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Journal of Rock Mechanics and Geotechnical Engineering. 2011, 3 (4): 329–342
Key issues in rock mechanics of the Three Gorges Project in China Qixiang Fan*, Hongbing Zhu, Xuchun Chen
China Three Gorges Corporation, Yichang, 443002, China Received 29 April 2011; received in revised form 4 July 2011; accepted 26 July 2011
Abstract: The Three Gorges Project is one of the essential key projects for flood controlling and water resources regulation in the Yangtze River. The project includes a river-crossing dam, underground powerhouses, and navigation structures. Because of the huge size and complicated construction technologies, the project faced a series of challenging engineering issues. In terms of rock mechanics, there are many key technical issues, including the sliding resistance and stability of the dam section along the foundations of powerhouses No.1–5, the slope stability of the double-line five-stage shiplock, excavation of large-scale underground powerhouses, and curtain grouting under the dam. With decades of scientific research and 16 years of practical construction experiences and reservoir operations, these key technical issues in construction of the Three Gorges Project are successfully resolved, which will attribute to the development of hydropower technology. On the basis of the monitoring data during construction and normal operation periods of the Three Gorges Project, this paper presents a systematic analysis of these key rock mechanical issues in terms of behaviors, solutions, dynamic controlling, monitoring arrangement and integrated assessment. Key words: Three Gorges Project; rock mechanics; dam sliding resistance and stability; high shiplock slope; underground powerhouses; curtain grouting
1 Introduction 1.1 Description of the Three Gorges Project
The Three Gorges Project is located in Yichang City, Hubei Province, China. The project has a normal pool level of 175 m, a storage capacity of 3.93×1010 m3, and a flood storage capacity of 2.215×1010 m3. It is able to effectively regulate floods from the upper reaches of the Yangtze River and protect the plains in the lower reaches against floods. The Three Gorges dam, a concrete gravity dam, has a crest height of 181 m and a total axial length of 2 309.5 m. The hydropower station has an installed capacity of 22.50 GW, with an average electricity generation of 1.024×1011 kW·h in a year, making it the world’s largest hydropower plant. The double-line five-stage continuous shiplock enables 10 000 t (ton) freighters to sail through in one time and the one-stage vertical ship lift can accommodate 3 000 t ships in the tank. Figure 1 gives a panoramic view of the Three Gorges Project after completion.
Doi: 10.3724/SP.J.1235.2011.00329 *Corresponding author. Tel: +86-717-6276666;
E-mail: [email protected]
Fig.1 A panoramic view of the Three Gorges Project after completion.
The Three Gorges Project was built in three phases.
Official construction started on December 14, 1994, and river closure was completed on November 8, 1997, and closure for the diversion channels was completed on November 6, 2002. In June 2003, the third phase stated as the roller compacted concrete cofferdam was put into service to block the river, and the water level in the reservoir rose to 135 m. After that, the double-line five-stage shiplock began trial operations. On July 18, 2003, the first group of generating units was connected to the power grid and began to operate. On May 20, 2006, the dam reached its designed elevation of 185 m. In October, water level in the reservoir rose to 156 m, and the hydropower complex began its initial operative phase. On June 11, 2007, the first group of generating units in the right bank powerhouse started their commercial operation. In late 2008, all of the twenty six 700 MW hydro-turbine
330 Qixiang Fan et al. / J Rock Mech Geotech Eng. 2011, 3 (4): 329–342
generating units in the powerhouses on the left and right banks of the river behind the dam were put into operation. Thus, the hydropower complex was completed (excluding the construction-postponed ship lift), and the reservoir was ready to impound to the normal level of 175 m. The Three Gorges Reservoir started a trial water storage at the end of rainy season in 2008, and in December 2010, the Three Gorges dam had experienced three years of trial water impoundment. The water level in the reservoir reached 172 and 175 m after the end of rainy season of 2009 and 2010, respectively. All monitoring data and analytical results show that the operation of the Three Gorges Project is normal. Since its water impoundment from 2003, the Three Gorges Project has shown comprehensive benefits in flood control, power generation, navigation, water supply, and environmental protection, greatly contributing to the sustainable development and steady growth of the Chinese economy. 1.2 Key rock mechanical issues
Key rock mechanical issues of the Three Gorges Project include the sliding resistance and stability of the deep layers of the dam sections of powerhouses No.1–5, the slope stability in the double-line five- stage continuous shiplock, the excavation of large- scale underground powerhouses, and the curtain grouting under the dam. To cope with these technical issues, numerous field geotechnical investigations and scientific experiments were conducted in the 1950s, allowing the engineers to identify the characteristics and fundamental parameters of the rock masses in the dam area. In addition, a number of research institutes, universities, design institutes and several foreign
experts were engaged to the scientific research, computational analyses and optimization design of the project. Moreover, the required productive experiments were conducted, and related feedback information was achieved. With these efforts, systematic and scientific technical requirements for engineering design and construction of the project were formulated [1–6].
2 Sliding resistance and stability of left-bank powerhouses No.1–5
2.1 Solutions The dam sections of powerhouses No.1–5, on the
left bank of the Three Gorges dam site, are located in the riverfront slopes of mountainous region. The host rock masses in the dam foundation are primarily composed of fresh flash-cloud plagioclase granite, containing crevices with long and large low-angle dips with an occurrence of NE-NNE, inclined SE (inclined towards the left bank of the lower reach of the river), as well as a small number of crevices with intermediate inclination towards the lower reach of the river. The powerhouses are arranged behind the dam. The dam section of the powerhouses has a foundation elevation of 90 m. The powerhouses have an elevation of 22.2 m, resulting in a steep slope behind the dam, which has a gradient of 54°, a temporary height of 67.8 m and a permanent height of 39.0 m, as shown in Fig.2. These factors make it possible for the deep layers at the dam section of the left-bank powerhouses No.1–5 to slide.
