SEISMIC PERFORMANCE OF MID-RISE LIGHT WOOD

FRAME STRUCTURE CONNECTED WITH REINFORCED

MASONRY CORE

Lina Zhou1, Zhiyong

Chen

2, Ying Hei Chui

3, Chun Ni

4 , Andi Asiz

5

ABSTRACT: Recent changes in building regulations in the province of British Columbia have raised the storey limit

of residential light wood frame buildings to 6 storeys. The increase in height leads to more flexible buildings,

potentially necessitating the need to rely on the stiffer elevator shaft and stair well core to reduce building deflection

under lateral loads. A study is undertaken to investigate the seismic response of light wood frame structures (LWFS)

connected to a reinforced masonry core by a ductile connection system. Numerical modelling approach is adopted

through the use of commercial software ABAQUS together with a subroutine describing hysteretic performance of

shear walls and connections as user-defined elements. The research effort presented in this paper has provided a

preliminary indication of the interaction between the reinforced masonry core and LWFS and may eventually lead to

design guidelines that can be adopted by design professionals to effectively deal with the design of mid-rise hybrid

wood frame buildings.

KEYWORDS: Wood-masonry hybrid structure, Mid-rise light wood frame building, Seismic performance, Numerical

analysis

1 INTRODUCTION 123

The North American platform wood frame building has

traditionally been restricted to 4 storeys or lower due to

fireproof restriction of building codes. The elevator shaft

and stair well cores in multi-storey light wood frame

structure (LWFS) are usually constructed with non-

combustible materials like reinforced concrete or

masonry, making the construction a multi-material

hybrid building system. In design practice of low-rise

(up to 4-storey) LWFS, engineers prefer to design wood

frame and the concrete/masonry core independently. One

of the reasons is that LWFS, unlike heavy timber frame,

1 Lina Zhou, Faculty of Forestry and Environmental

Management, University of New Brunswick, 28 Dineen Drive,

Fredericton NB, Canada. Email: lina.zhou@unb.ca 2 Zhiyong Chen, Faculty of Forestry and Environmental

Management, University of New Brunswick, 28 Dineen Drive,

Fredericton NB, Canada. Email: zchen3@unb.ca 3 Ying Hei Chui, Faculty of Forestry and Environmental

Management, University of New Brunswick, 28 Dineen Drive,

Fredericton NB, Canada. Email: yhc@unb.ca 4 Chun Ni, Building Systems, FPInnovations, 2665 East Mall,

Vancouver, BC, Canada. Email: Chun.Ni@fpinovations.ca 5 Andi Asiz, Department of Civil Engineering, Prince

Mohammad Bin Fahd University, PO BOX 1664, Al Khobar

31952, Saudi Arabia. Emial: aasiz@pmu.edu.sa

can resist lateral loads by its own shear wall system.

Another reason is the design seismic force of wood

structure will be substantially higher if the wood frame

and reinforced concrete/masonry core are connected

together. In this case, the lower value of the force

modification factor, RdRo (Rd: ductility-related force

modification factor; Ro: over-strength related force

modification factor) for concrete or masonry structure

has to be assigned for the whole building according to

the National Building Code of Canada [1]. For instance,

the value of RdRo for timber structure constructed with

nailed shear walls and wood-based panels is 3.0 × 1.7 =

5.1, while for reinforced masonry structures built with

moderately ductile shear walls it is 2.0 × 1.5 = 3.0 [1].

In April 2009, the government of British Columbia in

Canada announced that it had amended its building code

to allow 6-storey residential LWFS to be constructed in

its jurisdiction. The increase in height leads to more

flexible buildings. For example, the lateral drift of a six

storey light wood frame building could reach up to

300mm as estimated by design engineers. Therefore it

may be necessary to utilize the existing stiffer

concrete/masonry core to reduce building deflection

under lateral loads. A ductile connection system is

desired to transfer loads from wood frame to stiff core to

ensure acceptable performance under lateral loads and to

allow the wood frame subsystem to retain the values for

seismic force modification as if it was designed as a

stand-alone structure.

