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FEDERAL EMERGENCY MANAGEMENT AGENCY FEMA 267b / June, 1999

Interim GuidelinesAdvisory No. 2

Supplement to FEMA-267

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INTERIM GUIDELINES ADVISORY NO. 2Supplement to FEMA-267 Interim Guidelines:

Evaluation, Repair, Modification and Design ofWelded Steel Moment Frame Structures

Report No. SAC-99-01

SAC Joint Venturea partnership of:

Structural Engineers Association of California (SEAOC)Applied Technology Council (ATC)

California Universities for Research in Earthquake Engineering (CUREe)

Prepared for SAC Joint Venture Partnership byGuidelines Development Committee

Ronald O. Hamburger, Chair

John D. HooperRobert E. Shaw

Lawrence D. Reaveley

Thomas SabolC. Mark Saunders

Raymond H.R. Tide

Project Oversight CommitteeWilliam J. Hall, Chair

John N. BarsomShirin Ader

John BarsomRoger Ferch

Theodore V. GalambosJohn Gross

James R. Harris

Richard HolguinNestor IwankiwRoy G. Johnston

Len Joseph Duane K. Miller

John TheissJohn H. Wiggins

SAC Project Management CommitteeSEAOC: William T. HolmesATC: Christoper RojahnCUREe: Robin Shepherd

Program Manager: Stephen A. MahinInvestigations Director: James O. MalleyProduct Director: Ronald O. Hamburger

Federal Emergency Management AgencyProject Officer: Michael Mahoney Technical Advisor: Robert D. Hanson

SAC Joint Venture555 University Avenue, Suite 126

Sacramento, California 95825916-427-3647

June, 1999

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THE SAC JOINT VENTURE

SAC is a joint venture of the Structural Engineers Association of California (SEAOC), the Applied Technology Council(ATC), and California Universities for Research in Earthquake Engineering (CUREe,) formed specifically to address bothimmediate and long-term needs related to solving problems of the Welded Steel Moment Frame (WSMF) connection thatbecame apparent as a result of the 1994 Northridge earthquake. SEAOC is a professional organization composed ofmore than 3,000 practicing structural engineers in California. The volunteer efforts of SEAOC’s members on varioustechnical committees have been instrumental in the development of the earthquake design provisions contained in theUniform Building Code as well as the National Earthquake Hazards Reduction Program (NEHRP) Provisions for SeismicRegulations for New Buildings. The Applied Technology Council is a non-profit organization founded specifically toperform problem-focused research related to structural engineering and to bridge the gap between civil engineeringresearch and engineering practice. It has developed a number of publications of national significance including ATC 3-06, which serves as the basis for the NEHRP Recommended Provisions. CUREe is a nonprofit organization formed topromote and conduct research and educational activities related to earthquake hazard mitigation. CUREe’s eightinstitutional members are: the California Institute of Technology, Stanford University, the University of California atBerkeley, the University of California at Davis, the University of California at Irvine, the University of California at LosAngeles, the University of California at San Diego, and the University of Southern California. This collection ofuniversity earthquake research laboratory, library, computer and faculty resources is among the most extensive in theUnited States. The SAC Joint Venture allows these three organizations to combine their extensive and unique resources,augmented by subcontractor universities and organizations from around the nation, into an integrated team ofpractitioners and researchers, uniquely qualified to solve problems related to the seismic performance of WSMFstructures.

DISCLAIMER

The purpose of this document is to serve as a supplement to the FEMA-267 publication Interim Guidelines: Evaluation,Repair, Modification and Design of Welded Steel Moment Frame Structures. This Advisory, which is intended to be usedin conjunction with FEMA-267, supercedes and entirely replaces Interim Guidelines Advisory No. 1 (FEMA 267a). FEMA-267 was published to provide engineers and building officials with guidance on engineering procedures forevaluation, repair, modification and design of welded steel moment frame structures, to reduce the risks associated withearthquake-induced damage. The recommendations were developed by practicing engineers based on professionaljudgment and experience and a preliminary program of laboratory, field and analytical research. This preliminaryresearch, known as the SAC Phase 1 program, commenced in November, 1994 and continued through the publication ofthe Interim Guidelines document. This Interim Guidelines Advisory No. 2, which updates and replaces InterimGuidelines Advisory No. 1, is based on supplementary data developed under a program of continuing research, known asthe SAC Phase 2 program, as well as findings developed by other, independent researchers. Final designrecommendations, superceding both FEMA-267 and this document are scheduled for publication in early 2000. Independent review and guidance in the production of both the FEMA-267, Interim Guidelines and the advisories wasprovided by a project oversight panel comprised of experts from industry, practice and academia. Users are cautioned thatresearch into the behavior of these structures is continuing. Interpretation of the results of this research may invalidate orsuggest the need for modification of recommendations contained herein. No warranty is offered with regard to therecommendations contained herein, either by the Federal Emergency Management Agency, the SAC JointVenture, the individual joint venture partners, their directors, members or employees. These organizations andtheir employees do not assume any legal liability or responsibility for the accuracy, completeness, or usefulness ofany of the information, products or processes included in this publication. The reader is cautioned to carefullyreview the material presented herein. Such information must be used together with sound engineering judgment whenapplied to specific engineering projects. This Interim Guidelines Advisory has been prepared by the SAC Joint Venturewith funding provided by the Federal Emergency Management Agency, under contract number EMW-95-C-4770. TheSAC Joint Venture gratefully acknowledges the support of FEMA and the leadership of Michael Mahoney and RobertHanson, Project Officer and Technical Advisor, respectively. The SAC Joint Venture also wishes to express its gratitudeto the large numbers of engineers, building officials, organizations and firms that provided substantial efforts, materials,and advice and who have contributed significantly to the progress of the Phase 2 effort.

Interim Guidelines Advisory No. 2 SAC 99-01

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PREFACE

Purpose

The purpose of the Interim Guidelines Advisory series is to provide engineers and buildingofficials with timely information and guidance resulting from ongoing problem-focused studies ofthe seismic behavior of moment-resisting steel frame structures. These advisories are intended tobe supplements to FEMA-267 Interim Guidelines: Evaluation, Repair, Modification and Designof Welded Steel Moment Frame Structures first published in August 1995.

The first Interim Guidelines Advisory, FEMA-267a, was published in January 1997. Thespecific revisions and updates to the Interim Guidelines contained in FEMA-267a were developedbased on input obtained from a group of engineers and building officials actively engaged in theuse of the FEMA-267 document, in the period since its initial publication in August 1995. Thatinput was obtained during a workshop held in August 1996, in Los Angeles, California.

This second Interim Guidelines Advisory has been prepared as a series of updates andrevisions both to the FEMA-267, Interim Guidelines which it supplements and to the FEMA-267a, Interim Guidelines Advisory publication, which it supercedes. The material contained inthis Interim Guidelines Advisory No. 2 is based on the extensive analytical and laboratoryresearch that has been conducted by the SAC Joint Venture and other researchers during theintervening period, along with recent developments in the steel construction industry. Thematerial contained in this Advisory has been formatted to match that contained in the originalInterim Guidelines, to permit the user to insert this material directly into appropriate sections ofthat document. This Advisory is not intended to serve as a self-contained text and should not beused as such. It does, however, completely replace the material contained in FEMA-267a.

A new set of recommendations for the design, analysis, evaluation repair, retrofit andconstruction of moment-resisting steel frames is currently being prepared as part of the Phase 2Program to Reduce Earthquake Hazards in Steel Moment Frame Structures. These new SeismicDesign Criteria, which are anticipated to be completed early in the year 2000, will replace in theirentirety the FEMA-267 Interim Guidelines and this Interim Guidelines Advisory No. 2.

