedure and equations explained earlier, the cross sectional loss of
transverse reinforcements can be taken into consideration by reducing confined concrete
strength. As mentioned by Ghosh (2013), confined concrete strength based on time can be
calculated using equations 7 and 8 where K(t) is the confinement factor K(t) at time t and
was used based on the equation proposed by (Park et al. 1982). fyh is the yield strength of
the transverse steel reinforcement and ρs(t) is the volume ratio of corroding steel hoops at
time t.
fcc = K (t)* fc
(7)
K (t) = 1 + s (t) fyh/fc
(8)
2.5. FRAGILITY CURVES BASED ON AGING
Ghosh (2013) performed a study on seismic fragility of bridges based on time-dependent
deterioration. He generated 96 models at different ages (e.g. 0,25,50,75 years) and
performed a nonlinear time history analysis on them. The peak median demands of
deteriorated components of bridges were related to ground motion intensity at that time
instant by developing probabilistic seismic demand model through the use of linear
regression analysis. A bridge is considered as a series of systems and components, so failure
41
of any element is assumed as the whole bridge failure. Two different type of bridges were
used for this study. The first one being MSC steel girder bridge under deicing salt exposure
and the second case being MSSS concrete girder under three different exposures of deicing
salt, marine splash and atmospheric zones. The ultimate lateral strength of fixed bearing
over the time due to anchor bolts’ area loss is shown in the Figure 18. (Ghosh and Padgett
2010)
Figure 18: Ultimate lateral strength of fixed bearing over the time due to anchor bolts' area
loss (Ghosh and Padgett 2010)
2.6. AGING IN RC COLUMNS
The study by (Ghosh and Padgett 2010) showed that the load carrying capacity and yield
curvature of RC columns were significantly affected by the corrosion and steel area loss
which revealed 21% reduction in the yield moment of 50 year old bridge and 16.6%
reduction in yield curvature of 50 year old bridge compared to the pristine column as
presented in Figure 19. Once both bridges (Pristine and the 50 year old) were subjected to a
ground motion, the 50 year old bridge showed greater demand on the RC column as the
peak curvature ductility of 5.4 was obtained comparing to the peak ductility of 3.3 in the
pristine case as shown in Figure 20.
42
Figure 19: Reduction in column load resisting capacity and yield curvature (Ghosh and
Padgett 2010)
Figure 20: Ductility demand of 50 year old bridge vs. pristine bridge (Ghosh and Padgett
2010)
The increase in fragility on other components in case of having a corroded RC column in the
50 year old bridge were negligible as it caused only about 3% in expansion bearing fragility
and 1% on the abutments’ fragility.
2.7. AGING OF BEARINGS
In Case of fixed bearings, the peak displacement increased by 16% in the longitudinal
direction and by 11% in the transverse direction. In case of expansion bearings, the
longitudinal deformation was reduced by 19% due to accumulation of debris which caused
increase in friction coefficient. However, in the transverse direction, the peak deformation
43
was increased by 18% due to reduced ultimate strength of the bearing comparing to a
pristine model (Ghosh and Padgett 2010).
2.8. COMBINED EFFECTS OF AGING IN BEARINGS AND
COLUMNS
Under individual degradation effects, for most of the components a steady increase in
seismic fragility of the components were observed. However, few components such as steel
fixed and expansion bearings revealed reduced vulnerability in the longitudinal direction
over the time. But at the global level, the overall fragility of MSC steel girder bridges
increased as it is shown in Figure 21 for all the damage states, particularly at complete
damage state with a 32% increase in the average value of fragility toward the bridge’s end
life. (Ghosh and Padgett 2010)
Figure 21: System level time-dependent seismic fragility curves corresponding to different
damage states for the case study MSC steel girder bridge (Ghosh and Padgett 2010)
44
For the MSSS concrete girder bridge, three different environmental exposures were studied.
The results showed that the increased in fragility due to aging under atmospheric zone
exposure had less effect (5% increase) comparing to splash zone exposure with 9% increase
and deicing salt exposure which showed 44% increase in fragility for the complete damage
state as shown in Figure 22. (Ghosh and Padgett 2012)
Figure 22: Comparison of fragility curves for MSSS Concrete Bridge for complete damage
state under different exposure conditions (Ghosh and Padgett 2012)
45
CHAPTER III
SEISMCI VULNERABILITY RISK ASSESSMENT AND
COST-BENEFIT ANALYSIS
3.1. LIFE CYCLE COST-BENEFIT
Due to limited funds and also uncertainty associated with seismic events and bridges’
fragility, life-cycle cost-benefit evaluation based on risk analysis could be the best option in
prioritizing bridges for retrofit and also for evaluating different retrofit measures. Key
factors in cost-benefit analysis are economics and social costs. Economic losses can include
losses due to replacement cost, repair cost and travel time losses. Social cost include
downtime which is the time associated with repair and restoring of a bridge after a seismic
event and fatalities. Social cost are much more difficult to predict comparing to economic
losses. The probabilistic approach is based on the fragility of the as built bridges and fragility
of the retrofitted bridges. In this procedure, the difference between the present value of
46
losses based on the as built fragilities and also the present value of losses based on the
retrofitted fragilities can be used to evaluate a retrofit measure benefit as shown in the
following equation:
Benefit = Expected losses (as-built) - Expected losses (retrofitted)
(9)
Expected Loss = Repair Cost + Travel Time Cost
(10)
The as-built costs includes repair and replacement of the bridge along with the delay in
travel time and retrofitted costs includes the damage to the bridge and the cost of retrofit.