Since construction of mid-rise hybrid LWFS is a

relatively new experience in Canada and the world, there

is a knowledge gap on the structural performance of this

hybrid building system when LWFS is structurally

attached to a stiff core. This project is part of Theme 2 –

Hybrid building systems of NSERC Strategic Research

Network on Innovative Wood Products and Building

Systems (NEWBuildS). The goal of the project is to

study, through numerical modelling, seismic

performance of multi-storey light wood frame building

connected with reinforced masonry core and how the

responses are influenced by the characteristics of the

building systems and inter-connections. Reinforced

masonry core was selected instead of reinforced concrete

core because it is the more conservative choice due to its

lower ductility performance. Results described in this

paper are from preliminary analysis of a designed six-

storey light wood frame building located in three

Canadian cities and its relative hybrid wood-masonry

structure under a series of seismic excitations.

2 DESCRIPTION OF HYBRID

BUILDING SYSTEM

2.1 DESIGN OF LWFS

A procedure to establish the seismic behavior factor of

timber building was proposed by Ceccotti and Sandhaas

[2]. According to their procedure, the structure would be

designed elastically with RdRo=1 and a certain peak

ground acceleration (PGAdesign). Fictitious shear wall

length was assigned to each storey of the building based

on a typical hysteresis loop of wood shear wall used as a

standard and the design storey shear. So the building was

designed to just resist the seismic force assuming linear-

elastic behavior. Analysis of building performance was

performed through the use of a numerical model. A

series of earthquake excitations were applied to the

building model. For each of the excitations, the PGA

was scaled until a near-collapse criterion is reached. The

ratio of PGAnear-collapse and PGAdesign is the R value.

The design approach and numerical modeling in this

project is based on Ceccotti and Sandhaas’s proposed

procedure. A six-storey light wood frame building is

designed with elastic performance assumption according

to appropriate design standards and building codes in

Canada, including the National Building Code of Canada

(NBCC) [1], Canadian design standard for masonry

structures [3] and engineering design standard for wood

structures [4]. The building is assumed to be located in

three Canadian cities: Victoria, Ottawa and Halifax to

cover a range of seismic intensities. The total building

height of the structure is 16.8m above ground level with

2.8m for each storey. The rough floor area is 696m2.

Dead load of 0.7kpa for roof, 1.3kpa for floor and 0.5kpa

for partitions are assigned to this building. So the total

building weight in three cities is: 7113kN, 7329kN and

7294kN respectively. Figure 1 shows the plan view of

layout of wood shear wall and the reinforced masonry

core of the building.

Figure 1: Layout of wood shear wall and masonry core

To simplify the design procedure, the building is

assumed to only resist seismic load without considering

other load combinations. And only the shear walls along

east-west direction are designed and analyzed. Site class

D is assumed to all of the three cities with peak ground

acceleration of 0.61g, 0.32g and 0.086g respectively.

Based on the above information, the seismic coefficient

expressed as the ratio of design base shear to total weight

of the building are 0.816, 0.488 and 0.199 respectively.

So the base shear of the building in the three cities is

5804kN, 3577kN and 1454kN which are distributed to

each storey by base shear method (See Table 1). This

design ensures that the subsequent numerical analysis

covers a range of building characteristics, including

natural period and flexibility, and a range of seismic

intensity.

Table 1: Storey shear of wood structure

Storey shear (kN)

Level Victoria Ottawa Halifax

6 1238 908 360

5 2760 1798 725

4 3978 2509 1016

3 4891 3043 1235

2 5500 3399 1381

1 5804 3577 1454

There was no wood shear wall test conducted in this

project. For the numerical modeling purpose according

to Ceccotti and Sandhaas [2], what is required is shear

wall hysteresis loop data for a standard length. Total

shear wall length could be calculated for each storey of

the building based on the load-carrying capacity of the

standard wall and the design storey base shear.