Fig.2 Typical foundation profile at the dam section of left-bank powerhouses No.1–5 and dam foundation reinforcement (unit: m).
Qixiang Fan et al. / J Rock Mech Geotech Eng. 2011, 3 (4): 329–342 331
To meet the required stability of deep layers in the
dam, the following engineering measures were introduced: (1) slightly reducing the foundation elevation, and setting a cutoff trench at the dam heel; (2) increasing the width of the dam bottom towards the upstream, and moving the curtain water drainage forward; (3) setting up a contact grouting system between the powerhouses and the rock slope of the dam to ensure the integrity of concrete and the rock slope; (4) setting keyways at the cross joints on the dam section of the left-bank powerhouses No.1–5, and filling these joints with grouting to strengthen the overall functions of the dam sections; (5) carrying out solid grouting in certain locations of the tail water channels of the hydropower plant to ensure the functions of the downstream resistance body; (6) adding prestressed anchorage cables on the crevices with long and large low-angle dips from the downstream slope surface; (7) considering a down- stream drainage gallery with an elevation of 26 m below the foundation of the dam section of the left-bank powerhouse No.3; (8) adding 3 000 kN-grade prestressed anchorage cables on the structural plane of crevices with long and large low-angle dips at the left-bank powerhouses No.1–3; and (9) carrying out backfill concrete treatment to the unfavorable rock structural plane that protrudes above the surface of the dam foundation. With these structural design and foundation treatment measures, the dam and its rock foundation have an anti-sliding factor of safety above 4.0, which can meet the design requirements and provide a considerable safety margin. 2.2 Monitoring results of the dam sections of the left-bank powerhouses No.1–5 2.2.1 Monitoring of horizontal displacements at the dam foundation
Figure 3 shows the measured horizontal displacements of the foundation for dam sections of left-bank powerhouses No.1, 3 and 5.
Fig.3 Time-dependent curves of horizontal displacements of the dam foundation from 2001 to 2010.
It can be observed from Fig.3 that before the
commencement of water impoundment in the Three Gorges Reservoir in 2003, horizontal displacements of
the dam foundation ranged from –0.7 to 0.6 mm. After reservoir impoundment, the variation slightly increased, ranging from –0.8 to 3 mm. After the experimental water impoundment in the Three Gorges Reservoir, the water level reached 175 m in 2010, and the variation in horizontal displacements of the dam foundation ranged from 1.89 to 2.69 mm, with an annual change of 0.44–0.88 mm. The horizontal displacements of deep foundation rock masses for the left-bank powerhouses No.1–5, which have an elevation below 95 m, ranged from –1.27 to 3.3 mm, and the increment in the horizontal displacements at the foundation drainage tunnels before and after the experimental impoundments, in 2008 and 2010, ranged from –0.09 to 0.17 mm. The annual increments in displacement are basically consistent. The minimum increment occurred in 2010. Foundation deformation on the dam section of the left-bank powerhouses No.1–5 is converged. 2.2.2 Monitoring of settlements at dam foundation
Figure 4 shows the measured settlements of the upstream and downstream grouting galleries of the left-bank powerhouses No.1, 3 and 5.
(a) Upstream.
(b) Downstream. Fig.4 Time-dependent displacement curves of the upstream and downstream foundation galleries of left-bank powerhouses No.1, 3 and 5.
It can be observed from Fig.4 that the upstream
foundation grouting gallery had a settlement of 18.88–23.48 mm in December 2010, with an annual change of 0.81–1.09 mm, while the downstream foundation gallery had a settlement of 20.83–22.94 mm in December 2010, with an annual change of 1.03–1.32 mm. It is clear that the two adjacent dam sections have approximately the same settlement,
5
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1
1
3
IP02ZC01 (No.1)
EX03ZC033 (No.3)
IP02ZC05 (No.5)
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LD01ZC013 (No.1)
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1997
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typically about 0.5 mm, indicating that there was no uneven settlement in the dam foundation. 2.2.3 Monitoring of deformations inside rock foundation
It can be observed from Fig.5 that horizontal displacements inside the deep rock masses (with a depth of 45–55 m) measured by a clinograph were small. Measuring holes in direction A (upstream and downstream) had a maximum cumulative downstream displacement of 15.6 mm, and the rest had a displacement of –5.8–9.3 mm. The measured values fluctuated around the median value, while measurement had a measuring error of ±6 mm. Instrument IN1CF5 did not detect any dislocation of the rock strata in a hole at depth of 50 m, indicating that the rock masses remained stable.
Fig.5 Time-dependent curves of cumulative horizontal displacements of holes inside the rock foundation of the dam section of left-bank powerhouse No.4.
Displacement of the none-steel-pipe segment of the
left-bank powerhouse No.5 was measured using multi-point borehole extensometers. The typical cumulative displacement curves are shown in Fig.6.
Fig.6 Time-dependent curves of cumulative displacement at each measuring point measured by multi-point borehole extensometer.
It can be observed from Fig.6 that horizontal displacement of rock masses was small at the depth of 35.0 m, and had a springback of less than 1.1 mm in the downstream direction. The other eight boreholes also had a small magnitude of deformation, ranging from 0.6 to 2.3 mm with an annual change of –0.03– 0.06 mm. The displacement curves indicate that the deformation of the rock masses behind the dam has
been basically stabilized since 2001. Statistical analyses of the downstream horizontal
displacements of the left-bank powerhouses No.1 and 5 indicate that the downstream horizontal displacements in the foundation can be divided into two parts, one is approximately 1.4 mm induced by water pressure and another one is less than 2.5 mm induced by temporal effect. The temporal effect component varied mildly, but not completely constrained. It is consistent with the pattern of changes in the temporal effect component at the foundation galleries of right-bank powerhouses No.24–26 and left-bank powerhouse No.14. The incomplete displacement constraint of the temporal effect is largely attributed to the relatively short duration of impoundment in the reservoir. Therefore, further monitoring is needed. Generally, the rock masses at the foundation of the left-bank powerhouses No.1–5 are stable. 2.2.4 Monitoring of deformation at dam crest
During the experimental impoundment, under the effects of temperature drop and water pressure, the crests of the left-bank powerhouses No.1–5 had a maximum downstream horizontal displacement, ranging from –3.07 to 10.32 mm, and the displacement increment in dam crest before and after impoundment ranged from 6.66 to 7.3 mm, which was smaller than the deformation at the crest on the riverbed, but was consistent with the changing patterns. The downstream horizontal displacement of the dam crest was largely attributed to temperature changes. 2.2.5 Monitoring of stresses of anchorage cables
The slopes behind the left-bank powerhouses No.1–5 were reinforced with prestressed anchorage cables and rods. Figure 7 shows the results of the measurements of anchorage cables in pipe trench No.5 of the left-bank powerhouse using a dynamometer.