As shown in Figure 23, Cost benefit ratio can then be used to evaluate the ratio of the
current present benefit gained from the retrofitting over the initial retrofitting cost. This is a
mean to evaluate the expected return per amount of dollars invested in retrofitting, and the
retrofit measure with the largest value of the CBR can be referred to as the measure with
the highest return per amount of money invested in the retrofitting. So basically the CBR is a
financial return measure per invested dollars and a CBR greater than one can be referred as
positive return on investment. CBRs less than one also might be also favorable based on
non-monetary benefits such as avoiding loss of lives. The whole cost benefit analysis
procedure can be shown in Figure 23.
47
CBRr = Benefitr / Costr
(11)
Figure 23: Cost-benefit analysis procedure (ODOT 2015)
Retrofit cost estimates can be done based on the Table 11 (Padgett, DesRoches) or based on
the deck area similar to what has been used by ODOT which is shown in Equation 12 and 13.
Retrofitting prices might be different based on the geographic location. ODOT refers to
superstructure retrofit as Phase I and the substructure retrofit as Phase II.
48
Table 11: Retrofit measure's cost estimate (Padgett, DesRoches)
(12)
(13)
3.2. REPLACEMENT COST
Replacement cost can also be calculated using new construction cost based on the deck
area and type of the bridge at each location based on local approximate unit prices. In order
to account for and incorporate different associated costs such as traffic control, approaches,
etc., ODOT uses a factor of 3.2 times the construction cost of the bridge and considers a
minimum cost of $3 million. It further multiplies the amount by a factor of 1.2 to account for
expected larger dimensions of the new bridge comparing to the old bridge (ODOT 2009).
3.3. FRAGILITY CURVES OF RETROFITTED BRIDGES
Based on the work done on the effectiveness of retrofit measures in the available literature,
as-built fragility curves can be scaled using median value modification factors developed by
(Padgett JE, DesRoches) to account and consider the effects of different retrofit measures
49
for different type of bridges and different type of retrofit measures used. Another way of
obtaining retrofitted fragility cures is by developing analytical models and modifying the as
built condition of the bridge by implementing retrofit measures in the analysis. This method
is applicable in the final decision stage for a particular bridge, however it is tedious and non-
applicable in network analysis when considering all the bridges in the network.
3.4. NETWORK BASED SEISMIC RISK ANALYSIS USING
REDARS2
Seismic risk assessment can be done for a region by using REDARS2 software by estimating
potential impacts of a selected earthquake event. It can also be used to assess the
effectiveness of retrofitting measures on a highway segment and also associated impacts on
travel time and traffic flows at post-earthquake stage. The network based analysis of
REDARS2 can consider redundancy or lack of redundancy in a system to highlight highways
which are more critical and gives higher return in investment by measuring the system-wide
economic losses and traffic disruptions. REDARS can estimate direct and indirect losses due
to system disruption. It can consider repair costs, delays in travel time and also losses due to
forgone trips in case of traffic congestion caused by earthquake. (OTREC 444 SRS 500-480)
50
CHAPTER IV
SEISMIC RETROFIT MEASURES
For the purpose of this study, seismic retrofit measures are categorized into three
categories such as Superstructure retrofit measures, substructure retrofit measures, and
lastly retrofit measures for seismic response modification of structures.
4.1. SUPER-STRUCTURE RETROFIT MEASURES
One of the most common types of failure of bridges during seismic events are due to
unseating of the spans. Most of the bridges built before the implementation of seismic
design might not satisfy the required seat length based on the equations proposed in the
code. The most common superstructure retrofit measures include: 1) seat extenders, 2)
catcher blocks, 3) cable restrainers, 4) restraining bars, 5) shear keys, and 6) bumpers or
stoppers. Seat extenders provide additional support length in order to avoid unseating of
spans during an earthquake event. This measure is relatively inexpensive and easy. This
retrofit measure allows the superstructure to float over the substructure and prevent
unseating. Seat extension can be done using different methods such implementing concrete
51
block, steel bracket, steel beams or even extended seating frame as shown in Figure 24 and
Figure 25 respectively.
Figure 24: Concrete block & steel bracket seat extenders (DesRoches 2008)
Figure 25: Seat extenders: steel bracket, steel beam and extended seating frame
(DesRoches 2008) The purpose of using catcher blocks is similar to the use of seat extenders with a difference
that instead of extending the seat length, a catcher block can catch the girders supported by
high-type (tall) bearings in the events they become unstable. Other than the use of catcher
blocks at the location of tall bearings, they can also be used in situations where anchoring
seat extenders might not be possible due to not enough space being available. A catcher
block is shown in Figure 26.
52
Figure 26: Catcher block installed at the location of a tall bearing (DesRoches 2008)
Another retrofit measure which can be implemented to avoid unseating of the
superstructure is by use of restrainer cables to limit relative hinge displacement at locations
where sufficient support length were not provided. These type of restrainers can be
connected in two ways; either through the bent cap or exactly between neighboring girders
as shown in Figure 27.