Background

The Northridge earthquake of January 17, 1994, dramatically demonstrated that theprequalified, welded beam-to-column moment connection commonly used in the construction ofwelded steel moment resisting frames (WSMFs) in the period 1965-1994 was much moresusceptible to damage than previously thought. The stability of moment frame structures inearthquakes is dependent on the capacity of the beam-column connection to remain intact and toresist tendencies of the beams and columns to rotate with respect to each other under theinfluence of lateral deflection of the structure. The prequalified connections were believed to beductile and capable of withstanding the repeated cycles of large inelastic deformation explicitlyrelied upon in the building code provisions for the design of these structures. Although manyaffected connections were not damaged, a wide spectrum of unexpected brittle connection


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fractures did occur, ranging from isolated fractures through or adjacent to the welds of beamflanges to columns, to large fractures extending across the full depth of the columns. At the timethis damage was discovered, the structural steel industry and engineering profession had littleunderstanding of the specific causes of this damage, the implications of this damage for buildingsafety, or even if reliable methods existed to repair the damage which had been discovered. Although the connection failures did not result in any casualties or collapses, and many WSMFbuildings were not damaged, the incidence of damage was sufficiently pervasive in regions ofstrong ground motion to cause wide-spread concern by structural engineers and building officialswith regard to the safety of these structures in future earthquakes.

In response to these concerns, the Federal Emergency Management Agency (FEMA) enteredinto a cooperative agreement with the SAC Joint Venture to perform problem-focused study ofthe seismic performance of welded steel moment connections and to develop interimrecommendations for professional practice. Specifically, these recommendations were intended toaddress the inspection of earthquake affected buildings to determine if they had sustainedsignificant damage; the repair of damaged buildings; the upgrade of existing buildings to improvetheir probable future performance; and the design of new structures to provide more reliableseismic performance. Within weeks of receipt of notification of FEMA’s intent to enter into thisagreement, the SAC Joint Venture published a series of two design advisories (SAC, 1994a; SAC,1994b). These design advisories presented a series of papers, prepared by engineers andresearchers engaged in the investigation of the damaged structures and presenting individualopinions as to the causes of the damage, potential methods of repair, and possible designs formore reliable connections in the future. In February 1995, Design Advisory No. 3 (SAC, 1995a)was published. This third advisory presented a synthesis of the data presented in the earlierpublications, together with the preliminary recommendations developed in an industry workshop,attended by more than 50 practicing engineers, industry representatives and researchers, onmethods of inspecting, repairing and designing WSMF structures. At the time this third advisorywas published, significant disagreement remained within the industry and the profession as to thespecific causes of the damage observed and appropriate methods of repair given that the damagehad occurred. Consequently, the preliminary recommendations were presented as a series of issuestatements, followed by the consensus opinions of the workshop attendees, where consensusexisted, and by majority and dissenting opinions where such consensus could not be formed.

During the first half of 1995, an intensive program of research was conducted to moredefinitively explore the pertinent issues. This research included literature surveys, data collectionon affected structures, statistical evaluation of the collected data, analytical studies of damagedand undamaged buildings and laboratory testing of a series of full-scale beam-column assembliesrepresenting typical pre-Northridge design and construction practice as well as various repair,upgrade and alternative design details. The findings of this research (SAC 1995c, SAC 1995d,SAC 1995e, SAC 1995f, SAC 1995g, SAC 1996) formed the basis for the development of FEMA267 - Interim Guidelines: Evaluation, Repair, Modification, and Design of Welded Steel MomentFrame Structures (SAC, 1995b), which was published in August, 1995. FEMA 267 provided thefirst definite, albeit interim, recommendations for practice, following the discovery of connectiondamage in the Northridge earthquake.


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As a result of these and supplemental studies conducted by the SAC Joint Venture, as well asindependent research conducted by others, it is now known that a large number of factorscontributed to the damage sustained by steel frame buildings in the Northridge earthquake. Theseincluded:

• design practice that favored the use of relatively few frame bays to resist lateralseismic demands, resulting in much larger member and connection geometries than hadpreviously been tested;

• standard detailing practice which resulted in the development of large inelasticdemands at the beam to column connections;

• detailing practice that often resulted in large stress concentrations in the beam-columnconnection, as well as inherent stress risers and notches in zones of high stress;

• the common use of welding procedures that resulted in deposition of low toughnessweld metal in the critical beam flange to column flange joints;

• relatively poor levels of quality control and assurance in the construction process,resulting in welded joints that did not conform to the applicable quality standards;

• excessively weak and flexible column panel zones that resulted in large secondarystresses in the beam flange to column flange joints;

• large variations in the strengths of rolled shape members relative to specified values;

• an inherent inability of material to yield under conditions of high tri-axial restraint suchas exist at the center of the beam flange to column flange joints.

With the identification of these factors it was possible for FEMA 267 to present arecommended methodology for the design and construction of moment-resisting steel frames toprovide connections capable of more reliable seismic performance. This methodology includedthe following recommendations:

• proportion the beam-column connection such that inelastic behavior occurs at adistance remote from the column face, minimizing demands on the highly restrainedcolumn material and the welded joints;

• specify weld filler metals with rated toughness values for critical welded joints;

• detail connections to incorporate beam flange continuity plates, to minimize stressconcentrations;

• remove backing bars and weld tabs from critical joints to minimize the potential forstress risers and notch effects and also to improve the reliability with which flaws atthe weld root can be observed and repaired;


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• qualify connection configurations through a program of full-scale inelastic testing ofrepresentative beam-column assemblies, fabricated in the same manner as is proposedfor use in the structure;

• increased participation of the design professional in the specification and surveillanceof welding procedures and the quality assurance process for welded joints.

In the time since the publication of FEMA-267, SAC has continued, under funding providedby FEMA, to perform problem-focused study of the performance of moment resisting connectionsof various configurations. This work, which is generally referred to as the SAC Phase II program,includes detailed analytical evaluations of buildings and connections, parametric studies into theeffects on connection performance of connection configuration, base and weld metal strength,toughness and ductility, as well as additional large scale testing of connection assemblies. Theintent of this study is to support development of final guidelines that will present more reliable andeconomical performance-based methods for:

• identification of damaged structures following an earthquake and determination of theextent, severity and consequences of such damage;

• design of effective repairs for damaged structures;

• identification of existing structures that are vulnerable to unacceptable levels ofdamage in future earthquakes;

• design of structural upgrades for existing vulnerable structures;

• design of new structures that are suitably resistant to earthquake induced damage;

• procedures for construction quality assurance that are consistent with the levels ofreliability intended by the design criteria.

This Phase II program of research, which is being conducted by the SAC Joint Venture inparallel and coordination with work by other researchers, is anticipated to be complete in late1999. It is the intent of FEMA and the SAC Joint Venture to ensure that pertinent informationand findings from this program are made available to the user community in a timely mannerthrough the publication of this series of design advisory documents. This Interim GuidelinesAdvisory No. 2 is the second such publication.

Format

This Advisory has been prepared as a series of updates and revisions to the FEMA-267,Interim Guidelines publication. It has been formatted in a manner intended to facilitate theidentification of changes to the original FEMA-267 text. Only those sections of FEMA-267 thatare being revised at this time are included. Other sections of FEMA-267 remain in effect as thecurrent best recommendations of the SAC Joint Venture. This Advisory replaces the earlierInterim Guidelines Advisory, FEMA-267a, in its entirety.


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To facilitate coordination of this Advisory with FEMA-267, the existing system of chapter andsection numbering has been retained. The Table of Contents lists all sections of the chaptersbeing revised, including those sections for which no revisions are included. Within the body ofthis document, a section heading is provided for each section of the chapter; however, if norevision to the section is currently being made, this is indicated immediately beneath the sectionheading.

To facilitate reading of this document, where a revision is made to a section in FEMA 267, theentire text of that section is included herein. Where existing text from FEMA-267 is reproducedin this document, without edit, it is shown in normal face type for guidelines, and in italicized typefor commentary. Where existing text is being deleted, this is shown in strike through format. Asingle strikethrough indicates text deleted in the first advisory, FEMA-267a. A doublestrikethrough indicates text deleted in this current advisory. New text is shown in underlineformat. A single underline identifies text added in the first advisory, FEMA-267a. A doubleunderline identifies text added in this current advisory. When a modification has been made to aportion of text, relative to FEMA-267, this will also be noted by the presence of a vertical line atthe outside margin of the page. The following two paragraphs illustrate these conventions forguideline and commentary text, respectively.