Fig.7 Time-history curves of force changes in anchorage cable measured by a dynamometer.
It can be observed from Fig.7 that over 1.5 years after installation, the prestressed anchorage cables experienced a considerable loss of prestress. Afterwards, the loss rate decreased, and the prestress change had a close relation with temperature, i.e. rising when temperature went up and falling when
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1.2
0.9
0.6
0.3
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0.3
#1 (5.0 m) #2 (10.0 m)
#3 (25.0 m) #4 (35.0 m)
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19
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1
2 950
2 790
2 630
2 470
2 310
2 150
Loss rate
Anchoring force
For
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kN)
Los
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%)
SF1CF05 No.5 (88.0 m) Tonnage: 3 030.5 kN
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Qixiang Fan et al. / J Rock Mech Geotech Eng. 2011, 3 (4): 329–342 333
temperature went down. After 2006, the prestress remained stable, indicating that the bedrocks behind the dam were stable. 2.2.6 Monitoring of drainage and seepage pressures at foundation
In a global sense, the dam sections of the left-bank powerhouses No.1–5 experienced insignificant changes in seepage pressure before and after experimental impoundment in each year. The seepage increased slightly after experimental impoundment and exhibited an upward tendency. The uphole of drainage tunnel No.1 had a flow of less than 2.0 L/min. It was generally free water, indicating that the dam foundation behind the main drainage curtain at an elevation of 74 m (and upward) was dry. The drainage tunnel had a total seepage rate of less than 135 L/min. The total seepage declined slightly after 2003, and the drainage tunnel had a measured seepage rate of approximately 85.1 L/min on November 30, 2008, 77.5 L/min in January 2010, and 66.85 L/min on October 30, 2010. Preliminary analysis suggests that the steady decline in total seepage at the dam foundation is basically attributed to the following factors: (1) silting was increased in front of the dam; and (2) seepage crevices inside the bedrocks became compacted due to the water pressure from the reservoir basin filled up by silt. The steady decline of total seepage at the foundation of the Three Gorges dam was basically consistent with the variation patterns of seepage in other projects after impoundment. Water levels in the piezometer tube at the left-bank powerhouses No.1–5 before and after experimental impoundment in each year did not show any significant change. The seepage pressure at the dam foundation behind the main drainage curtain on the upstream left-bank powerhouses No.1–5 was below the elevation of 52 m. Actually measured seepage pressure in the deep holes of the piezometer tube in the middle of the foundation of the left-bank powerhouse No.3 was 49–52 m (41–38 m below the foundation surface). No water was discharged into the drainage holes of the drainage tunnel No.1, indicating that the downstream dam foundations of the main drainage curtain and the upstream of drainage tunnel No.2 were basically dry, and that the water level at the dam foundation was underneath the structural plane at a low-angle dip.
Inside the dam foundation, there were crevices with low-angle dips towards the downstream, thus the designers assumed that a potential slip plane at depth may exist inside the dam foundation in terms of
geological conditions. A water pressure distribution chart was prepared on the basis of actual measurements after impoundment at 175 m, as shown in Fig.8. It can be observed from Fig.8 that the measured groundwater level is lower than the slip plane; therefore, no impact will be induced on the stability of the dam body. It also shows that the measured uplift pressure is smaller than the designed value. From 2008 to 2010 after experimental impoundment at 175 m, the measured uplift pressures at the foundation of the left-bank powerhouse No.3 were 48.81%, 49.38% and 49.6% of the designed uplift pressures, respectively. The measured uplift pressure was smaller than the designed uplift pressure. Analysis of data obtained from piezometer tubes on the left-bank powerhouses No.1–5 shows that the uplift pressure coefficients behind the upstream and downstream curtains at the dam foundation are smaller than corresponding designed values of 0.25 and 0.30. The water level between the upstream and downstream drainage tunnels is far below the deep-layer slip plane. The water levels between upstream and downstream drainage tunnels are basically the same and do not change with the rise and fall of the upstream water level. The measured uplift pressures on the two typical deep-layer slip planes are 40%–56% of the designed value.
Fig.8 Distribution of uplift pressure at the foundation of the left-bank powerhouse No.3.
The uplift pressure coefficient at the drainage curtain of the upstream grouting gallery on the left-bank powerhouses No.1–5 ranges from 0 to 0.11, smaller than the designed value of 0.25. The maximum seepage rate at the dam foundation is 67.96 L/min on October 31, 2010, which is smaller than the
I
F C
B
5
4
32
1
Measured pressureMeasured water levelDesigned pressureSliding path
▽185
▽171.26▽175
▽153.06
20+
000.
000
20+
23.0
00
20+
35.0
00
▽136.89
▽106.6▽108
▽90 ▽91.10
▽85 ▽74
A
▽51.89 ▽53.44 ▽50
175 m
▽26.34
▽81.7 ▽75
20+
118.
000
▽65
▽42 ▽36.47
▽65.4
▽23.6 ▽25 ▽23.5
▽0 20+
007.