Figure 27: Restrainer cables connected through the bent cap and restraining cables
connected directly to the adjacent girder (DesRoches 2008) This retrofit measure is also relatively simple and inexpensive. In west Coast, this retrofit
measure has been a common approach since 1970s after the San Fernando (1971)
earthquake and most of the restrainer cables performed effectively in the Northridge (1994)
and Loma Prieta (1989) earthquakes. (Priestley et al., 1996). These are usually high strength
steel cables with Modulus of Elasticity E=10,000 Ksi and Fy=39 Kips ( Caltrans 1997) and are
designed as 0.75 inch diameter cables with an effective area of 0.22 in2, length of 5 to 10 ft.
53
and slack of 0 to 0.75 inches based on the ambient temperature( Saiidi et al., 1996).
Examples of this retrofit measure are shown in Figure 28.
Figure 28: Restrainer cables applications (DesRoches 2008)
Restrainer Bars are also used for the same purpose as Restrainer Cables. Comparing to
cables, these are usually stiffer and also more ductile. As shown in Figure 29, they can be
used to connect girders directly to columns or they can be used to connect girders together
over the piers. Restrainer bars can be used to restrain the movement in both directions at
the same time as shown in Figure 30.
54
Figure 29: Restraining cables connecting girders to column/connecting girder to girder over
the pier (DesRoches 2008)
Figure 30: Restraining cables used to restrain movement in both directions (DesRoches
2008) In order to restrain the motion of the superstructure in the transverse direction, shear keys
might be used at each bearing location. These are usually concrete blocks that facilitate
shear force transfer from the superstructure to the substructure. In order to limit the force
transferred to the substructure, sometimes these concrete blocks might be designed as fuse
elements. Usually there will be an initial gap between the shear keys and bearing on the
order of ½ inches. In addition to concrete blocks, there might be other types of shear keys
used such as keeper brackets or transverse bumpers.
55
Figure 31: Concrete block shear key (DesRoches 2008)
Figure 32: Keeper bracket shear key (DesRoches 2008)
The goal of using bumpers or stoppers is also to limit movements at hinges or support
adjacent girders. They are relatively inexpensive and easy to implement. They are shown in
Figure 33.
Figure 33: Bumpers/stoppers applications (DesRoches 2008)
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4.2. SUBSTRUCTURE RETROFIT MEASURES
Substructure retrofit measures include bent cap and column retrofit measures. The goal of
bent cap retrofitting is to improve shear strength, flexural strength and ductility capacity of
these elements making sure that the plastic hinge will form in the columns before damage
occurs in the bent caps. Several different approaches might be used for retrofitting bent
caps based on what aspect of the element needs enhancement. Commonly used bent cap
retrofit methods includes external pre-stressing as shown in Figure 34, concrete bolster or
jacketing as shown in Figure 35, or steel Jacketing as shown in Figure 36 and providing
external shear reinforcement as shown in Figure 37.
Figure 34: Bent cap retrofit: external post tensioning (DesRoches 2008)
Figure 35: Bent cap retrofit: concrete bolster or jacketing (DesRoches 2008)
57
Figure 36: Bent cap retrofit: steel jacketing (DesRoches 2008)
Figure 37: Bent cap retrofit: external shear reinforcement and confinement (DesRoches
2008)
Most of the bridges built before implementation of seismic design, have columns vulnerable
to seismic forces due to insufficient ductility capacity and shear strength which were
resulted due to insufficient splice lengths and inadequate transverse reinforcements. As
mentioned also earlier in the aging and deterioration section, confinement of concrete
members directly affects the compressive strength and also the ultimate strain capacity of
the members. So the goal of column retrofit is to improve shear strength, deformation,
ductility capacity and lap splice. Column retrofit measures include methods such as steel
jacketing, concrete jacketing, pre-stressed high strength cables, composite wraps, seismic
isolation silo, and others. Steel Jacketing has been used as one of the earliest methods of
retrofitting columns for increasing the confinement of plastic hinge region, enhancing
58
ductility, shear strength, bond transfer and lap splice performance. Steel Jacketing might be
used over the full height of the column or just over the partial height (such as at the location
of plastic hinges) as shown in Figure 38.
Figure 38: Partial and full height steel jacketing (DesRoches 2008)
Steel Jackets are usually made of A 36 steel with a minimum recommended thickness of 0.4
inch for handling issues during construction. A typical details of a steel jacketed column is
shown in Figure 39.Use of steel jacket causes un-intended increase in stiffness of the
elements. For the partial height jacketing the increase in the stiffness can be around 10% to
15% (Chai et al., 1991) and in the full height column it can be from 20% to 40% (Priestley et
al., 1996).
Figure 39: Typical details of steel jacketed column (DesRoches 2008)
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Concrete overlays or encasements can also be used as retrofit measures for columns. This
method will be mainly used to provide confinement for improving ductility capacity rather
than increasing the flexural strength since often flexural strength is not an issue. Concrete
overlays can also be either full height or partial height and might contain both longitudinal
and transverse reinforcement in them. Application of concrete overlays is shown in Figure
40.
Figure 40: Application of concrete overlay retrofit (DesRoches 2008)
Another technique to improve the confinement and increase the ductility of columns is the
use of composite wraps. Usually these materials are made either with glass or carbon fibers.
Tests showed that retrofitted undamaged columns had typical displacement ductility of 6 to
8. These wraps can be also used for post-seismic event repair, since the post-damaged
retrofitted columns revealed a ductility of 2 to 4 under tests. According to the report by
James E.Roberts on Caltrans retrofit measures, even though under tests, displacement
ductility factors of 6 to 8 have been achieved, the strategy is to place a maximum limit on
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the moment and displacement ductility demand which should not be greater than 4. The
columns under the test condition and also application of the composite wraps on highway
bridge columns are shown in Figure 41. (James E.Roberts 2005).