Experimental hysteresis loop of shear wall cited by Xu

and Dolan [5] was adopted as a standard wall data in this

preliminary analysis (Figure 2). The ultimate lateral

resistance of the standard wall is 9.49kN. As the factored

design resistance of the shear wall is approximately half

A

B

C

D

1 2 3 4

9000

6200

9000

24200

10000 10000 10000

30000

SW-2

SW-1

SW-2

SW-1

SW-2

SW-1

SW-2

SW-1

SW-A SW-A

SW-A SW-ASW-BMW-C

MW-D

1600

3000

SW-B

SW-B

SW-B

of the ultimate lateral resistance accordance to the CSA

O86 commentary [6], it equals to 4.75kN. Based on this,

the total length of the shear walls on each storey of

buildings can be calculated.

Figure 2: 1.22m×2.44m wood shear wall (Xu and Dolan[5])

2.2 MASONRY STRUCTURE

There are two main masonry walls along the east-west

direction of the hybrid building in this project (See

Figure 1). One is 9.4m in length and the other 6.0m. The

total length of reinforced shear wall in each storey is

around 15.4m. According to the equation of factored

shear resistance of reinforced masonry wall described in

CSA S304.1 [3], the shear resistance of reinforced

masonry wall is proportional to its length if there is no

vertical load applied. In this project, the building weight

is resisted by the wood structure. The vertical load

resisted by masonry wall is its own weight, stairs and

elevators which can be ignored here. So the shear

resistance of the two masonry walls can be calculated

based on the length ratio with a standard wall by

assuming the material and reinforcement of these walls

along the whole building height are the same as the

standard masonry wall.

Figure 3: Hysteresis loop of 1.8m×3.6m masonry wall (Shedid et al [7])

Shedid et al [7] tested six 1.8m×3.6m fully grouted

reinforced masonry walls failing in flexure. These walls

were constructed with hollow 20cm blocks and type S

mortar. The average compressive strength of masonry

prisms is 14.8MPa. The hysteresis loop of the No.2 wall

in [7] was adopted in this project as a standard wall for

masonry wall modelling (See Figure 3). In wall No. 2,

the ratios of vertical and horizontal reinforcement were

0.78 and 0.13 respectively. Nine No.20 reinforcing bars

were placed vertically at every cell and No.10 bars were

placed every 400mm height horizontally.

With a height/length ratio of 2 and roughly moderate

reinforcement in both vertical and horizontal direction,

the No.2 wall exhibited flexural failure. Plastic hinges

were well developed at the lower part of the wall,

whereas the top portion of the wall deformed very little

and behaved as a relatively rigid body. Based on this

observation, the lateral deflection of masonry wall could

be assumed to be linearly increasing along the height. So

the hysteresis loop of the masonry wall in this project

can be derived from the test data of 1.8m × 3.6m No.2

Wall (Figure 3) by scaling the ultimate lateral load and

relative displacement.

2.3 WOOD-MASONRY CONNECTION TEST

A wood-masonry connection system suggested by

design engineers was tested in laboratory to generate

load-slip curves and hysteresis loops as initial input to

computer modelling analysis. In this connection system,

a grade 8 bolt was used to join the hollow masonry block

and 2×8 SPF dimension lumber, No. 2 grade or better.

The size of the masonry block was 380mm×190mm×

190mm. Diameters of the bolts tested were 12.5mm (½

in.) and 18.5mm (¾ in.). Two loading regimes

(monotonic load and reversed cyclic load) were applied

to the connection specimens. Two types of test

specimens with loading applied parallel and

perpendicular to wood grain were tested. Each loading

and specimen combination had three replicates.

Monotonic loading tests were conducted first to get the

reference yield displacement Dy for design of the

reversed cyclic loading protocol. Figure 4 shows the test

apparatus of parallel and perpendicular loading tests.