200
20+
015.
600
20+
11
0.00
0
20+
16
9.75
0 Unit: m
I
334 Qixiang Fan et al. / J Rock Mech Geotech Eng. 2011, 3 (4): 329–342
designed value. 2.3 Integrated estimation
An integrated analysis, from Figs.2–8, suggests that after the Three Gorges Project was completed and began to impound in December 2010, the foundation of the dam section of left-bank powerhouse No.5 had a cumulative maximum displacement of 2.68 mm towards downstream and –0.57 mm towards the left bank. The foundation grouting gallery had a maximum displacement of 23.48 mm. Surveys using clinometers and multi-point borehole extensometers display that rock strata that are 45–55 m below the foundation surface remain approximately stable. The uplift pressure coefficient at the discharge curtain of the upstream foundation grouting gallery and the uplift pressure at the dam foundation are smaller than the corresponding designed values. The groundwater level is lower than the elevation of the slip plane, indicating that all reinforcement measures are effective. Analyses of measured uplift pressure, using the definitive sliding model at a water level of 175 m, and the anti-sliding stability of a typical deep section of the dam show that the deep section of the dam has an anti-sliding stability coefficient slightly higher than originally designed value. The anti-sliding factors of safety under the two most unfavorable definitive sliding models at elevations of 85 m (ABE) and 106.6 m (ABCFI) on the left-bank powerhouse No.3 are 3.37 and 4.20, respectively. Therefore, the anti-sliding stability in the deep layer of the left-bank powerhouses No.1–5 can meet the design requirements [5]. This indicates that, with the engineering measures for the left-bank powerhouses, the anti-sliding stability coefficients of the dam and its rock foundation can satisfy the design requirements and have a considerable safety margin [3].
3 Stability analysis of the high slopes in the double-line five-stage shiplock 3.1 General description
The double-line five-stage shiplock of the Three Gorges Project is hosted in a hill body on the left bank of the Yangtze River. From its top, the hill was cut along its ridge, forming two chute-shaped slopes. The chutes contain the shiplock head and the sidewalls of the chambers. Each chamber has the dimensions of 280 m×34 m×5 m (length × width × minimum water depth), and spacing of the central axes of the chambers is 94 m. The length of main structure of the shiplock is 1 621 m, with total excavation of 55×106 m3. The highest slope reaches 170 m. The vertical walls of the
upper chamber range from 45 to 68 m in height, and a 57 m wide central divider stands between the two lines of chambers.
The high slopes of the shiplock are resultant from a deep cut of the hill, featured with large heights, complex forms, a broad extent and full release of stress, exhibiting heterogeneous characteristics. The stability and small deformation of the slopes are necessary for the normal operation of the miter gate. The complicated geological conditions, including faults and unstable monoliths, made the shiplock construction more difficult. The shiplock construction also encountered the interference of other works and had to be completed with a tight schedule. Excavation of deep and steep chambers with narrow floors had to be carried out in accordance with excavation of underground diversion tunnels and the valve shafts of the shiplock. Moreover, the rock masses of the central divider must keep intact. Both construction and blasting operations must ensure minimal damage to the rock mass and absolute safety [7]. 3.2 Integrated solution to high slope stability in the shiplock
After extensive research, an integrated solution was adopted for the high slopes in the shiplock. The instrument layout of cross-section 17-17 (stake 15+675) of the shiplock is shown in Fig.9 as an example.
(1) Based on an integrated analysis of the engineering geology, hydrological geology, field rock mass dewatering tests, and 2D and 3D underground seepage field tests of the high slopes, a scheme was formulated for the underground drainage design of the high slopes, which played a vital role in ensuring the stability of the high slopes.
(2) According to the characteristics and actual construction conditions of the high slopes, dynamic design approach and methodology were introduced to resolve various technical problems encountered during construction.
(3) Blasting operations were strictly controlled, and bolt bracing plus shotcrete was timely provided. Prestressed anchorage cables were also promptly applied, and systematic high-performance anchorage rod and other construction reinforcement measures were adopted.
(4) To ensure the stability of the rock masses in the vertical slopes, various anchoring schemes were considered. Eventually, 1 000 kN-grade double- protection unbonded anchorage cables (with corrugated tubes) and 3 000 kN-grade double- protection terminal unbonded anchorage cables were used [2].
Qixiang Fan et al. / J Rock Mech Geotech Eng. 2011, 3 (4): 329–342 335
Fig.9 Instrument layout of cross-section 17-17 (stake 15+675) of the shiplock (unit: m).
3.3 Monitoring Generally speaking, experimental impoundment in
the reservoir has not exerted a significant impact on the deformation or seepage of the shiplock. The surface deformation of the slopes in the shiplock, the prestress of the anchorage cables, and the stress of the high-performance anchorage rods all meet the requirements of design, and the measured data display a stable state of the slope. 3.3.1 Displacement monitoring of high slopes and shiplock head
In December 2010, the northern slope of the typical cross-section 17-17 (stake 15+675) in the shiplock had a maximum displacement of 50.18 mm and a minimum of 36.61 mm, while the southern slope had a maximum of 21.15 mm and a minimum of 21.15 mm (Fig.10). The rock masses in the southern and northern
(a) Northern slope.
(b) Southern slope.
Fig.10 Time-history deformation curves of cross-section 17-17 (stake 15+675) towards the chambers in northern and southern slopes of the shiplock.
slopes had maximum displacements of 71.57 and 53.90 mm, respectively, both towards the central axes of the chambers. The southern and northern sides of the central divider had maximum cumulative displacements of 22.01 and 31.30 mm, respectively, while those of the chambers were 6.08 and 4.71 mm, respectively. The shiplock head had a maximum displacement of 5.22 mm towards the chambers, indicating that the shiplock head was approximately stable.