This sentence is representative of typical guideline text, that has been reprintedfrom FEMA-267 without change.This sentence, is representative of the way inwhich text being deleted from FEMA-267 in this Interim Guidelines Advisory isidentified. This sentence illustrates the way in which text deleted from FEMA-267in the previous Interim Guidelines Advisory is identified. This sentence illustratesthe way in which text being added to FEMA-267 in this Interim GuidelinesAdvisory is identified.This sentence illustrates the way in which text added toFEMA-267 in the previous Interim Guidelines Advisory is identified.

Commentary: This sentence is representative of typical commentary text, that hasbeen reprinted from FEMA-267 without change. This sentence is representative ofthe way in which commentary text being deleted from FEMA-267 in this InterimGuidelines Advisory is identified. However, this sentence, is representative of theway in which text being deleted from FEMA-267 commentary in the previousAdvisory is identified. This sentence indicates the way in which text added to theFEMA-267 commentary in this Advisory is shown.This final sentence illustratesthe way in which text added in previous advisory, FEMA-267a, is identified.

Intent

This Interim Guidelines Advisory, together with the Interim Guidelines they modify, are primarilyintended for two different groups of potential users:

a) Engineers engaged in evaluation, repair, and upgrade of existing WSMF buildings and inthe design of new WSMF buildings incorporating either Special Moment-Resisting Framesor Ordinary Moment-Resisting Frames utilizing welded beam-column connections. The


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recommendations for new construction are applicable to all WSMF construction expectedto resist earthquake demands through plastic behavior.

b) Regulators and building departments responsible for control of the evaluation, repair, andoccupancy of WSMF buildings that have been subjected to strong ground motion and forregulation of the design, construction, and inspection of new WSMF buildings.

The fundamental goal of the information presented in the Interim Guidelines as modified by thisAdvisory is to help identify and reduce the risks associated with earthquake-induced fractures inWSMF buildings through provision of timely information on how to inspect existing buildings fordamage, repair damage if found, upgrade existing buildings and design new buildings. The informationpresented here primarily addresses the issue of beam-to-column connection integrity under the severeinelastic demands that can be produced by building response to strong ground motion. Users arereferred to the applicable provisions of the locally prevailing building code for information with regardto other aspects of building construction and earthquake damage control.

Limitations

The information presented in this Interim Guidelines Advisory, together with that contained in theInterim Guidelines it modifies, is based on limited research conducted since the NorthridgeEarthquake, review of past research and the considerable experience and judgment of the professionalsengaged by SAC to prepare and review this document. Additional research on such topics as the effectof floor slabs on frame behavior, the effect of weld metal and base metal toughness, the efficacy ofvarious beam-column connection details and the validity of current standard testing protocols forprediction of earthquake performance of structures is continuing as part of the Phase 2 program and isexpected to provide important information not available at the time this Advisory was formulated. Therefore, many of the recommendations cited herein may change as a result of forthcoming researchresults.

The recommendations presented herein represent the group consensus of the committee ofGuideline Writers retained by SAC following independent review by the Project OversightCommittee. They may not reflect the individual opinions of any single participant. They do notnecessarily represent the opinions of the SAC Joint Venture, the Joint Venture partners, or thesponsoring agencies. Users are cautioned that available information on the nature of the WSMFproblem is in a rapid stage of development and any information presented herein must be usedwith caution and sound engineering judgment.


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TABLE OF CONTENTSTHE SAC JOINT VENTURE iiDISCLAIMER iiPREFACE iii

Purpose iiiBackground iiiFormat viIntent viiLimitations viii

1 INTRODUCTION1.1 Purpose 1-11.2 Scope 1-11.3 Background 1-11.4 The SAC Joint Venture 1-81.5 Sponsors 1-81.6 Summary of Phase I Research 1-81.7 Intent 1-81.8 Limitations 1-91.9 Use of the Guidelines 1-9

3 CLASSIFICATIONS AND IMPLICATIONS OF DAMAGE3.1 Summary of Earthquake Damage 3-13.2 Damage Types 3-1

3.2.1 Girder Damage 3-13.2.2 Column Flange Damage 3-13.2.3 Weld Damage, Defects and Discontinuities 3-13.2.4 Shear Tab Damage 3-43.2.5 Panel Zone Damage 3-43.2.6 Other Damage 3-4

3.3 Safety Implications 3-53.4 Economic Implications 3-7

4 POST-EARTHQUAKE EVALUATION4.1 Scope 4-14.2 Preliminary Evaluation 4-1

4.2.1 Evaluation Process 4-14.2.1.1 Ground Motion 4-14.2.1.2 Additional Indicators 4-1

4.2.2 Evaluation Schedule 4-14.2.3 Connection Inspections 4-2

4.2.3.1 Analytical Evaluation 4-24.2.3.2 Buildings with Enhanced Connections 4-3

4.2.4 Previous Evaluations and Inspections 4-3


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4.3 Detailed Evaluation Procedure 4-34.3.1 Eight Step Inspection and Evaluation Procedure 4-34.3.2 Step 1 - Categorize Connections By Group 4-44.3.3 Step 2 - Select Samples of Connections for Inspection 4-4

4.3.3.1 Method A - Random Selection 4-54.3.3.2 Method B - Deterministic Selection 4-54.3.3.3 Method C - Analytical Selection 4-5

4.3.4 Step 3- Inspect the Selected Samples of Connections 4-54.3.4.1 Damage Characterization 4-5

4.3.5 Step 4 - Inspect Connections Adjacent to Damaged Connections 4-84.3.6 Step 5 - Determine Average Damage Index for the Group 4-84.3.7 Step 6 - Determine the Probability that the Connections in a

Group at a Floor Level Sustained Excessive Damage 4-94.3.7.1 Some Connections In Group Not Inspected 4-94.3.7.2 All Connections in Group Inspected 4-9

4.3.8 Step 7 - Determine Recommended RecoveryStrategies for the Building 4-9

4.3.9 Step 8 - Evaluation Report 4-94.4 Alternative Group Selection for Torsional Response 4-94.5 Qualified Independent Engineering Review 4-9

4.5.1 Timing of Independent Review 4-94.5.2 Qualifications and Terms of Employment 4-94.5.3 Scope of Review 4-94.5.4 Reports 4-94.5.5 Responses and Corrective Actions 4-104.5.6 Distribution of Reports 4-104.5.7 Engineer of Record 4-104.5.8 Resolution of Differences 4-10

5 POST-EARTHQUAKE INSPECTION5.1 Connection Types Requiring Inspection 5-1

5.1.1 Welded Steel Moment Frame (WSMF) Connections 5-15.1.2 Gravity Connections 5-35.1.3 Other Connection Types 5-3

5.2 Preparation 5-45.2.1 Preliminary Document Review and Evaluation 5-4

5.2.1.1 Document Collection and Review 5-45.2.1.2 Preliminary Building Walk-Through 5-45.2.1.3 Structural Analysis 5-45.2.1.4 Vertical Plumbness Check 5-4

5.2.2 Connection Exposure 5-45.3 Inspection Program 5-6

5.3.1 Visual Inspection (VI) 5-65.3.1.1 Top Flange 5-65.3.1.2 Bottom Flange 5-6


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5.3.1.3 Column and Continuity Plates 5-65.3.1.4 Beam Web Shear Connection 5-7

5.3.2 Nondestructive Testing (NDT) 5-75.3.3 Inspector Qualification 5-95.3.4 Post-Earthquake Field Inspection Report 5-95.3.5 Written Report 5-9

6 POST-EARTHQUAKE REPAIR AND MODIFICATION6.1 Scope 6-16.2 Shoring 6-16.3 Repair Details 6-16.4 Preparation 6-16.5 Execution 6-16.6 Structural Modification 6-1

6.6.1 Definition of Modification 6-16.6.2 Damaged vs. Undamaged Connections 6-16.6.3 Criteria 6-16.6.4 Strength and Stiffness 6-4