Figure 41: Composite wrap column retrofit (James E.Roberts 2005)
Usually during the seismic event, on bridges with variable column heights, shorter columns
which are stiffer, attract most of the seismic force near the abutments and might cause
failure. In order to fix this problem, seismic isolation silos might be used to extend the
elastic length of shorter columns by providing annular space around those columns. These
silos usually extend below the ground from 3 to 15 m. This is an effective way to distribute
the lateral force to all the columns equally be altering their heights and making them to
have the same stiffness. Example of this method are shown in Figure 42 James E.Roberts
2005).
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Figure 42: Seismic isolation silos for increasing elastic length of short column (FIB Bulletin
39)
4.3. SEISMIC RESPONSE MODIFICATION MEASURES
Another way of seismic retrofitting is by changing the response of the bridge such as
changing the vibration mode and by doing that protect the vulnerable substructure.
Vulnerable bearings not only can cause a failure, but they can also change the expected
seismic response of the bridge. Some of the vulnerable bearings are shown in Figure 43.
Figure 43: Examples of vulnerable bearings (DesRoches 2008)
There are two types of seismic response modification devices (SRMD); Isolation devices and
damping devices. Isolation devices are used to reduce the force transmitted to the
substructure by changing the period of the structure while damping devices are used to
reduce displacements.
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Figure 44: Seismic force reduction due to the increase in the period (Buckle 2016)
4.3.1. Seismic Isolation
Seismic isolation devices can be used to lengthen the period of the structure by making it
more flexible which results in substantial reduction of the demand on the substructure such
as the base shear. As shown in Figure 44, increasing the period from 0.5 sec to 1.5 sec
resulted in large reduction on the acceleration and accordingly the forces applied to the
structure.
By using seismic isolation devices, it would be feasible to have the structure perform
elastically during design earthquake and it can also significantly reduces repair costs and
accordingly continuing functionality is achievable. The increase in the period causes also
increase in the displacement. These larger displacements mainly happen in isolator
themselves not in the structure. So although this displacement might be large, the
displacement (drift) in columns is small. Additional damping devices might be used in order
to limit the displacement.
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Isolation Elastomeric Bearings
Use of Elastomeric bearings or laminated-rubber bearings is the simplest method of
isolation and it is usually a low-cost option. This method has been used almost for the past
35 years (Stanton and Roeder, 1992). This type of bearing is composed of horizontal
elastomer layers reinforced and separated by thin steel layers as shown in Figure 45. This
type of bearing require minimal maintenance but might face significant stiffness increase
due to cold temperature. Due to practical reasons, the use of elastomeric bearings is
restricted to lighter bearing loads.
Figure 45: Elastomeric bearing (DesRoches 2008)
Damping systems either built in into the isolator (Lead-Rubber bearings), or external
damping devices parallel with the isolator might be used to control the increased
displacement. Sometimes, keeper brackets might also be used to control the displacement.
Figure 46: Deformed elastomeric bearing due to a lateral Load (DesRoches 2008)
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Lead-Rubber Bearings
This type of bearing is similar to Elastomeric bearing with a difference that it has a lead core
which takes care of the displacements. It has been used on bridges requiring reasonably
small displacement at seismic events. It is made up with alternating layers of neoprene or
natural rubber and thin steel plates with a lead core of 100 to 150 mm diameter. An
example of this type of isolation device is shown in Figure 47.
Figure 47: Lead-rubber bearing
The pad can displace up to 100% of its rubber thickness without causing a failure at a
seismic event, and the lead core dissipates energy by heating up. The effect of damping is
shown in Figure 48.
Figure 48: Effect of damping on acceleration and displacement spectra when an isolation
system is used (DesRoches 2008)
65
Concave Friction Pendulum Bearings
This type of isolation bearing works as a simple pendulum principle. During a seismic event,
small simple harmonic motions is achieved due to moving of the articulated surface along
the concave surface. Due to sliding of structure along the concave inner surface, the natural
period will be increased. The frictional interface filters out the seismic forces by generating
damping friction force which acts as a damping device. This type of isolation device is shown
in Figure 49.
Figure 49: Concave friction pendulum bearing (Buckle 2016)
Adjusting the curvature and diameter of the bearing can be done to accommodate different
magnitude of displacement.
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4.3.2. Damping Devices
Shock Transmission Lock-Up Devises
These type of devises are usually used on large bridges for providing thermal expansion and
contraction. However, during a seismic event they will also lock up. These devises are
usually designed with small orifices which are used to prevent the rapid flow of the liquid
and to lock up enabling the shock waves to be transmitted to other parts of the structure.
Examples of these are shown in Figure 50.
Figure 50: Shock transmission lock-Up devices
Viscous Damper
This type of dampers are used to assist in resisting earthquake forces by absorbing energy. It
can also reduce the problems associated with soft soils. Figure 51 shows viscous dampers
that can be used to resist earthquake forces through energy absorption.