During the process of chamber filling, the shiplock head had a displacement not more than 0.5 mm. Thus, it is believed that the normal operations of the miter gates can be ensured, satisfying the requirements of design.
It can be observed from Fig.10 that deformation of the rock masses in the slopes primarily occurred during excavation and increased with increasing depth and further excavation. The deformation rate dropped gradually after excavation was completed. The current deformation rate is within ±2.5 mm per year, indicating that the rock masses in the slopes are stable.
Figure 11 shows the time-history curves of the maximum displacement of southern and northern slopes in the shiplock. It can be observed that the deformations of the high slopes were largely triggered
Fig.11 Time-history curves of the maximum displacement of southern and northern slopes in the shiplock.
10 0
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by excavation unloading and increased with depth and further excavation. After excavation was completed in September 1999, the deformation rate slowed down and was gradually stabilized. In December 2010, the northern and southern slopes of the shiplock had maximum displacements of 53.9 and 71.57 mm, respectively. After the experimental impoundment to elevation of 175 m in 2010, slope displacement varied within an observational error of ±1.5 mm, indicating that impoundment had no drastic effect on slope deformation. 3.3.2 Monitoring of stresses of anchorage cables for shiplock slopes
Figure 12 shows the time-history curve of stress changes in the anchorage cables at the cross-section 17-17 (stake 15+675) of the northern slope above the shiplock head.
Fig.12 Time-history curve of stress changes in the anchorage cables for cross-section 17-17 (stake 15+675) of the northern slope above the shiplock head.
It can be observed from Fig.12 that the stress
changes can be divided into three stages. The first stage, a stage of rapid stress loss, lasted for about six months, during which the anchorage cables had an average prestress loss of 6.87%. The second stage, a stage of stress fluctuation, lasted for about 2.5 years, during which the anchorage cables had an average prestress loss of 3.71%. The third stage was a stage of periodic stress changes. The prestress increased with increasing temperatures, and vice versa. After two years, the prestress was basically stable. From the measured results, the prestress of the anchorage cables on the high slopes had a loss rate of 2.9%–16.3%. In total, the prestress ranged from 881 to 3 342 kN, with an average loss of 11.3%. 3.3.3 Monitoring of the stresses of anchorage rods for shiplock slopes
Figure 13 shows the time-history curve of stress changes in the anchorage rods for the cross-section 17-17 (stake 15+675) of the southern slope above the shiplock head.
It can be observed from Fig.13 that the stresses of the anchorage rods declined with high-elevation excavation, and became stable after the completion of
Fig.13 Time-history curve of stress changes in the anchorage rods for cross-section 17-17 (stake 15+675) of the southern slope above the shiplock head.
excavation. The stresses of the anchorage rods, ranging from –116.38 to 136.65 MPa, had a negative relation with temperatures, and 52% of the total anchorage rods had a stress less than 50 MPa. The stresses of most anchorage rods fluctuated with temperature, with an annual change of 5.07–57.81 MPa and accumulative stress change of 15 MPa. This indicates that the anchorage rods and cables are effective in reinforcing the high slopes of the shiplock. The displacement has been controlled since the anchorage rods were installed, and the slopes were stable overall [7, 8]. 3.3.4 Monitoring of drainage system of the shiplock
Measures for water interception, prevention and drainage have been considered for the high slopes in the shiplock. Figure 9 shows the distribution of groundwater levels in the piezometer tubes in the drainage tunnels in various layers. Results indicate that almost all of the water levels in the drainage tunnels (from the 2nd to 7th layers) were below the elevation of bottom plates, and that the variation in accumulative water level was less than 2 m. The groundwater level is lower than that proposed in the design for slope stability computations. For example, the osmometers imbedded at the depth of 90–110 m of the drainage tunnel in the 7th layer indicated that the actual water pressure inside the rock masses was merely 40% of the designed value. The rock masses between the groundwater levels and slope excavation faces on both sides of the slopes remain dry, and it is favorable for the stability of the slopes. The groundwater levels also indicated that drainage tunnels with higher elevations had higher groundwater levels and vice versa, which meant that the micropores inside rock masses had poor connectivity. Water levels at the main drainage curtain in the foundation gallery of the shiplock head did not change much, indicating that the water levels were not significantly impacted by the water storage in the reservoir. The granites exhibit the characteristics of unsaturated pore flows. The groundwater levels also demonstrate that the drainage measures, such as drainage tunnels, drainage holes and
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drainage trenches, are effective for the project. Before the impoundment in September 2010, total
seepages in the drainage tunnels on the 1st to 7th layers in the southern and northern slopes were 981.07 and 920.99 L/min, respectively. After the impoundment in October 2010, the seepage flow dropped to 60.08 L/min, indicating that impoundment at elevation of 175 m had no impact on seepage in the high slopes. Seepage in the drainage tunnels was primarily affected by rainfall, and the duration of rainfall reduced agreed well with the period of impoundment. Generally, the seepage of shiplock drainage system exhibited a downward tendency before and after impoundment.
As the shiplock head and the chambers were supported by concrete lining, a crisscross drainage system was arranged at the high slopes and vertical walls behind the lined walls. Monitoring data indicated that the shiplock and the back of the chamber walls had a small seepage pressure, typically with water head less than 1.0 m. The shiplock head and the chambers basically remained dry. Only in a few locations, the measured water head reached 2.4 m, but still much lower than the designed value of 8.0 m. The water head on the bottom was typically below 4.0 m, with the maximum of 6.9 m (the northern line of Chamber 1), indicating that the drainage system was effective. At present, seepage flows in the foundation drainage galleries on the southern and northern lines were 2 537.12 and 2 215.64 L/min, respectively, highest in winter and lowest in summer. The change of seepage flow is related to temperatures, water-proof and impoundment. The back of the shiplock walls typically has a water head less than 1.0 m, and the highest one, as measured in certain location, is 0.023 MPa. The highest seepage pressure on the bottom of the chambers is 0.068 7 MPa.