(a) Parallel loading (b) Perpendicular loading

Figure 4: Apparatus of wood-masonry connection test

Eight groups of load-slip response curves have been

obtained from experiments, including 2 bolt diameters, 2

loading directions and 2 loading protocols. The most

common failure mode of these connection combinations

is the wood split as summarized in Table 2, except that

for the 12.5mm bolt under cyclic loading, the bolt broke

due to fatigue. The 12.5mm bolt always showed large

bending deformation during test, making the connection

-12

-8

-4

0

4

8

12

-60 -40 -20 0 20 40 60

Forc

e (k

N)

Displacement (mm)

Test Data

Load parallel

to wood grain

Load perpendicular

to wood grain

Load parallel

to wood grain

Load perpendicular

to wood grain

Bolt

system more flexible than 18.5mm bolt connection,

while the ultimate load resistance is much lower. In

some cases when the load is applied parallel to wood

grain, the masonry blocks fractured. Pictures of

connection failure modes are presented in Figure 5.

Generally, this wood-concrete block connection system

could reach large deformation before it failure (See

Figure 6) which is desirable for seismic performance of

hybrid buildings.

Table 2: Failure mode of wood-masonry connections

Failure

mode

Monotonic cyclic Para. Perp. Para. Perp.

12.5 18.5 12.5 18.5 12.5 18.5 12.5 18.5

Wood Split 3

* 3 3 3 3 3 3

Local crush 3 3 3 3 3

Bolt Bending 3 3 3 3

Broken 2 3

Block Broken 1 1

Local crush 1 1 3

Note: * Number of replicates

(a) Wood split (b) Wood crush (c) Bolt bending

(d) Bolt broken (e) Block crush (f) Block broken

Figure 5: Failure modes of wood-masonry connections

In this preliminary analysis, the connection is assumed to

transfer shear between wood structure and masonry core.

The hysteresis loop of 12.5mm bolt connection under

parallel wood grain loading was selected in the

numerical modelling analysis instead of 18.5mm bolt

connection because it has larger deformable ability (See

Figure 6). In the 12.5mm bolt connection test, significant

wood split had already been developed during the 7th

loading circle. The load resistance was dropped rapidly

on the negative loading direction after the 7th

circle for

both specimens CPa102 and Cpa103 (Figure 6a). So the

minimum load is chosen to be the ultimate lateral load

resistance of the connection even higher load are shown

of specimen CPa102 and Cpa103 after the 7th

circle. The

12.5mm bolt was assumed to be placed at each cell

length. So there are total 77 wood-masonry single-bolt

connections along the 15.4m masonry wall on each

storey of hybrid building.

(a) 12.5mm bolt connection

(b) 18.5 mm bolt connection

Figure 6: Hysteresis loop of wood-maonry connections under parallel loading

3 NUMERICAL MODELLING

APPROACH

The behavior of multi-storey wood-masonry hybrid

buildings under seismic loads is studied through the use

of numerical models. Wood shear walls, reinforced

masonry shear walls and wood-based connection

systems all exhibit nonlinear, inelastic and

strength/stiffness degradation behavior under reversed

cyclic load. An analysis program capable of performing

non-linear dynamic analysis is required. The general

finite element program, ABAQUS v6.10, was used for

this purpose. The hysteretic behavior of the wood,

masonry sub-systems and inter-connections is modeled

using a subroutine developed by Xu and Dolan [5] that

incorporates the Bouc-Wen-Baber-Noori (BWBN)

model. BWBN model is a versatile, smoothly varying

hysteresis model originally proposed by Bouc in 1967

[8]. This model has been continuously developed during

last half century and was first introduced to model wood

structures by Foliente in 1993[9] who generalized the

pinching function of wood products in the model. Xu

and Dolan in 2009 [10] made a further improvement of

the model to avoid the pinching lag phenomenon and be

capable of representing small loops during partial

loading-reloading cycles. The modified BWBN model

used by Xu and Dolan contains 16 parameters to

describe the hysteresis performance of shear walls and

13 parameters for nail connections under reversed cyclic

loads. A genetic algorithm is used as a fitting method to

obtain these parameters from load-slip test data. Detailed

-20

-15

-10

-5

0

5

10

15

20

-40 -30 -20 -10 0 10 20 30 40

Load

(kN

)

Displacement (mm)

CPa101Cpa102CPa103

-30

-20

-10

0

10

20

30

-40 -30 -20 -10 0 10 20 30 40Lo

ad (

kN)

Displacement (mm)

Cpa201Cpa202Cpa203

explanation of the importance of various parameters and

equations of BWBN hysteresis loop can be found in

reference [5, 10-11].