Figure 51: Viscous damper
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CHAPTER V
ANALYTICAL INVESTIGATION OF SEISMIC
VULNERABILITY OF BRIDGES
5.1. INTRODUCTION TO ANALYTICAL STUDY
In this study, different types of concrete bridges were analyzed using elastic response
spectrum and nonlinear push-over analysis to investigate the effect of multiple spans,
bridge bent type and bent continuity. Two different bent types were analyzed: single
column bent and multiple column bent. The multiple column bent selected for this study
has three (3) columns. Two types of bent continuity were evaluated in this investigation: an
integral (monolithic bent) in which the bent is integrally connected with the bridge
superstructure and a non-integral (fixed) bent in which the bent is pin connected to the
superstructure. The bridge models were subjected to three different levels of earthquake
ground motion: Low, medium, and high.
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The geometry of the considered bridges were selected based on the typical bridge spans
and element dimensions mainly in the eastern part of the country. The two span bridges
consisted of two (2) 130 ft. spans while the three span bridges consisted of 100 ft., 140 ft.
and 100 ft. spans. The single column bent has 5’ diameter concrete column and the multi-
column bents consisted of three (3) columns each having a diameter equal to three (3) ft.
The bridge superstructure has six (6) girders made of Bulb Tee sections (BT-72) and a nine
(9) in thick concrete deck. The girders are spaced 8 ft. 4 in and overhangs were three (3) ft.
on each side of the deck. The total width of the supper structure was forty eight (48) ft.
from outside of the deck to outside of the deck. The girders’ adequacy to carry the gravity
loads for the selected spans were confirmed by the PCI Bridge Design Manual (PCI 2011) as
shown in Figure 52. Element dimensions and reinforcements are summarized in Table 12.
Table 12: Bridge Elements Dimension
SINGLE Column Bent MULTIPLE Column Bent
COLUMN
DIA. 5' 3'
VERTICAL BARS 30 # 10 14 # 9
TRANSVERSE TIES #4 @ 4 " #4 @ 4 "
CLEAR COVER TO TIES 2" 2"
HEIGHT(TO THE MIDDLE OF CAP BEAM)
29' 29'
BENT CAP HEIGHT 72" 72"
WIDTH 6' 4'
LENGTH 48' 48'
DECK WIDTH 48' 48'
THICKNESS 9" 9"
OVERHANG 3' 3'
69
Figure 52: Bulb Tee Girders Maximum Span length Vs. Spacing (PCI 2011)
Bridge elevations, bent cross sections and column cross sections are shown in Figure 53
though Figure 56.
Figure 53: Elevation of a Two Span Bridge
Figure 54: Elevation of a Three Span Bridge
70
Figure 55: Bent Cross-Sections a) Single-Column Bent, b) Multi-Column Bent
Figure 56: Column Cross-Sections and Reinforcements a) 5' Diameter Column, b) 3' Diameter Column
5.2. SEISMICITY LEVELS
Three deferent seismic levels used were based on the US Department of Veteran
Affairs (Siegel,2016) seismicity levels shown in Table 13.
71
Table 13: US Department of Veteran Affairs Seismicity Levels
The levels given in Table 13 were used to select seismicity level that resemble those
experiences in the Eastern United States. The three seismicity levels used in this study are
Medium to High, Medium, and Low are the average values of are the average values of
Moderate to High, Moderate to Low, and Low respectively. These values are shown in Table
14.
Table 14: Seismicity Levels Used for the Study
Based on the seismic coefficients summarized in Table 14 and AASHTO 2012 design
spectrum function, three response spectrum curves were developed by assuming soil type
category D and a damping ratio of 5%.The response spectrum curves are presented in Figure
57, Figure 58, and Figure 59 respectively.
72
Figure 57: Generated Response Spectrum Curve for Low Seismicity Regions
Figure 58: Generated Response Spectrum Curve for Medium Seismicity Regions
73
Figure 59: Generated Response Spectrum Curve for Medium to High Seismicity Regions
5.3. MATERIALS
The material properties used to model the elements are summarized in Table 15.
Table 15: Material Properties Used for the Study
COMPRESSIVE STRENGTH
f'c (psi)
MODULUS OF ELASTICITY (psi)
COLUMN 4,000 3,604,997
BENT CAP 6,000 4,415,201
DECK 4,000 3,604,997
GIRDER 8,000 5,098,235
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5.4. FINITE ELEMENT MODELING
Three- dimensional analytical models were created using CSI Bridge 2017 to simulate the
geometry, boundary conditions and material behavior of the considered bridges. In order to
increase the accuracy of the dynamic analysis, continuous mass distribution was selected
over lumped mass option. Multi- modal linear elastic response spectrum analysis were
performed using the response spectrum curves generated based on the seismicity levels in
Table 14.
The bridge abutments were modeled as bents but their bearings were assumed to have free
movement in both longitudinal and transverse directions. According to the AASHTO Guide
Specification for LRFD Seismic Bridge Design, a bridge element can be part of the
Earthquake Resisting System (ERS), if it provides a load path. The ERS depends on the types
of supports used at the abutments and bents and the bearing properties assigned to each of
them. Since the bearings at the abutments do not have restrained degree of freedom (DOF),
they won’t provide a load path and they won’t be considered as part of the bridge ERS.
Bent columns consist of fiber-sections and are modeled using nonlinear elements. The
concrete enclosed by the transverse reinforcements is modeled as confined concrete and
the outer concrete is modeled as unconfined. Both confined and unconfined concrete
materials’ stress strain curves were generated based on Mander models as shown by yellow
and green in Figure 56 respectively. The Stress-Strain curves for 3’ diameter column is
shown in Figure 60.