Actual monitoring data show that the deformation of the rock masses in the slopes primarily occurred during excavation and increased with further excavation. The deformation rate dropped gradually after excavation was completed. The slopes can be seen as stable as all monitoring data are within the designed ranges. 4 Key excavation technologies for underground powerhouses 4.1 Excavation technologies for the Three Gorges underground powerhouses
The main underground powerhouse of the Three Gorges Project is equipped with six 700 MW hydro-turbine generating units. The cross-section of the chamber of the main powerhouse was roughly rectangular with an arch-curved roof. The crest of the chamber is set at the elevation of 105.3 m. Powerhouses below the rock anchorage beams have a span of 31.00 m, and 32.60 m for those above the rock anchorage beams. The dimensions of powerhouse are 87.3 m high and 311.30 m long. The cross-section of underground powerhouse No.4 and distribution of deformation of surrounding rock masses are shown in Fig.14.
The main underground powerhouse is located at the right bank of the Three Gorges. The surrounding rock masses of the powerhouses are primarily fresh plagiogranite and diorite rocks of the Presinian system. Fractures were developed in the rock masses, and the major faults include F20 and F22 with a NNW orientation, and F84 (a fault zone) and F10 with a NE-NEE orientation. The surrounding rock masses of the powerhouses are slightly permeable strata due to faults and strong weathering. Groundwater mainly comes from the upstream reservoir, right hill, atmospheric precipitation, and construction-induced water.
To ensure that the excavation quality of the main powerhouse could meet the design requirements, a series of indoor and field experiments were carried out. Significant results were achieved: (1) The excavation precision of the main powerhouse was ensured by adopting a series of advanced construction technologies. A significant improvement on excavation precision was observed. The average over-excavation at the top arch was controlled at 8.5 cm. More than half of over-excavation was controlled within 20 cm. (2) To further improve the effect of tensile anchorage rods, experiments on grout injection equipment and techniques for anchorage rods, as well as water- cement ratio, were conducted and optimized. As a result, the compactness of cement around the anchorage rods is more than 90%. (3) Some significant breakthroughs were made in environmental protection and occupational health by introducing new concepts and measures. (4) Adding access road, smooth blasting and presplitting blasting was adopted to ensure normal construction of the main powerhouse. These efforts ensured the top arch deformation to be controlled within 2 mm. Thus, the stability of the surrounding rock masses of the large chambers during construction is ensured [10, 11].
338 Qixiang Fan et al. / J Rock Mech Geotech Eng. 2011, 3 (4): 329–342
Fig.14 Cross-section of underground powerhouse No.4 and distribution of deformation of surrounding rock masses (unit: mm).
4.2 Monitoring of underground powerhouse 4.2.1 Monitoring of deformation of surrounding rock masses
Figure 15 shows the time-history curves of displacement and excavation changes in M13DC04.
Fig.15 Time-history curves of displacement and excavation changes in M13DC04.
It can be observed from Fig.15 that the deformation
of surrounding rock masses of the main powerhouse was primarily induced by blasting operations during construction, and it increased with the depth of excavation. After the excavation was completed and surrounding rock masses were reinforced, deformation was totally controlled. At present, deformation of the surrounding rock masses was 5.72–26.16 mm. The deformation of the top arch was 0.45–2.14 mm, within the allowed value of ±10.0 mm in design. The deformation of the upstream arch abutment ranged from 0.68 to 7.79 mm, while that of the downstream arch abutment was 0.20–5.42 mm, mainly of tensile strain. The horizontal deformations of upstream sidewalls at elevations of 86.0 and 76.74 m reached 6.16 and 15.64 mm, respectively. Deformation was considerably large, ranging from 9.67 to 16.67 mm, in the vicinity of downstream sidewalls at elevations of 85.0–87.0 m. To monitor the deformation of unstable
blocks, six multi-point borehole extensometers were set on top of the arch of the powerhouses. The measured displacement was relatively small, ranging from 1.24 to 1.86 mm. 4.2.2 Monitoring of anchorage rod stresses
Figure 16 shows the time-history curves of stress changes in the anchorage rods on the upstream and downstream sidewalls and arch crown of generating unit No.4.
Fig.16 Time-history curves of stress changes in the anchorage rods on the upstream and downstream sidewalls and arch crown of generating unit No.4.
After excavation and supporting work were completed in December 2007, anchorage rod stresses were stabilized and typically below the allowed value of 175 MPa in design. At some locations, however, the stress of anchorage rod exceeded the limit. For example, the stress at elevation of 99.1 m on the upstream arch abutment of generating unit No.30 reached 228.86 MPa, and the tensile stress was about 50 MPa, with a tensile stress increase of 178.86 MPa. Fortunately, the stress was stable thereafter. 47 anchorage rod stress gauges were set in 14 key blocks on the arch crown of the main powerhouse, with stresses ranging from 1.94 to 105.98 MPa.
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4.2.3 Monitoring of anchorage cable stresses Monitoring was carried out from 2006 to 2011 on
the section of generating unit No.1 (the 4th layer) of the upstream sidewalls, key block No.7 of the main powerhouse, the section of generating unit No.4 (the 2nd layer) of the downstream sidewalls. The measured data of downstream keyway of generating unit No.4 were used to analyze the process of changes in anchoring force of the anchorage cables, as shown in Fig.17.
Fig.17 Time-history curves of changes in the anchoring force of prestressed anchorage cables.