In view of the connection of a large number of non-

linear hysteretic elements and the need to evaluate mid-

rise buildings, it was considered more efficient in the

first instance to study the seismic performance using

concept of simplified shear wall model by eliminating

individual nails and 2 dimension (2-d) buildings. In a 2-d

building model, all wood shear walls in a storey are

grouped into one super shear wall element. Likewise, the

reinforced masonry walls are represented by another

super element. The super element contains four rigid

truss elements pin connected at each corner and two

diagonal springs simulating the lateral hysteretic

performance of the walls. These two super elements are

connected by a pair of hysteretic springs that represent

the bolted connections used to connect the two sub-

systems. Figure 7 shows a schematic of the 2-d modeling

approach.

Figure 7: Schematic of 2-d modeling approach

Hysteretic loops of the 3 super elements of designed

buildings are scaled from test data conducted in this

project (Figure 6) or by other researchers (Figure 3).

Figure 8 shows the parameter estimation of hysteresis

loops of a standard wood shear wall, scaled 15.4m

masonry wall and lumped 77 wood-masonry single-bolt

connections used in this project and the comparison of

hysteretic loops between test data and numerical

modeling. As shown in Figure 8, good agreement is

found between the test data and fitting model for all of

the three components on the level of ultimate lateral load

resistance and displacement, while the unloading and

reloading tracing curves of the models of masonry wall

and wood-masonry connections is not as good as that of

the standard wood shear wall in this preliminary

analysis. Further optimization of parameter estimation is

needed to more accurately model the hysteresis

properties of building components.

(a) 1.22m×2.44m wood shear wall (Xu and Dolan[5])

(b) 15.4m masonry shear wall

(c) lumped 77 wood-masonry connections

Figure 8: Parameter estimation and comparison of fitted and test hysteresis loops

4 NUMERICAL MODELLING

ANALYSIS

Six-storey light wood frame buildings located at three

cities: Victoria, Ottawa and Halifax with and without

connecting to masonry structure are analyzed under three

ground motions: El Centro, Taft and Nahanni (Figure 9).

In this preliminary analysis, the first 10 seconds of all

three ground motions are intercepted to control the

computer running time. For each earthquake excitation

and building combination, the PGA is scaled from 1 to

1.5, 2, 2.2, 2.4, 2.6, 2.8 and 3.0 times of the local design

value. In total 144 dynamic time-history analyses were

conducted in this project.

Failure is considered to be happened when one of the

ultimate lateral deformation of wood shear wall,

masonry wall or wood-masonry connection is reached.

The ultimate lateral deformation of masonry wall and

connection are defined as the displacement at which the

lateral load resistance dropped to 80% of the ultimate

MasonryCore

Woodframe

Connection

Trusselement

Diagonalspring

Pin

-12

-8

-4

0

4

8

12

-60 -40 -20 0 20 40 60

Forc

e (k

N)

Displacement (mm)

Test DataSuper Element

p = 0.106 q = 0.117 a = 0.0478 b = 0.06 g = -0.00596 w = 0.868 z10 = 0.907 n = 1.103 y0 = 0.871 dy = 0.218 dn = 0.000002 x0 = 0.000015 dh = 0.00011

-3000

-2000

-1000

0

1000

2000

3000

-60 -40 -20 0 20 40 60

Load

(kN

)

Displacment (mm)

Scaled testModel

p = 0.10 q = 0.10 a = 0.02 b = 0.45 g = -0.030 w = 18.0 z10 = 0.98 n = 1.05 y0 = 1.0 dy = 0.2 dn = 0.0000002 x0 = 0.000015 dh = 0.0000001

-1500

-1000

-500

0

500

1000

1500

-30 -20 -10 0 10 20 30

Load

(kN

)

Displacement (mm)

Scaled testModel

p = 0.15 q = 0.05 a = 0.02 b = 0.7 g = -0.55 w = 15 z10 = 1.0 n = 1.05 y0 = 4 dy = 0.02 dn = 0.0000002 x0 = 0.000015 dh = 0.000001

lateral load resistance. In this project, 40mm is used for

masonry wall and 30mm for connection. The ultimate

lateral deformation of wood structure is defined as the

lateral storey drift reaching 1/40 of storey height and

60mm is used for safe.