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Figure 60: Material models for confined and unconfined concrete (CSI, 2017)
The base of the bents’ columns were assumed to be fixed. The mass source used in the
modal analysis, consists of the actual mass of the structure calculated by the software based
on the geometry and material properties. An additional 2-inch wearing surface was applied
uniformly as a surface load across the bridge deck. The program internally calculates the
mass of this wearing surface and adds it to the effective seismic mass.
5.5. ANALYSIS
The Analysis were performed using AASHTO Guide Specification for LRFD Seismic Bridge
Design 2011, interim 2014. Even though, the code allows the capacity to be determined
using code equations for Seismic Design categories B and C, non-linear pushover analysis
were performed for all the cases.
0
1
2
3
4
5
6
0 0.005 0.01 0.015
Stre
ss (
Ksi
)
Strain (in.)
Stress-Strain Curves:Confined Vs Un-Confined
Unconfined-4000 Psi- 3' DiaColumn
Confined-4000 PSI-3' DiaColumn
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The plastic hinge locations were generated by the software using the AUTO mode based on
the AASHTO/Caltrans hinges for concrete columns. For doing so, the relative height of the
columns in both longitudinal and transverse directions (RH Long and RH Trans) were set as
to 1.0. The shorter hinge length option was used in determining the hinge locations. The CSI
manual defines the shorter options as “The smaller value of RH Long and RH Trans is used to
determine the hinge length at the base of the column. Then the shorter remaining column
length (but not less than half of the clear height) is used to determine the hinge length at
the top of the column.”
Cracked section properties of the bent columns were obtained through iterative gravity load
analysis. The analysis calculates the axial force at the top and bottom on the column to
determine the cracked moment of inertia in the positive and negative transverse and
longitudinal directions. In the modal load case, which is the basis of the response spectrum
analysis, the cracked stiffness properties are used. The CQC method was used to combine
modes for both the transverse and longitudinal response spectrum load cases. Directional
combinations were performed by using the Absolute Sum Method such that the
acceleration loads applied in the longitudinal and transverse directions of the bridge, were
combined using the 100% + 30% rule. Once the push over curve slope became negative, it
was considered as the bent failure criterion. The analysis procedures can be summarized as
below:
-Perform iterative dead load analysis for evaluating cracked section properties
-Identify Response Spectrum and Demand Displacements
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-Determine plastic hinge properties and location
-Perform the capacity displacement analysis
-Evaluate Demand/ Capacity Ratios.
5.6. RESULTS AND DISCUSSION
The analysis results are summarized in Table 16 Table 17 for Single-column bent and Multi-
Column bent respectively. These tables show many parameters. These parameters include:
1) the number of spans (two span and three span bridges), 2) integral (monolithic) and non-
integral (fixed or pinned), 3) transverse and longitudinal earthquake loading, 4) period of
vibration for cracked and for un-cracked sections, 5) demand displacements and capacity
displacements, and 6) demand to capacity ratio (D/C).
Table 16: Analysis Results for Single-Column Bents
78
Table 17: Analysis Results for Multi-Column Bents
Table 18 shows a comparison of the response between bridges with single column bents
and three column bents. The single column has a diameter of 5 ft. and the columns in the
three-column bent each has a diameter equal to 3 ft. By looking at the results tabulated in
Table 18, and comparing the percent changes in the values when choosing multi-column
bent compared to the single-column bent, it can be observed that in the transverse
direction the demand decreased by an average of 2.25 % in the two span bridges and it
decreased by an average of 12.08 % in the three span bridges. On the other hand, the
capacity increased by an average of 27.49 % overall in all cases. The increase in the capacity
in the transverse direction would be due the framing action between individual columns of
the multi-column bents. In general, the D/C ratio dropped for multi-column bent system
compared to the single-column bent system in the transverse direction by 24.39% for two
span bridges, and by 29.99% for the three span bridges. It can be seen that the benefit of
having multi-column bent over the single-column bent in the transverse direction can
79
become more as the number of spans in the bridge increases. It can also be observed that in
the transverse direction the percent changes in the D/C ratio values seems to be
independent of the seismicity levels and bridge connectivity for the two span bridges and
the three span bridges.
Table 18: Comparison of Multi-Column Bents vs. Single-Column
In the longitudinal direction, the demand increased by adapting multi-column bents over
the single-column bents due to the fact that the multi-column bent is more rigid and has
shorter period compared to the single-column bent. However, the capacity was increased
by an average value of 79.24% for Integral bents and by an average value of 105.65% for
non-integral bents. It was observed that the benefit of using multiple-column bents over
single-column bents in the longitudinal direction is more significant in non-integral bent
compared to the integral bents. This increase in the capacity was further investigated by
checking the moment-curvature capacity of the 3’ diameter and 5’diameter columns used in
the bents. The single-column bent has lower displacement capacity compared to the multi-
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column bent, due to the fact that it consisted of larger diameter column which has less
curvature compared to columns with smaller diameters. The displacement capacity is
calculated by the following equation:
∆𝑐= 𝜃𝑝. 𝐿
(14)
Where L is the height of the column in the bent and
𝜃𝑝 = (∅𝑢 − ∅𝑦). 𝐿𝑝
(15)
Figure 61 shows the moment curvature of the 3 ft. and 5 ft. diameter columns respectively.
The 3 ft. column was reinforced with 14 # 9 bars and the 5 ft. column was reinforced with 30
# 10 bars. The curves show much higher ultimate curvature ∅𝑢for the 5 ft. diameter
compared to the 3 ft. diameter hence higher displacement capacity for the same column
length and plastic hinge length.