6.6.4.1 Strength 6-46.6.4.2 Stiffness 6-6

6.6.5 Plastic Rotation Capacity 6-76.6.6 Connection Qualification and Design 6-10

6.6.6.1 Qualification Test Protocol 6-116.6.6.2 Acceptance Criteria 6-116.6.6.3 Calculations 6-12

6.6.6.3.1 Material Strength Properties 6-136.6.6.3.2 Determine Plastic Hinge Location 6-166.6.6.3.3 Determine Probable Plastic Moment at Hinges 6-186.6.6.3.4 Determine Beam Shear 6-196.6.6.3.5 Determine Strength Demands on Connection 6-206.6.6.3.6 Check Strong Column - Weak Beam Conditions 6-216.6.6.3.7 Check Column Panel Zone 6-23

6.6.7 Modification Details 6-246.6.7.1 Haunch at Bottom Flange 6-246.6.7.2 Top and Bottom Haunch 6-266.6.7.3 Cover Plate Sections 6-266.6.7.4 Upstanding Ribs 6-286.6.7.5 Side-Plate Connections 6-296.6.7.6 Bolted Brackets 6-29

7 NEW CONSTRUCTION7.1 Scope 7-17.2 General - Welded Steel Frame Design Criteria 7-3

7.2.1 Criteria 7-37.2.2 Strength and Stiffness 7-4


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7.2.2.1 Strength 7-47.2.2.2 Stiffness 7-5

7.2.3 Configuration 7-67.2.4 Plastic Rotation Capacity 7-97.2.5 Redundancy 7-137.2.6 System Performance 7-157.2.7 Special Systems 7-15

7.3 Connection Design and Qualification Procedures - General 7-157.3.1 Connection Performance Intent 7-157.3.2 Qualification by Testing 7-167.3.3 Design by Calculation 7-16

7.4 Guidelines for Connection Qualification by Testing 7-167.4.1 Testing Protocol 7-167.4.2 Acceptance Criteria 7-16

7.5 Guidelines for Connection Design by Calculation 7-187.5.1 Material Strength Properties 7-187.5.2 Design Procedure - Strengthened Connections 7-23

7.5.2.1 Determine Plastic Hinge Locations 7-237.5.2.2 Determine Probable Plastic Moment at Hinge 7-247.5.2.3 Determine Shear at Plastic Hinge 7-267.5.2.4 Determine Strength Demands at Critical Sections 7-267.5.2.5 Check for Strong Column - Weak Beam Condition 7-277.5.2.6 Check Column Panel Zone 7-29

7.5.3 Design Procedure - Reduced Beam Section Connections 7-307.5.3.1 Determine Reduced Section and Plastic Hinge Locations 7-337.5.3.2 Determine Strength and Probable Plastic Moment in RBS 7-337.5.3.3 Strong Column - Weak Beam Condition 7-357.5.3.4 Column Panel Zone 7-367.5.3.5 Lateral Bracing 7-367.5.3.6 Welded Attachments 7-37

7.6 Metallurgy & Welding 7-387.7 Quality Control / Quality Assurance 7-387.8 Guidelines on Other Connection Design Issues 7-38

7.8.1 Design of Panel Zones 7-397.8.2 Design of Web Connections to Column Flanges 7-397.8.3 Design of Continuity Plates 7-407.8.4 Design of Weak Column and Weak Way Connections 7-40

7.9 Moment Frame Connections for Consideration in New Construction 7-407.9.1 Cover Plate Connections 7-407.9.2 Flange Rib Connections 7-437.9.3 Bottom Haunch Connections 7-447.9.4 Top and Bottom Haunch Connections 7-467.9.5 Side-Plate Connections 7-467.9.6 Reduced Beam Section Connections 7-467.9.7 Slip-Friction Energy Dissipating Connections 7-48


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7.9.8 Column Tree Connections 7-487.9.9 Slotted Web Connections 7-487.9.10 Bolted Bracket Connections 7-50

7.10 Other Types of Welded Connection Structures 7-527.10.1 Eccentrically Braced Frames (EBF) 7-527.10.2 Dual Systems 7-527.10.3 Welded Base Plate Details 7-527.10.4 Vierendeel Truss Systems 7-527.10.5 Moment Frame Tubular Systems 7-527.10.6 Welded Connections of Collectors, Ties and Diaphragm Chords 7-537.10.7 Welded Column Splices 7-537.10.8 Built-up Moment Frame Members 7-53

8 METALLURGY & WELDING

8.1 Parent Materials 8-18.1.1 Steels 8-18.1.2 Chemistry 8-38.1.3 Tensile/Elongation Properties 8-38.1.4 Toughness Properties 8-108.1.5 Lamellar Discontinuities 8-108.1.6 K-Area Fractures 8-10

8.2 Welding 8-118.2.1 Welding Process 8-118.2.2 Welding Procedures 8-128.2.3 Welding Filler Metals 8-138.2.4 Preheat and Interpass Temperatures 8-178.2.5 Postheat 8-178.2.6 Controlled Cooling 8-178.2.7 Metallurgical Stress Risers 8-178.2.8 Welding Preparation & Fit-up 8-17

12. REFERENCES 12-1


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Introduction

1-1

1. INTRODUCTION

1.1 Purpose

There are no modifications to the Guidelines or Commentary of Section 1.1 at this time.

1.2 Scope


1.3 Background

Following the January 17, 1994 Northridge, California Earthquake, more than 100 steel buildingswith welded moment-resisting frames were found to have experienced beam-to-column connectionfractures. The damaged structures cover a wide range of heights ranging from one story to 26 stories;and a wide range of ages spanning from buildings as old as 30 years of age to structures just beingerected at the time of the earthquake. The damaged structures are were spread over a largegeographical area, including sites that experienced only moderate levels of ground shaking. Althoughrelatively few such buildings were located on sites that experienced the strongest ground shaking,damage to these buildings was quite severe. Discovery of these extensive connection fractures, oftenwith little associated architectural damage to the buildings, was has been alarming. The discovery hasalso caused some concern that similar, but undiscovered damage may have occurred in other buildingsaffected by past earthquakes. Indeed, there are now confirmed isolated reports of such damage. Inparticular, a publicly owned building at Big Bear Lake is known to have been was damaged by theLanders-Big Bear, California sequence of earthquakes, and at least one building, under construction inOakland, California at the time fo the several buildings were damaged during the 1989 Loma PrietaEarthquake, was reported to have experienced such damage in the San Francisco Bay Area.

WSMF construction is used commonly throughout the United States and the world, particularlyfor mid- and high-rise construction. Prior to the Northridge Earthquake, this type of construction wasconsidered one of the most seismic-resistant structural systems, due to the fact that severe damage tosuch structures had rarely been reported in past earthquakes and there was no record of earthquake-induced collapse of such buildings, constructed in accordance with contemporary US practice.However, the widespread severe structural damage which occurred to such structures in theNorthridge Earthquake calleds for re-examination of this premise.

The basic intent of the earthquake resistive design provisions contained in the building codes is toprotect the public safety, however, there is also an intent to control damage. The developers of thebuilding code provisions have explicitly set forth three specific performance goals for buildingsdesigned and constructed to the code provisions (SEAOC - 1990). These are to provide buildings withthe capacity to

• resist minor earthquake ground motion without damage;


Introduction

1-2

• resist moderate earthquake ground motion without structural damage but possibly somenonstructural damage; and

• resist major levels of earthquake ground motion, having an intensity equal to the strongesteither experienced or forecast for the building site, without collapse, but possibly with somestructural as well as nonstructural damage.

In general, WSMF buildings in the Northridge Earthquake met the basic intent of the buildingcodes, to protect life safety. However, the ground shaking intensity experienced by most of thesebuildings was significantly less than that anticipated by the building codes. Many buildings thatexperienced moderate intensity ground shaking experienced significant damage that could be viewed asfailing to meet the intended performance goals with respect to damage control. Further, somemembers of the engineering profession (SEAOC - 1995b) and government agencies (Seismic SafetyCommission - 1995) have stated that even these performance goals are inadequate for society’s currentneeds.