Figure 9: First 10 seconds ground motions of El Centro, Taft and Nahanni

5 PRELIMINARY RESULTS AND

DISCUSSION

5.1 FAILURE ANALYSIS

In this preliminary analysis, only up to 3 time of design

PGA were applied to both pure wood and hybrid

structure. In most of the cases, no failure was noted

according to the failure criterions described in Section 4,

especially for buildings located at Ottawa and Halifax.

For buildings located at Victoria, pure wood structure

failed as the top storey drift reached 67.15mm and -

74.88mm (beyond 60mm) under El Centro and Taft

earthquake respectively when the PGA was scaled to 2

times the design value, while for hybrid buildings at

Victoria, the masonry structure failed first as the 1st

storey drift of masonry core reached 37.21mm (almost

40mm) under El Centro earthquake with PGA scaled to

2.4 times the design value and -37.19mm under Taft

earthquake with PGA scaled to 2 times the design value.

Because of limited failure statistics, there is no R value

result presented in this paper and further analysis is

required with higher PGA applied to those structures

until one of the near-collapse criterions is reached.

(a) Ottawa, El Centro (b) Ottawa, Taft

(c) Ottawa, Nahanni (d) Victoria, El Centro

(e) Victoria, Taft (f) Victoria, Nahanni

(g) Halifax, El Centro (h) Halifax, Taft

(i) Halifax, Nahanni

Figure 10: Comparison of lateral drift of pure wood and hybrid structure

-1.2-0.9-0.6-0.3

00.30.60.91.2

0 1 2 3 4 5 6 7 8 9 10

Acc

eler

atio

n ,

g

Time (s)

Nahanni

-0.4-0.3-0.2-0.1

00.10.20.30.4

0 1 2 3 4 5 6 7 8 9 10

Acc

eler

atio

n ,

g

Time (s)

EL Centro

-0.2-0.15

-0.1-0.05

00.05

0.10.15

0.2

0 1 2 3 4 5 6 7 8 9 10

Acc

eler

atio

n ,

g

Time (s)

Taft

0

1

2

3

4

5

6

7

-300 -150 0 150 300

Sto

rey

Displacement (mm)

0

1

2

3

4

5

6

7

-300 -150 0 150 300

Sto

rey

Displacement (mm)

0

1

2

3

4

5

6

7

-300 -150 0 150 300St

ore

y Displacement (mm)

0

1

2

3

4

5

6

7

-300 -150 0 150 300

Sto

rey

Displacement (mm)

0

1

2

3

4

5

6

7

-300 -150 0 150 300

Sto

rey

Displacement (mm)

0

1

2

3

4

5

6

7

-300 -150 0 150 300St

ore

y

Displacement (mm)

0

1

2

3

4

5

6

7

-100 -50 0 50 100

Sto

rey

Displacement (mm)

0

1

2

3

4

5

6

7

-100 -50 0 50 100

Sto

rey

Displacement (mm)

0

1

2

3

4

5

6

7

-100 -50 0 50 100

Sto

rey

Displacement (mm)

5.2 LATERAL DRIFT OF PURE WOOD AND

HYBRID BUILDING

Figure 10 shows a comparison of maximum positive and

negative lateral drift on each storey of mid-rise light

wood frame building and hybrid wood-masonry building

under the same PGA: 3 times the design value, except

for buildings located at Victoria under El Centro and

Taft earthquakes where 2 times design PGA was picked.

Analytical result of hybrid buildings located at Halifax

under Taft earthquake is not available because the

analysis can’t get convergence at the moment. From

Figure 10, significant lateral drift reduction is observed

for almost every city and earthquake combination. And

the drift reduction is cumulated as the storey height

increases. Table 3 summarizes the ratio of lateral drift of

wood structure in hybrid system to pure wood system at

the top of the building.