From Fig.17, we know that the prestress loss of the
anchorage cables primarily took place in the first year of construction and decreased thereafter, displaying a negative correlation with temperature. Measured data from 34 dynamometers in the main powerhouse showed that the anchorage cables had a prestress of 2 008.0–2 921.2 kN. The total prestress loss was 2.02%–25.47% with an average of 10.96%, and a post-locking average loss was 4.39%, which were all within the design requirement (not larger than 15%). The forces of the anchorage cables were basically below 2 610 kN. Measured data of 10 dynameters installed on the structural blocks of arch crown of the main powerhouse indicated that the forces of the anchorage cables were 1 958.7–2 444.2 kN, with a total loss of 10.22%–27.6%, an average loss of 19.07%, and a post-locking total loss of 7.4%. The forces have been stable since 2006. 4.2.4 Monitoring of rock anchorage beams
Measured data from 31 dynamometers mounted on the upstream and downstream rock anchorage beams for the generators and installation section II in the main powerhouse showed that the anchorage beams had a stress of 225.37–350.62 MPa, and a tensile stress not more than 30 kN. Two dynamometers and 18 joint meters were mounted on the upstream and downstream rock anchorage beams of the main powerhouse. Both of the two dynamometers were in a compression state. At present, the maximum stress was 0.22 MPa. With regard to the opening in the concrete of the rock anchorage beams and the
sidewalls, the opening width of generating unit No.4 was 5.59 mm, while 0.0–0.58 mm at other locations (0.58 mm at arch crown of access tunnel). Currently, the rock anchorage beams on K0+30.0 to K0+471.0, upstream segment of installation section II of the main powerhouse, have an internal temperature of 6.0 °C– 11.5 °C, while those on K0+51.0 to K0+61.0 have an internal temperature of 8.3 °C–13.0 °C. Measured data on the rock anchorage beams showed that the stress ranged from 9.7 to 33.99 MPa, within the allowable limits. 4.2.5 Monitoring of seepage
A total of 56 piezometer tubes were set inside the drainage tunnels around the powerhouse. Water levels in all these tubes were lower than the elevation of the openings, and groundwater levels ranged from 58.99 to 129.05 m. Drainage tunnels at higher elevations had higher groundwater levels, for example, drainage tunnel A had an elevation of approximately 129.0 m and a groundwater level of 114.71–129.05 m, and drainage hole C2 had an elevation of 60.31–67.78 m and a groundwater level of 58.99–67.54 m. At present, groundwater levels and seepages in surrounding rock masses around the underground powerhouse remain stable.
An integrated analysis shows that, in December 2010, the surrounding rock masses of the main structure of the underground powerhouse were stable and deformations of the arch crown and upstream sidewalls were controlled.
5 Grouting technology for the dam foundation curtain 5.1 Profile description
The impervious curtain of the Three Gorges dam consists of upstream main curtain and downstream enclosed curtain. During curtain construction, the depth of holes for the main curtain was generally 60–80 m. The depth of holes on the deep-channel dam sections was 125 m, and the rocks at this point were fractured. Other locations were with strong water impermeability, and the deepest hole was 140 m. The depth of holes for the enclosed curtain typically ranged from 40 to 60 m, but the depth of holes on the deep-channel sections reached 84 m. Layout of the impervious curtain and drainage arrangements on the foundation of a typical dam section are shown in Fig.18. 5.2 Grouting technology for the dam curtain
To ensure a high-quality grouting for the dam curtain, pre-grouting tests were carried out prior to
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Fig.18 Layout of the impervious curtain and drainage arrangements on the foundation of a typical dam section (unit: m). construction, and high-pressure grouting tests were conducted during construction. During dam curtain grouting, a variety of measures were adopted, including pressure-increasing grouting, intense grouting, postponed treatment and re-grouting, acrylate grouting, and high-performance chemical (elastic polyurethane) grouting, etc. During grouting, the complex geological conditions were carefully considered, such as water inrush in the drilling holes, grouting intensity due to water loss, poor adhesion of grouting, strongly weathered strata, pervious deep channels, and weak fault belts. In general, curtain grouting was conducted in three phases, with a total grouting length of approximately 2.0×105 m. The main curtain of the dam had a total grouting length of 1.307×105 m, and the holes in phases I, II and III had an average cement injection of 18.42, 6.90 and 4.49 kg/m, respectively. The enclosed curtain of the dam had a total drilling routing length of 6.81×104 m. In addition, the holes in phases I, II and III had an
average cement injection of 28.32, 9.96 and 6.23 kg/m, respectively.
After grouting was completed, water was filled up for checking. Results showed that the main curtain had a seepage rate of 0.01–0.11 Lu (L/min), and 99.56% of the sections had a seepage rate no more than 1 Lu. For the enclosed curtain, 99.06% of the sections had a seepage rate no more than 1 Lu, which was below the allowable limit of 1 Lu. It indicated that the micro-crevices had been effectively filled up by grouting. Single-hole and cross-hole sound wave tests were conducted. A pre-grouting average velocity of 5 100–5 400 m/s and a post-grouting average velocity of 5 300–5 600 m/s were observed, increased by 2.45%–8.40%. Measured results of large-diameter drilling holes showed that the rocks after cement grouting were compacted, integrated and well-bonded [12, 13].
Table 1 presents the results of an analysis after the dam curtain grouting.
Table 1 Results of analysis of hole permeability of the dam curtain with filled water after grouting.