Figure 61: Moment Curvature Curves of a) 3' Diameter Column, b) 5' Diameter column
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It can be observed that in the longitudinal direction the benefit of having multi-column bent
over single-column bents in integral bridges is dependent on the seismicity levels. Table 18
shows that for the two span bridges the percent reduction in the D/C ratios were -2.08%, -
15.6% and -22.65% for Low, Medium and Medium to High Seismicity levels respectively. This
shows that having multi-column bents rather than single-column bents in the longitudinal
direction of Integral bridges is more beneficial in areas with higher seismic acceleration.
However, for non-integral bridges, the D/C ratios for the single column compared to the
multi-column bents decreased by -17.97% for all cases with no influence of seismicity levels.
The benefit of having integral bents over non-integral bents also was investigated by
comparing the percent changes in demand displacements, capacity displacements, and D/C
rations as shown in Table 19 and Table 20. The D/C ratio of single-column bents with
integral connection and single column bent with non-integral connection were similar in the
transverse direction. However, in the longitudinal direction, the D/C ratio decreased
significantly for the non-integral bent (With an average drop of -50.32% for two span
bridges and an average drop of 44.98% for three span bridges). It can be seen that this
reduction in the D/C ratio for the two span bridges was independent of the seismicity levels
while for the three span bridges the reduction in the D/C ratio increased by the increase in
seismic acceleration.
For multi-Column bents, having an integral bent compared to non-integral bent, increased
the D/C ratio in the transverse direction by about 3.05% for all the cases while it dropped
the D/C ratio by an average of 47.38% in the two span bridges and by an average of 37.86%
in the three span bridges in the longitudinal direction. For multi-column bents, the percent
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change by having integral bents over non-integral bents was dependent on the seismicity
levels. For high seismicity zones, the benefits of having Integral bents becomes more
significant.
Table 19: Comparison Single-Col Integral Bent Vs. Single-Col Non-Integral
Table 20: Comparison Multi-Col Integral Bent Vs. Multi-Col Non-Integral
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5.7. SIMULATION OF AGING EFFECTS AND RETROFIT OF
BRIDGE COMPONENTS
Finite element modeling and analysis is a great tool, which can be used to better design and
retrofit structures. However, care should be taken with modeling assumptions and
simplifications. Nonlinear material modeling and staged construction analysis, allow the
designers to account for time dependent material properties and also to check the
performance of the structure during construction, or at any other time during the life of the
structure under different loads, structural elements conditions, and other situations. The
time dependent material behavior becomes more crucial in concrete structures as the
concrete undergoes creep and shrinkage and also the compressive strength and modulus of
elasticity change over time. As discussed in Chapter 2, aging of the structure can also
significantly change the behavior and capacity of the structure. This section is aimed to
present how the aging conditions, material behaviors, retrofit measures and preventive
actions can be incorporated in finite element modeling of bridge structures.
Finite element software packages like CSI Bridge, can be used to model time-dependent
behaviors according to CEB-FIP parameters (Comite Euro-International Du Beton, 1993), ACI
209R, and others. Creep formulation may follow full integration or an expedited Dirichlet
series approximation (Ketchum, 1986). As mentioned in earlier, concrete sub-structure
cracks under gravity loads and this cracking will change the structure period and accordingly
the demand forces. In addition, it can also change the capacity of the members since the
push over curves will no longer be based on the gross moment of inertia of the substructure
elements. In order to account for the cracked section properties, iterative analysis should be
84
performed for finding the proper cracked moment of inertia of the bents in both directions
as explained earlier in earlier sections. Property modifiers can also be applied to modify the
member properties if the designer wishes to consider percentage of the gross moment of
inertia based on the empirical equations available in the codes.
Figure 62: Screen shot showing time dependent concrete model selection (CSI, 2017)
In case of seismic retrofit analysis and load rating of existing bridges, if a bridge has already
undergone a settlement or displacement, these effects can be analyzed in the FE model.
Displacement can be applied as external force at desired nodes. As mentioned earlier in
Chapter 2, aging in bridges can cause locking of bearings and also increase in shear stiffness
due to oxidation (Itoh et. Al 2006). In order to account for these long term effects, boundary
conditions need to be modified, or springs can be modeled using stiffness of the bearings
obtained by the equations presented in aging section to resemble the actual boundary
condition of the bridge. The boundary conditions of the bridges affects the load path and
might change the bridge behavior by making a bridge element part of ERS. So in case of
85
locked bearings, the appropriate degree of freedom should become restrained in the model
or spring elements should be used to accurately model the bearing stiffness.
Section loss of the reinforcements due to corrosion and aging can be calculated by using the
equations presented in Chapter 2. These equations take into account the age of the
structure and the time elapsed after the corrosion starting time (Thoft-Christensen et al.
1996; Enright and Frangopol 1998). Then the area of these reinforcement can by adjusted in
the finite element model to account for the section losses. Corrosion of the transverse bars
not only affects the shear capacity of the bent columns, but also affects the confined
compressive strength of the concrete as explained earlier. Similar to the vertical bars, the
area of the transverse bars should be reduced based on the section loss equations and
corrosion rates presented earlier. In addition, the confined compressive strength of the
concrete should be modified based on the factor and the model presented earlier by (Park
et al. 1982). The stress strain curves of either confined or unconfined concrete can be
modified, or a curve can be generated based on the user-input values. The change in
concrete confined strength accordingly affect the moment-curvature relationship of the
affected member followed by the member push-over curves.