WSMF buildings are designed to resist earthquake ground shaking based on the assumption thatthey are capable of extensive yielding and plastic deformation, without loss of strength. The intendedplastic deformation is intended to be developed through a combination of consists of plastic rotationsdeveloping within the beams, at their connections to the columns, and plastic shear yielding of thecolumn panel zones,. and is tTheoretically these mechanisms should be capable of resulting in benigndissipation of the earthquake energy delivered to the building. Damage is expected to consist ofmoderate yielding and localized buckling of the steel elements, not brittle fractures. Based on thispresumed behavior, building codes require a minimum lateral design strength for WSMF structures thatis approximately 1/8 that which would be required for the structure to remain fully elastic. Supplemental provisions within the building code, intended to control the amount of interstory driftsustained by these flexible frame buildings, typically result in structures which are substantially strongerthan this minimum requirement and in zones of moderate seismicity, substantial overstrength may bepresent to accommodate wind and gravity load design conditions. In zones of high seismicity, mostsuch structures designed to minimum code criteria will not start to exhibit plastic behavior until groundmotions are experienced that are 1/3 to 1/2 the severity anticipated as a design basis. This designapproach has been developed based on historical precedent, the observation of steel buildingperformance in past earthquakes, and limited research that has included laboratory testing of beam-column models, albeit with mixed results, and non-linear analytical studies.

Observation of damage sustained by buildings in the Northridge Earthquake indicates that contraryto the intended behavior, in some many cases brittle fractures initiated within the connections at verylow levels of plastic demand, and in some cases, while the structures remained essentially elastic. Typically, but not always, fractures initiated at, or near, the complete joint penetration (CJP) weldbetween the beam bottom flange and column flange (Figure 1-1). Once initiated, these fracturesprogressed along a number of different paths, depending on the individual joint and stress conditions. Figure 1-1 indicates just one of these potential fracture growth patterns. Investigators initially identifieda number of factors which may have contributed to the initiation of fractures at the weld root including:notch effects created by the backing bar which was commonly left in place following joint completion;sub-standard welding that included excessive porosity and slag inclusions as well as incomplete fusion;


Introduction

1-3

and potentially, pre-earthquake fractures resulting from initial shrinkage of the highly restrained weldduring cool-down. Such problems could be minimized in future construction, with the application ofappropriate welding procedures and more careful exercise of quality control during the constructionprocess. However, it is now known that these were not the only causes of the fractures whichoccurred.

Backing bar

Column flange

Beam flange

Fused zone

Fracture

Figure 1-1 - Common Zone of Fracture Initiation in Beam -Column Connection

Current production processes for structural steel shapes result in inconsistent strength anddeformation capacities for the material in the through-thickness direction. Non-metallic inclusions inthe material, together with anisotropic properties introduced by the rolling process can lead to lamellarweakness in the material. Further, the distribution of stress across the girder flange, at the connectionto the column is not uniform. Even in connections stiffened by continuity plates across the panel zone,significantly higher stresses tend to occur at the center of the flange, where the column web produces alocal stiffness concentration. Large secondary stresses are also induced into the girder flange tocolumn flange joint by kinking of the column flanges resulting from shear deformation of the columnpanel zone.

The dynamic loading experienced by the moment-resisting connections in earthquakes ischaracterized by high strain tension-compression cycling. Bridge engineers have long recognized thatthe dynamic loading associated with bridges necessitates different connection details in order to provideimproved fatigue resistance, as compared to traditional building design that is subject to “static”loading due to gravity and wind loads. While the nature of the dynamic loads resulting fromearthquakes is somewhat different than the high cycle dynamic loads for which fatigue-prone structuresare designed, similar detailing may be desirable for buildings subject to seismic loading.

In design and construction practice for welded steel bridges, mechanical and metallurgical notchesshould be avoided because they may be the initiators of fatigue cracking. As fatigue cracks grow underrepetitive loading, a critical crack size may be reached whereupon the material toughness (which is afunction of temperature) may be unable to resist the onset of brittle (unstable) crack growth. Thebeam-to-column connections in WSMF buildings are comparable to category C or D bridge details thathave a reduced allowable stress range as opposed to category B details for which special metallurgical,inspection and testing requirements are applied. The rapid rate of loading imposed by seismic events,and the complete inelastic range of tension-compression-tension loading applied to these connections is


Introduction

1-4

much more severe than typical bridge loading applications. The mechanical and metallurgical notchesor stress risers created by the beam-column weld joints are a logical point for fracture problems toinitiate. This, coupled with the tri-axial restraint provided by the beam web and the column flange, is arecipe for brittle fracture.

During the Northridge Earthquake, oOnce fractures initiated in beam-column joints, theyprogressed in a number of different ways. In some cases, the fractures initiated but did not grow, andcould not be detected by visual observation. In other cases, In many cases, the fractures progressedcompletely directly through the thickness of the weld, and if fireproofing was removed, the fractureswere evident as a crack through exposed faces of the weld, or the metal just behind the weld (Figure 1-2a). Other fracture patterns also developed. In some cases, the fracture developed into a surface thatresembled a through-thickness failure of the column flange material behind the CJP weld (Figure 1-2b). In these cases, a portion of the column flange remained bonded to the beam flange, but pulled freefrom the remainder of the column. This fracture pattern has sometimes been termed a “divot” or“nugget” failure.

A number of fractures progressed completely through the column flange, along a near horizontalplane that aligns approximately with the beam lower flange (Figure 1-3a). In some cases, thesefractures extended into the column web and progressed across the panel zone Figure (1-3b). Investigators have reported some instances where columns fractured entirely across the section.

a. Fracture at Fused Zone b. Column Flange “Divot” Fracture

Figure 1-2 - Fractures of Beam to Column Joints

a. Fractures through Column Flange b. Fracture Progresses into Column Web

Figure 1-3 - Column Fractures


Introduction

1-5

Once these fractures have occurred, the beam - column connection has experienced a significantloss of flexural rigidity and capacity. Residual flexural strength and rigidity must be developed througha couple consisting of forces transmitted through the remaining top flange connection and the webbolts. Initial rResearch suggests that residual stiffness is approximately 20% of that of the undamagedconnection and that residual strength varies from 10% to 40% of the undamaged capacity, whenloading results in tensile stress normal to the fracture plane. When loading produces compressionacross the fracture plane, much of the original strength and stiffness remain. However, in providingthis residual strength and stiffness, the beam shear connections can themselves be subject to failures,consisting of fracturing of the welds of the shear plate to the column, fracturing of supplemental weldsto the beam web or fracturing through the weak section of shear plate aligning with the bolt holes(Figure 1-4).

Figure 1-4 - Vertical Fracture through Beam Shear Plate Connection

It is now known that these fractures were the result of a number of complex factors that were notwell understood either when these connections were first adopted as a standard design approach, orwhen the damage was discovered immediately following the Northridge earthquake. Engineers hadcommonly assumed that when these connections were loaded to yield levels, flexural stresses in thebeam would be transferred to the column through a force couple comprised of nearly uniform yieldlevel tensile and compressive stresses in the beam flanges. It was similarly assumed that nearly all ofthe shear stress in the beam was transferred to the column through the shear tab connection to thebeam web. In fact, the actual behavior is quite different from this. As a result of local deformationsthat occur in the column at the location of the beam connection, a significant portion of the shear stressin the beam is actually transferred to the column through the beam flanges. This causes large localizedsecondary stresses in the beam flanges, both at the toe of the weld access hole and also in the completejoint penetration weld at the face of the column. The presence of the column web behind the columnflange tends to locally stiffen the joint of the beam flange to the column flange, further concentratingthe distribution of connection stresses and strains. Finally, the presence of the heavy beam and columnflange plates, arranged in a “+” shaped pattern at the beam flange to column flange joint produces acondition of very high restraint, which retards the onset of yielding, by raising the effective yieldstrength of the material, and allowing the development of very large stresses.


Introduction

1-6

The most severe stresses typically occur at the root of the complete joint penetration weld of thebeam bottom flange to the column flange. This is precisely the region of this welded joint that is mostdifficult for the welder to properly complete, as the access to the weld is restricted by the presence ofthe beam web and the welder often performs this weld while seated on the top flange, in the so-called“wildcat” position. The welder must therefore work from both sides of the beam web, starting andterminating the weld near the center of the joint, a practice that often results in poor fusion and thepresence of slag inclusions at this location. These conditions, which are very difficult to detect whenthe weld backing is left in place, as was the typical practice, are ready-made crack initiators. When thisregion of the welded joints is subjected to the large concentrated tensile stresses, the weld defects beginto grow into cracks and these cracks can quickly become unstable and propagate as brittle fractures. Once these brittle fractures initiate, they can grow in a variety of patterns, as described above, underthe influence of the stress field and the properties of the base and weld metals present at the zone of thefracture.