Seepage rate ≤1 Lu Seepage rate >1 Lu Permeability (Lu) Curtain Location Dam section Hole No. Total section
Sections Frequency
(%) Sections
Frequency (%)
Maximum Average
Left non-overflow section
35 421 416 98.81 5 1.19 3.70 0.09
Powerhouse section 53 773 769 99.48 4 0.52 4.40 0.11 Left bank
Flood discharge section
56 866 862 99.54 4 0.46 1.95 0.31
Right powerhouse section
56 831 831 100 0 0 0.74 0.02 Right bank Right non-overflow
section 3 39 39 100 0 0 0.11 0.01
Main curtain
Total 203 2 930 2 917 99.56 13 0.44 4.40 — Left powerhouse section
21 268 267 99.63 1 0.37 1.47 0.32
Flood discharge section
45 550 529 96.18 21 3.82 4.30 0.63 Left bank
Left powerhouse 26 239 239 100 0 0 0.20 0.02 Right Powerhouse section
59 867 867 100 57 0.94 0.00 — Right bank
Right powerhouse 18 160 160 100 0 0 0.40 0.01
Enclosed curtain
Total 464 6 071 6 014 99.06 44 0.72 0.40 —
Dam axis
Dam axis185.00
Drainage hole (hole distance 300 cm)
185.00
Drainage gallery (300 cm 350 cm)
Grouting hole (hole distance 200 cm,row distance 80 cm)
Drainage gallery
Grouting hole
Discharging section profile No.2
Left powerhouse section profile No.3
120.00
98.00
45.00
20.00
85.00
0.00 Grouting hole (hole distance 250 cm)
80.00
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27.00 66.00
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120.00
108.00
90.00 75.00
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82.00
Drainage gallery (250 cm 300 cm)
50.00
Drainage hole (hole distance 200 cm)
23.00
Drainage gallery(250 cm 300 cm)
25.0015.00
2.0014.00
Qixiang Fan et al. / J Rock Mech Geotech Eng. 2011, 3 (4): 329–342 341
5.3 Monitoring during dam curtain operations 5.3.1 Seepage monitoring
Before and after the impoundment of the reservoir to level 175 m, the seepage rates on the left-bank dam foundation (including that of the powerhouse) were 261.75 and 275.46 L/min, respectively, with an increase of 13.71 L/min. The seepage rates on the right-bank dam foundation (including that of the powerhouse) were 310.58 and 351.33 L/min, respectively, with an increase of 40.75 L/min. Figure 19 shows the time-history curves of seepage rates on the left- and right-bank dam foundations.
(a) Left-bank dam.
(b) Right-bank dam.
Fig.19 Time-history curves of seepage rate on the left- and right-bank dam foundations.
5.3.2 Seepage pressure
Water levels in the piezometer tubes in front of the curtain rose with increasing reservoir water level. After water level reached 175 m, the curtain on the left-bank dam experienced the maximum water rising of 16.13 m to hit 171.58 m, 3.42 m lower than the reservoir water level. The water level behind the curtain fluctuated within a narrow range of 0.27–5.0 m. Water level in front of the curtain on the right-bank dam experienced a maximum increase of 13.26 m to hit 160.87 m. Water level behind the curtain fluctuated within the range of 0.0–1.48 m.
Figure 20 shows the distribution of the uplift pressure coefficients at the drainage curtain of the upstream grouting gallery of the Three Gorges dam. Monitoring results of the impervious curtain of the dam show that all uplift pressure coefficients at the upstream drainage curtain were below the designed limit of 0.25. The maximum uplift pressure coefficients
Fig.20 Distribution of the uplift pressure coefficients at the drainage curtain of the upstream grouting gallery of the Three Gorges dam.
on the left-bank non-overflow dam section, the left-bank dam section, and the flood discharge dam section were 0.18, 0.25 and 0.11, respectively. The maximum uplift pressure at the drainage curtain on section of the dam between the right-bank powerhouse and right-bank dam section No.3 was 0.14. Figure 21 shows the distribution of the uplift pressure coefficients at the downstream drainage curtain. Fig.21 Time-history curves of distribution of uplift pressure coefficients at the downstream drainage curtain on the left diversion-right horizontal dam section.
In Fig.21, all uplift pressure coefficients at the
upstream drainage curtain of the downstream enclosed gallery were smaller than 0.50. The maximum uplift pressure coefficients on the left-bank powerhouse dam section, the flood discharge dam section, the left-bank powerhouse, and the right-bank powerhouse dam section were within the ranges of 0.00–0.37, 0.04–0.34, 0.06–0.30, 0.00–0.37, and 0.08–050, respectively.
Figure 22 shows the distribution of uplift pressure on discharge dam section No.2.
Fig.22 Distribution of uplift pressure on discharge dam section No.2.
On October 26, 2010, when water level reached 175 m,
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the actual uplift pressure on discharge dam section No.2 was 35 555.2 kN/m, 63.31% of the designed value. The actual uplift pressure was smaller than the designed level, helpful for the anti-sliding stability of the dam site. Figure 23 shows the time-history curves of seepage in left- and right-bank dam foundations during the impoundment periods from 2002 to 2010.
Fig.23 Time-history curves of seepage rate at dam foundation.
It can be observed from Fig.23 that the seepage exhibited a downward tendency after water level reached 135 m, possibly due to the effects of silting on the reservoir bottom near the dam. The forefront blocking line of the dam currently had a seepage rate of 614.53 L/min, and those in the left- and right-bank dam foundations were 276.22 and 338.31 L/min, respectively. In total, all uplift pressure coefficients at the drainage curtain in the upstream and downstream enclosed galleries were within the permitted ranges in the design. Actual uplift pressures were smaller than the values allowed, and seepage declined gradually, indicating that the impervious curtain at the foundation of the Three Gorges dam was effective [13, 14].
6 Conclusions
There are many rock mechanical issues in the Three Gorges Project, but the sliding resistance and stability of the dam section of the foundation for powerhouses No.1–5, the slope stability of the double-line five-stage shiplock, excavation for large-scale underground powerhouses, and curtain grouting for the dam are more complicated.
(1) Some special measures were adopted at powerhouses No.1–5. New technologies such as dynamic design approach and methodology, the 1 000 kN-grade double-protection unbonded anchorage cables (with corrugated tubes) and 3 000 kN-grade double-protection unbonded anchorage cables, were used to ensure the stability of the vertical sections of the slopes of the shiplock.
(2) With new slip-casting method, significant breakthroughs in environmental concepts were applied to underground powerhouses to ensure stability of the surrounding rock masses.
(3) Various grouting methods were considered to ensure the quality of dam curtains.
The monitoring data from 16 years of construction experiences and related tests show that the optimized schemes are very successful. It will also contribute to the advancement and development of hydropower technology in the future.
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