In case of investigating the vulnerability of existing bridges, the time dependent shear
strength of the of the bearing dowels at fixed bearings should be calculated based on the
equations presented in Chapter 2 (Hwang et al. 2001; Ghosh and Padgett 2012), the current
capacity should be compared with the bearing reaction forces taken from the finite element
model to verify the adequacy of the dowels to transfer the forces to the substructure
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components. Moreover, the existing seat lengths should be compared with the required
seat length values from the analytical analysis to make sure unseating will not happen.
Seismic retrofit measures can be applied to finite element models simultaneously with
considering the aging effects either for protective design of new bridges or risk assessment
and retrofit design of the existing structure. As discussed earlier, the most common type of
bridge failure during earthquake events is the unseating of the bridge superstructure. The
displacement of the superstructure at the support locations can be obtained by finite
element means and can be compared to the existing seat length for the existing bridges or
can be compared with the proposed seat lengths to avoid unseating. One of the retrofit
measures that can prevent unseating is the use of restrainers either as bars or cables. These
restrainers can be modeled in the finite element models as shown in Figure 63 by defining
their length, area, modulus of elasticity and slack length to see their effects in reducing the
displacements over the supports. They can be used to attach girders from one span to
girders of the adjacent spans or they can be used to attach individual girders to the bents.
Figure 63: Screen shot showing restrainer selection (CSI, 2017)
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Response modification devices such as seismic isolator and dampers can be implemented in
finite element models through the use of spring or link elements. Joints can be connected to
the ground using spring supports which are link elements. These link elements can be either
linear or nonlinear depending on the analysis level. Nonlinear supports can be modeled to
include gaps (compression only), base isolators and viscous dampers.
Link elements can be linear, nonlinear and can have frequency dependent behavior. The
software which was used for this research, has the following link element: Linear, Multi-
linear elastic, Multi-linear plastic, Gaps, Hooks, dampers, Friction Isolators, Rubber Isolators,
T/C Isolators, Frequency-dependent springs, and Frequency-dependent dampers as shown
in Figure 64. These elements can be modeled based on their characteristics presented by
the manufacturer such as stiffness, area, length, mass and the degree of freedom or the
direction that they will be used for.
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Figure 64: Screen shot showing link element selection (CSI, 2017)
Bent columns in finite element program are comprised of fiber-sections and can be modeled
using nonlinear elements as mentioned earlier. This feature allows the designer to
investigate the benefits of column jacketing and its effects on the overall performance of
the bridge. Column jacketing changes the bent stiffness and can be modeled by defining
steel or concrete jackets( casings) and using discretized fiber sections as shown in Figure 65.
The bent caps’ shear and flexural capacities and ductility are the key factors to ensure that
the bent cap won’t fail before the formation of the plastic hinge in the columns.
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Figure 65: Screen shot showing selection menu for modeling of column casing (CSI, 2017)
To strengthen bent caps, external posttensioning is one of the retrofit techniques that can
be used. External post tensioning can be modeled by using rigid links and tendons element
and by applying the external post tension force to the tendons.
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CHAPTER VI
CONCLUSION AND RECOMMENDATIONS
6.1. Conclusions
1. The effect of aging is an important factor that needs to be considered when
evaluating load demand and component capacities for seismic design. Locking of
bearings can modify bridge stiffness and corrosion of confinement steel in columns
can reduce its displacement capacity.
2. Fragility curves can be used for evaluating the seismic performance of new bridges
and retrofitted bridges for various bridge types subjected to different peak ground
acceleration levels. They can also be used to identify the most effective retrofit
measures along with their cost-benefit analysis for a certain accepted damage level.
3. As it was recommended by other researchers, (Agarwal et al, 2010), providing
examples of bridge retrofit that considers alternatives of retrofit measures with
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their cost-benefit comparisons would be useful for engineers and will provide them
with data and guidance to model and perform bridge retrofit.
4. The Long Term Bridge Performance (LTBP) Portal can be used to identify seismically
vulnerable bridges based on the year built (age), bridge type, location, daily traffic,
and climate condition. This data can be very helpful to DOTs to classify bridges and
prioritize seismic retrofit and upgrades.
5. The study reviled that Multi-column Bents are more efficient over the Single
Column-Bent due to framing action in the transverse direction and higher moment
curvature capacity in the longitudinal directions.
6. By using Multi-Column bent instead of Single-Column bent, the D/C ratio dropped
significantly for all the cases independent of seismicity levels in the transverse
direction and also in the longitudinal direction for Non-Integral cases. However, D/C
ratio drop in the longitudinal direction for the Integral bents was dependent on
seismicity levels.
7. The study showed that integral bents perform better in the longitudinal direction
comparing to non-integral bents. The benefit of utilizing integral bents over non-
integral bents was dependent on seismicity levels except for two span single-column
bent case. In general, integral bents are more efficient as seismicity levels increases.
6.2. Recommendations
1. Explore the use of fragility concepts for various zone sin the United States to
evaluate bridge vulnerabilities and prioritize retrofits were needed.
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2. Need to establish a seismic database for bridges that can be included in the bridge
inventory. This will help bridge engineers and state officials in evaluation of bridge
seismic vulnerabilities, retrofitting, and aging.
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