Despite the obvious local strength impairment resulting from these fractures, many damagedbuildings did not display overt signs of structural damage, such as permanent drifts or extreme damageto architectural elements. Until news of the discovery of connection fractures in some buildings beganto spread through the engineering community, it was relatively common for engineers to performcursory post-earthquake evaluations of WSMF buildings and declare that they were undamaged. Inorder to reliably determine if a building has sustained connection damage, it is necessary to removearchitectural finishes and fireproofing and perform nondestructive examination including visualinspection and ultrasonic testing careful visual inspection of the welded joints supplemented, in somecases, by nondestructive testing. Even if no damage is found, this is a costly process. Repair ofdamaged connections is even more costly. A few WSMF buildings have sustained so much connectiondamage that it has been deemed more practical to demolish the structures rather than to repair them. In the case of one WSMF building, damaged by the Northridge earthquake, repair costs weresufficiently large that the owner elected to demolish rather than replace than building.

Immediately following the Northridge Earthquake, a series of tests of beam-column subassemblieswere performed at the University of Texas at Austin, under funding provided by the AISC as well asprivate sources. The test specimens used heavy W14 column sections and deep (W36) beam sectionscommonly employed in some California construction. Initial specimens were fabricated using thestandard prequalified connection specified by the Uniform Building Code (UBC). Section 2211.7.1.2of UBC-94 {NEHRP-91 Section 10.10.2.3} specified this prequalified connection as follows:

“2211.7.1.2 Connection strength. The girder top column connection may be considered to be adequateto develop the flexural strength of the girder if it conforms to the following:

1. the flanges have full penetration butt welds to the columns.

2. the girder web to column connection shall be capable of resisting the girder shear determined for thecombination of gravity loads and the seismic shear forces which result from compliance with Section2211.7.2.1. This connection strength need not exceed that required to develop gravity loads plus3(Rw/8) times the girder shear resulting from the prescribed seismic forces.


Introduction

1-7

Where the flexural strength of the girder flanges is greater than 70 percent of the flexural strength ofthe entire section, (i.e. btf/(d-tf)Fy>0.7ZxFy) the web connection may be made by means of welding orhigh-strength bolting.

For girders not meeting the criteria in the paragraph above, the girder web-to-column connection shallbe made by means of welding the web directly or through shear tabs to the column. That welding shallhave a strength capable of developing at least 20 percent of the flexural strength of the girder web. Thegirder shear shall be resisted by means of additional welds or friction-type slip-critical high strength boltsor both.

and:

2211.7.2.1 Strength. The panel zone of the joint shall be capable of resisting the shear induced by beambending moments due to gravity loads plus 1.85 times the prescribed seismic forces, but the shearstrength need not exceed that required to develop 0.8ΣMs of the girders framing into the column flangesat the joint...”

In order to investigate the effects that backing bars and weld tabs had on connection performance,these were removed from the specimens prior to testing. Despite these precautions, the test specimensfailed at very low levels of plastic loading. Following these tests at the University of Texas at Austin,reviews of literature on historic tests of these connection types indicated a significant failure rate in pasttests as well, although these had often been ascribed to poor quality in the specimen fabrication. It wasconcluded that the prequalified connection, specified by the building code, was fundamentally flawedand should not be used for new construction in the future.

In retrospect, this conclusion may have been somewhat premature. More recent testing ofconnections having configurations similar to those of the prequalified connection, but incorporatingtougher weld metals, having backing bars removed from the bottom flange joint, and fabricated withgreater care to avoid the defects that can result in crack initiation, have performed better than thoseinitially tested at the University of Texas. However, as a class, when fabricated using currentlyprevailing construction practice, these connections still do not appear to be capable of consistentlydeveloping the levels of ductility presumed by the building codes for service in moment-resisting framesthat are subjected to large inelastic demands.When the first test specimens for that series werefabricated, the welder failed to follow the intended welding procedures. Further, no special precautionswere taken to assure that the materials incorporated in the work had specified toughness. Someengineers, with knowledge of fracture mechanics, have suggested that if materials with adequatetoughness are used, and welding procedures are carefully specified and followed, adequate reliabilitycan be obtained from the traditional connection details. Others believe that the conditions of high tri-axial restraint present in the beam flange to column flange joint (Blodgett - 1995) would prevent ductilebehavior of these joints regardless of the procedure used to make the welds. Further they point to theimportant influence of the relative yield and tensile strengths of beam and column materials, and othervariables, that can affect connection behavior. To date, there has not been sufficient researchconducted to resolve this issue.


Introduction

1-8

In reaction to the University of Texas tests as well as the widespread damage discovered followingthe Northridge Earthquake, and the urging of the California Seismic Safety Commission, in September,1994 the International Conference of Building Officials (ICBO) adopted an emergency code change tothe 1994 edition of the Uniform Building Code (UBC-94) {1994 NEHRP Recommended ProvisionsSection 5.2}. This code change, jointly developed by the Structural Engineers Association ofCalifornia, AISI and ICBO staff, deleted the prequalified connection and substituted the following in itsplace:

“2211.7.1.2 Connection Strength. Connection configurations utilizing welds or high-strengthbolts shall demonstrate, by approved cyclic test results or calculation, the ability to sustaininelastic rotation and develop the strength criteria in Section 2211.7.1.1 considering the effect ofsteel overstrength and strain hardening.”

“2211.7.1.1 Required strength. The girder-to-column connection shall be adequate to develop thelesser of the following:

1. The strength of the girder in flexure.

2. The moment corresponding to development of the panel zone shear strength as determined fromformula 11-1.”

Unfortunately, neither the required “inelastic rotation”, or calculation and test procedures are welldefined by these code provisions. Design Advisory No. 3 (SAC-1995) included an InterimRecommendation (SEAOC-1995) that attempted to clarify the intent of this code change, and thepreferred methods of design in the interim period until additional research could be performed andreliable acceptance criteria for designs re-established. The State of California similarly published a jointInterpretation of Regulations (DSA-OSHPD - 1994) indicating the interpretation of the current coderequirements which would be enforced by the state for construction under its control. This appliedonly to the construction of schools and hospitals in the State of California. The intent of these InterimGuidelines is to supplement these previously published documents and to provide updatedrecommendations based on the results of the limited directed research performed to date.

1.4 The SAC Joint Venture


1.5 Sponsors


1.6 Summary of Phase 1 Research


1.7 Intent


Introduction

1-9


1.8 Limitations


1.9 Use of the Guidelines



Introduction

1-10

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References

12-1

12. REFERENCES

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ATLSS, Testing of Welded “T” Specimens, (publication pending, title not exact), SAC, 1995 AStudy of the Effects of Material and Welding Factors on Moment-Frame Weld JointPerformance Using a Small-Scale Tension Specimen. Kauffman, E.J., and Fisher, J.W., SAC95-08 1995.

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Campbell, S., “Modeling of Weld Fractures Using the Drain Programs”, Technical Report:Parametric Analytical Investigations of Ground Motion and Structural Response, NorthridgeEarthquake of January 17, 1994 SAC95-05. SAC, 1995.

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Diererlein, G. “Summary of Building Analysis Studies” Analytical and Field Investigations ofBuildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 1, SAC,December, 1995

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Engelhardt, M. D., Keedong, K.M., Uzarski, J., Abunassar, H., Sabol, T.A., Ho, L., and Kim, H.“Parametric Studies on Inelastic Modeling of Steel Moment Frames”, Technical Report:Parametric Analytical Investigations of Ground Motion and Structural Response, NorthridgeEarthquake of January 17, 1994 SAC95-05. SAC, 1995.

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Paret, T.F., Sasaki, K.K., “Analysis of a 17 Story Steel Moment Frame Building Damaged by theNorthridge Earthquake”, Analytical and Field Investigations of Buildings Affected by theNorthridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC, December, 1995.