Table 3: Ratio of lateral drift of hybrid wood to pure wood system

Victoria Ottawa Halifax

El Centro 41 37 29

Taft 75 57 N/A

Nahanni 76 61 40

Table 3 shows that the lateral drift reduction varies

depending on the earthquake and location. The ratio of

lateral drift of wood structure at Victoria is much higher

than Ottawa and Halifax which means a lower lateral

drift reduction is obtained for buildings at Victoria.

Because the wood structure at Victoria is much stiffer

than the ones at Ottawa and Halifax, while the design

details of masonry structure and connection system are

the same for all of the buildings at the three cities which

leads to a different stiffness ratio of masonry core to

wood structure. Figure 11 shows the designed shear

resistance of wood structure at these cities and the

masonry core. Design shear resistance of 15.4m masonry

core is around 1582kN, and the design shear resistance

of wood structure for each storey can be found in Table

1.

Figure 11: Lateral shear resistance of wood and maonry structrue

Higher stiffness ratio of masonry to wood structure

produces higher lateral drift reduction. Table 3 also

shows that the lateral drift reduction may be influenced

by earthquake excitations, as the reduction under El

Centro earthquake is much more than that under Taft and

Nahanni earthquakes.

5.3 DISTRIBUTION OF BASE SHEAR

Table 4 presents the distribution of base shear between

wood and masonry structure in hybrid building system.

More than 80% of base shear is resisted by the wood

sub-system at Victoria, while only around 70% and 40%

of the base shear is resisted by wood frame in Ottawa

and Halifax. The main reason is due to the ratio of lateral

stiffness between wood structure and masonry core. A

larger base shear is distributed to the masonry sub-

system as the total length of shear walls in wood frame

structure is getting less. The effect of earthquake

excitation appears small.

Table 4: Distribution of base shear to wood structure

Victoria Ottawa Halifax

El Centro 81 74 44

Taft 84 77 N/A

Nahanni 80 67 43

5.4 WOOD-MASONRY CONNECTION

There are a total of 77 wood-masonry connections

distributed at each storey in the hybrid buildings. If half

of the ultimate lateral resistance of the connection is

regarded as the design shear resistance, the total design

shear resistance of connection on each storey is about

585kN. Figure 10 shows the ultimate storey drift of

wood and masonry structure in the same building is

almost the same. Actually, in both pure wood and hybrid

building system, the lateral drift on each storey reached

its positive and negative peak value simultaneously

under the selected three ground motions. Figure 12

shows an example of inter-storey drift versus time of

wood and masonry sub-systems in the same building,

where S/M dui means the inter-storey drift on ith

floor of

wood/masonry structure.

(a) Inter-storey drift of wood structure

-30

-20

-10

0

10

20

30

0 1 2 3 4 5 6

Dis

pla

cem

ent

(mm

)

Time (s)

Sdu1 Sdu2 Sdu3

Sdu4 Sdu5 Sdu6

(b) Inter-storey drift of masonry structure

Figure 12: Inter-storey drift of hybrid building

The deformation of the connection system is very small.

Most of them remain within 10mm. Even in the cases of

building at Victoria under El Centro and Taft earthquake

where failure has already been developed, the

deformation of connection only reaches up to 13.6mm

and 15.7mm respectively. Table 5 summarizes the

maximum positive and negative deformation of

connections. The connection system in this preliminary

analysis is quite strong and no failure happened at the

connection. It could transfer load from wood structure to

masonry core effectively and works more like a rigid

link which explains why the masonry wall fails first in

hybrid building. Since the ultimate displacement of

wood shear walls is larger than the ultimate displacement

of masonry core (60mm versus 40mm), changing the

number of the connections on each storey may change

the failure mode of the hybrid system and the

distribution of the base shear between wood and

masonry sub-systems.