Popov, E.P. and Yang, T.S. Steel Seismic Moment Resisting Connections. University ofCalifornia at Berkeley, May, 1995.

Popov, E.P. Blondet, M., Stepanov, L, and Stodjadinovic, B. Full-Scale Beam-ColumnConnection Tests. et. al. Testing of Repaired Welded Beam Column Assemblies, SAC,publication pending (title not exact), 1995 SAC 96-01. 1996..

SAC, Proceedings of the International Workshop on Steel Moment Frames, October 23-24, 1994SAC-94-01. Sacramento, CA, December, 1994.


References

12-7

SAC . Steel Moment Frame Advisory No. 1. September, Sacramento, CA, 1994.

SAC . Steel Moment Frame Advisory No. 2. October, Sacramento, CA, 1994.

SAC . Steel Moment Frame Advisory No. 3 SAC-95-01, February, Sacramento, CA, 1995.

Shonafelt, G.O., and Horn, W.B.. Guidelines for Evaluation and Repair of Damaged SteelBridge Members, NCHRP Report 271, Transportation Research Board, 1984.

Skiles, J.L. and Campbell, H.H., “Why Steel Fractured in the Northridge Earthquake” SACAdvisory No. 1, October, 1994.

Seismic Safety Commission, Northridge Earthquake Turning Loss to Gain, Report to theGovernor, Sacramento, CA, 1995.

Smith Emery Company. Report of Test, July, 1995.

Sommerville, P, Graves, R., Chandan, S. Technical Report: Characterization of Ground MotionDuring the Northridge Earthquake of January 17, 1994, SAC 95-03, SAC, December, 1995.

State of California. Division of the State Architect (DSA) and Office of Statewide HealthPlanning and Development (OSHPD). Interpretation of Regulations Steel Moment ResistingFrames, Sacramento, CA, 1994.

Structural Engineers Association of California (SEAOC), Seismology Committee, RecommendedLateral Force Requirements and Commentary, Sacramento, CA. 1990.

Structural Engineers Association of California (SEAOC), Seismology Committee, InterimRecommendations for Design of Steel Moment Resisting Connection,. Sacramento, CA, January,1995.

Structural Engineers Association of California (SEAOC), Vision 2000: A Framework forPerformance Based Engineering of Buildings, Sacramento, CA, April, 1995.

Structural Shape Producers Council, Statistical Analysis of Tensile Data for Wide FlangeStructural Shapes, 1994.

Thiel, C.C., and Zsutty, T.C., “Earthquake Characteristics and Damage Statistics,” EarthquakeSpectra, Volume 3, No. 4., Earthquake Engineering Research Institute, Oakland, Ca. 1987.

Tremblay, R., Tchebotarev, N., and Filiatrault, A., “Seismic Performance of RBS Connections forSteel Moment Resisting Frames: Influence of Loading Rate and Floor Slab,” Proceedings of theSecond International Conference on the Behavior of Steel Structures in Seismic Area, Kyoto,Japan, August, 1997


References

12-8

Tsai, K.C. and Popov, E. P. “Seismic Steel Beam-Column Moment Connections” State of ArtPapers: Metallurgy, Fracture Mechanics, Welding, Moment Connections and Frame SystemBehavior SAC 95-09. SAC, September, 1996

Uang, C.M. and Latham, C.T. Cyclic Testing of Full-Scale MNH-SMRF Moment Connections,Structural Systems Research, University of California, San Diego, March, 1995.

Tsai, K.C. and Popov, E.P., Steel Beam - Column Joints In Seismic Moment Resisting Frames,Report No. UCB/EERC-88/19, Earthquake Engineering Research Center, University ofCalifornia, Berkeley, Nov., 1988.

Uang, C.M., Yu, Q.S., Sadre, A., Bonowitz, D., Youssef, N. “Performance of a 13 Story SteelMoment-Resisting Frame Damaged in the 1994 Northridge Earthquake”, ” Analytical and FieldInvestigations of Buildings Affected by the Northridge Earthquake of January 17, 1994, SAC 95-04 Part 2 SAC, December, 1995.

Uang, C.M. and Bondad, D. Progress Report on Cyclic Testing of Three Repaired UCSDSpecimens, SAC, 1995.

Uang, C.M. and Lee, C.H. “Seismic Response of Haunch Repaired Steel SMRFs: AnalyticalModelling and Case Studies” ” Analytical and Field Investigations of Buildings Affected by theNorthridge Earthquake of January 17, 1994, SAC 95-04 Part 2, SAC., December, 1995

Wald, D.J., Heaton, T.H., and Hudnut, K.W., The Slip History of the 1994 Northridge,California, Earthquake Determined from Strong-Motion, Teleseismic, GPS, and Leveling Data,United Sates Geologic Survey, 1995.

Watabe, M. Peformance of Wooden Houses and Steel Buildings during the Great HanshinEarthquake, Architectural Institute of Japan, May, 1995.

Youssef, N.F.G, Bonowitz, D., and Gross, J.L., A Survey of Steel Moment-Resisting FrameBuildings Affected by the 1994 Northridge Earthquake, NISTR 5625, Gaithersburg Md, April,1995.


Classifications and Implications of Damage

3-1

3. CLASSIFICATION AND IMPLICATIONS OF DAMAGE

3.1 Summary of Earthquake Damage


3.2 Damage Types


3.2.1 Girder Damage

There are no modifications to the Guidelines or Commentary of Section 3.2.1 at this time.

3.2.2 Column Flange Damage


3.2.3 Weld Damage, Defects and Discontinuities

Six types of weld discontinuities, defects and damage are defined in Table 3-3 and illustratedin Figure 3-4. All apply to the complete joint penetration (CJP) welds between the girder flangesand the column flanges. This category of damage was the most commonly reported typefFollowing the Northridge Earthquake, many instances of W1a and W1b conditions were reportedas damage. These conditions, which are detectable only by ultrasonic testing or by removal ofweld backing, are now thought more likely to be construction defects than damage.

Table 3-3 - Types of Weld Damage, Defects and Discontinuities

Type DescriptionW1 Weld root indications

W1a Incipient indications -– depth <3/16” ortf/4; width < bf/4

W1b Root indications larger than that for W1aW2 Crack through weld metal thicknessW3 Fracture at column interfaceW4 Fracture at girder flange interfaceW5 UT detectable indication - non-rejectable



3-2

W1, W5W2

W3W4

Note: See Figure 3-2 for related column damage and Figure 3-3 for girder damageFigure 3-4 - Types of Weld Damage

Commentary: Despite significant controversy, type W1 and W5 discovered inbuildings following the Northridge earthquake, were commonly reported asdamage. These small discontinuities and defects located at the roots of the CJPwelds are detectable only by ultrasonic testing (UT) when the weld backing is leftin place or by visual testing (VT) or magnetic particle testing (MT) when weldbacking is removed. It now seems likely that most such conditions are notdamage at all, but rather, are pre-existing construction defects. A number offactors point to this conclusion. First, statistical surveys of damage sustained bybuildings in the Northridge earthquake show that if type W1 and W5 conditionsare not considered, there was a much greater incidence of damage in framesresisting north-south ground shaking than in frames resisting east-west shaking.This appears to be correlated with the relative strength of the ground shakingexperienced along these two directional axes. However, there is no significantdifference between the incidence rate of reported W1 and W5 conditions in thesetwo directions, suggesting that these conditions are not correlated with shakingintensity.

The discovery of W1 conditions in welds for which original constructionquality assurance documentation is available, indicating that no such defectswere present when the building was originally constructed, tends to contradictthis argument. However, investigations conducted by SAC under the Phase 2project have indicated that as a result of the joint geometry, UT techniques areoften unable to detect W1 conditions at the weld root, when scanning of the jointis conducted from the top surface of the beam bottom flange. It is important tonote that this is the most common method of conducting UT as part ofconstruction quality assurance. When UT scanning of a joint is conducted fromthe bottom surface of the flange, as is commonly done when inspecting for



3-3

earthquake damage, it becomes more likely that such conditions will be detected,since the geometric constraints present for top flange scanning are altered. Thisleads to the conclusion that it is probable that typical construction qualityassurance UT of welded joints would be likely to miss W1 conditions, allowingthem to be discovered in later post-earthquake surveys.