Table 5: Deformation of connections (mm)

Victoria Ottawa Halifax

El Centro 13.6 -5.8 6.1 -3.8 2.3 -1.5

Taft 15.7 -8.2 8.2 -8.4 NA NA

Nahanni 8.4 -6.9 4.2 -4.0 1.3 -1.1

5.5 EFFECT OF GROUND MOTION

There are three ground motions considered in this paper:

El Centro, Taft and Nahanni. Although the PGA of

Nahanni in un-scaled motion is much larger than that of

El Centro and Taft motions (Figure 9), the earthquake

effect under scaled motion is less significant as the

ground motion are scaled based on its PGA. The other

observation about the Nahanni motion is that near the

PGA value the acceleration values are much smaller than

the PGA meaning that the motion was spikier. So the

lateral drift of building system under scaled Nahanni

motion is much smaller than that under El Centro and

Taft motion (See Figure 10 c, f, i) which will lead to

higher R value.

6 CONCLUSIONS

Dynamic time-history analyses of six-storey light wood

frame building designed elastically at three Canadian

cities, with and without connection to a reinforced

masonry core were conducted through the use of

numerical modelling approach under three earthquake

excitations. All the storeys in both pure and hybrid

buildings oscillate with the same phase (Figure 12). The

first storey drift is always the key character that controls

the failure load except for buildings located at Victoria

where top storey drift controls due to bullwhip effect

(Figure 10 d, e). Wood-masonry connections exhibit a

relatively rigid link performance, leading to masonry

sub-system reaching its deformation capacity earlier than

the wood sub-system. The attachment of wood structure

to a reinforced masonry core does reduce the lateral drift

of wood structure substantially, especially at the top

floor of the building. The base shear distribution

between wood and masonry structure is affected by

relative stiffness of wood and masonry structure. Ground

motions show no significant influence on it. The lateral

drift reduction is influenced by both stiffness ratio and

ground motion. To extract general conclusions about

seismic performance of hybrid building system, more

ground motions are required in further analysis to cover

a range of earthquake characteristics. Design procedure

and numerical modelling approach used in this project

are not limited to light wood frame structure connected

to a reinforced masonry core. Other types of core sub-

systems, such as reinforced concrete and cross laminated

timber (CLT), can be analysed using the same approach

as long as appropriate mechanical properties are

available.

7 LIMITATION OF THE WORK

The results described in this paper are based on the

preliminary analysis with limited earthquake excitations

and lower level of peak ground accelerations. Further

modeling analysis with higher PGA is required to

investigate the R value of the hybrid LWFS. The two-

dimensional modeling approach essentially assumes that

the wood roof and floor diaphragms behave rigidly

which may not be true in practice. The stiffness of floor

diaphragm could influence the lateral load sharing

between wood shear walls and wood-masonry sub-

systems. The simplified hybrid wood-masonry model

only considers the racking deformation of shear wall

system, which may provide a lower estimate of the total

lateral drift of building as the uplift deformation is

ignored and is cumulative with the increase of storey

height. The deformable ability of structural components

is always desired for seismic performance of buildings to

dissipate earthquake energy, while control of total lateral

drift of building is another aspect of design consideration.

Connecting wood structure with a stiffer core will

significantly reduce the lateral drift of the building;

while at the same time its deformability and energy

dissipation capacity will also be reduced. The masonry

structure adopted in this project with moderately

-30

-20

-10

0

10

20

30

0 1 2 3 4 5 6

Dis

pla

cem

ent

(mm

)

Time (s)

Mdu1 Mdu2 Mdu3

Mdu4 Mdu5 Mdu6

reinforcement produces quite ductile behavior as

indicated in Figure 3, while it may not be suitable to

connect wood structure to a really stiff structure with

small deformability. Further investigation is required to

achieve a deeper understanding of this important issue.

Basically, the preliminary analysis reported in this paper

provides a good indication on how light wood frame

structure interacts with a stiff sub-system in hybrid

buildings. Reducing the lateral drift of wood structure

by attaching it to a stiffer sub-system is feasible if proper

ductility ability of the stiffer sub-system and connection

system are considered in design.

ACKNOWLEDGEMENTS

Funding of this project is provided by NSERC Strategic

Network on Innovative Wood Products and Building

Systems (NEWBuildS).

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