When FEMA-267 was first published, it was recommended that W1 conditionsbe treated as damage and that UT be used as a routine part of the post-earthquake investigation process, in order to discover these conditions. However,more recent investigations conducted by SAC have revealed that even the carefulscanning typically conducted as part of a post-earthquake inspection is not ableto reliably detect these conditions. Given that it is both expensive and difficult tolocate W1 conditions as part of a post-earthquake investigation, and also, thatmost of these conditions are unlikely to be damage at all, it is no longerrecommended that exhaustive investigations for these conditions be conducted aspart of the earthquake damage investigation process.

Type W1 damage, discontinuities and defects and type W5 discontinuities aredetectable only by NDT, unless the backing bar is removed, allowing directdetection by visual inspection or magnetic particle testing. Type W5 consists ofsmall discontinuities and may or may not actually be earthquake damage. AWSD1.1 permits small discontinuities in welds. Larger discontinuities are termeddefects, and are rejectable per criteria given in the Welding Code. It is likelytherefore that some weld indications detected by NDT in a post-earthquakeinspection may be discontinuities which pre-existed the earthquake and do notconstitute a rejectable condition, per the AWS standards. Repair of thesediscontinuities, designated as type W5 is not generally recommended. Some typeW1 indications are small planar defects, which are rejectable per the AWS D1.1criteria, but are not large enough to be classified as one of the types W2 throughW4. Type W1 is the single most commonly reported non-conforming conditionreported in the post-Northridge statistical data survey, and in some structures,represents more than 80 per cent of the total damage reported. The W1classification is split into two types, W1a and W1b, based on their severity. TypeW1a “incipient” root indications are defined as being nominal in extent, less than3/16” deep or 1/4 of the flange thickness, whichever is less, and having a lengthless than 1/4 of the flange width. Some engineers believe that type W1aindications are not earthquake damage at all, but rather, previously undetecteddefects from the original construction process. A W1b indication is one thatexceeds these limits but is not clearly characterized by one of the other types. Itis more likely that W1b indications are a result of the earthquake than theconstruction process.

As previously stated, some engineers believe that both type W1a and sometype W1b conditions are not earthquake related damage at all, but instead, are



3-4

rejectable conditions not detected by the quality control and assurance programsin effect during the original construction. However, in recent large-scale sub-assembly testing of the inelastic rotation capacity of girder-column connectionsconducted in SAC Phase 1 at the University of Texas at Austin and theEarthquake Engineering Research Center of the University of California atBerkeley, it was reported that significantly more indications were detectable inunfailed CJP welds following the testing than were detectable prior to the test.This tends to indicate that type W1 damage may be related to stresses induced inthe structures by their response to the earthquake ground motions. Regardless ofwhether or not type W1 conditions are directly attributable to earthquakeresponse, it is clear that these conditions result in a reduced capacity for the CJPwelds and can act as stress risers, or notches, to initiate fracture in the event offuture strong demands.

Type W2 fractures extend completely through the thickness of the weld metaland can be detected by either MT or VI techniques. Type W3 and W4 fracturesoccur at the zone of fusion between the weld filler metal and base material of thegirder and column flanges, respectively. All three types of damage result in aloss of tensile capacity of the girder flange to column flange joint and should berepaired.

As with girder damage, damage to welds has most commonly been reported atthe bottom girder to column connection, with fewer instances of reported damageat the top flange. Available data indicates that approximately 25 per cent of thetotal damage in this category occurs at the top flange, and most often, top flangedamage occurs in connections which also have bottom flange damage. For thesame reasons previously described for girder damage, less weld damage may beexpected at the top flange. However, it is likely that there is a significant amountof damage to welds at the top girder flange which have never been discovered dueto the difficulty of accessing this joint. Later sections of these Interim Guidelinesprovide recommendations for situations when such inspection should beperformed.

3.2.4 Shear Tab Damage


3.2.5 Panel Zone Damage


3.2.6 Other Damage




3-5

3.3 Safety Implications

The implications of the damage described above with regard to building safety are discussed inthis section. As part of the SAC Phase 2 program, extensive nonlinear analyses have beenconducted of WSMF buildings to determine the effects of connection fractures on buildingperformance and also to develop an understanding of the risk of earthquake-induced buildingcollapse. These studies indicate that risk of collapse of WSMF buildings designed to modernstandards and having connections capable of ductile behavior is quite low. Even in regions ofvery high seismicity, such as those areas of coastal California adjacent to major active faults, theprobability that such a building would experience earthquake-induced collapse appears to be onthe order of one occurrence per building, every 20,000 years. For buildings that have brittleconnections such as those commonly constructed prior to 1994, the probability of collapseincreases somewhat. If only the bottom flange connections of beams to columns is subject tofracture, the risk of global collapse of buildings increases to perhaps one occurrence in 15,000years, presuming that the fractures do not jeopardize column capacity. However, if both flangesof the connections are subject to fracture, or if substantial column damage occurs, the risk ofcollapse increases significantly. Also, it is important to note that severe connection fractures canresult in significant risk of local collapse and life safety endangerment.

While these studies have been helpful in providing an understanding of the level of riskinherent in WSMF structures with brittle connections, they do not provide sufficient informationto There is insufficient knowledge at this time to permit determination of the assess the degree ofrisk with any real confidence. However, based on the historic performance of modern WSMFbuildings, typical of those constructed in the United States, it appears that the risk of collapse inmoderate magnitude earthquakes, ranging up to perhaps M7, is very low for buildings which havebeen properly designed and constructed according to prevailing standards. A possible exceptionto this may be buildings located in the near field (< 10 km from the surface projection of the faultrupture) of such earthquakes (Heaton, et. al. - 1995), however, this is not uniquely a problemassociated with steel buildings. Our current building codes in general, may not be adequate toprovide for reliable performance of buildings within the near field of large earthquakes. As is alsothe case with all other types of construction, buildings with incomplete lateral force resistingsystems, severe configuration irregularities, inadequate strength or stiffness, poor constructionquality, or deteriorated condition are at higher risk than buildings not possessing thesecharacteristics.

No modern WSMF buildings have been sited within the areas of very strong ground motionfrom earthquakes larger than M7, or for that matter, within the very near field for eventsexceeding M6.5. This style of construction has been in wide use only in the past few decades.Consequently, it is not possible to state what level of risk may exist with regard to buildingresponse to such events. This same lack of performance data for large magnitude, long durationevents exists for virtually all forms of contemporary construction. Consequently, there isconsiderable uncertainty in assigning levels of risk to any building designed to minimum coderequirements for these larger events.



3-6

Commentary: Research conducted to date has not been conclusive with regard tothe risk of collapse of WSMF buildings. Some testing of damaged connectionsfrom a building in Santa Clarita, California have been conducted at theUniversity of Southern California (Anderson - 1995). In these tests, connectionassemblies which had experienced type P6 damage were subjected to repeatedcycles of flexural loading, while the column was maintained under axialcompression. Under these conditions, the specimens were capable of resisting asmuch as 40 per cent of the nominal plastic strength of the girder for severalcycles of slowly applied loading, at plastic deformation levels as large as 0.025radians. However, damage did progress in the specimen, as this testing wasperformed. It is not known how these assemblies would have performed if thecolumns were permitted to experience tensile loading. Data from other testssuggests that the residual strength of connections which have experienced typesG1, G4, W2, W3, and W4 damage is on the order of 15 per cent of theundamaged strength. Some analytical research (Hall - 1995) in which nonlineartime history analyses simulating the effects of connection degradation due tofractures were included, indicates that typical ground motions resulting in thenear field of large earthquakes can cause sufficient drift in these structures toinduce instability and collapse. Other researchers (Astaneh - 1995) suggest thatdamaged structures, even if unrepaired, have the ability to survive additionalground motion similar to that of the Northridge Earthquake.

Even though there were no collapses of WSMF buildings in the 1994Northridge Earthquake, it should not be assumed that no risk of such collapseexists. Indeed, a number of WSMF buildings did experience collapse in the 1995Kobe Earthquake. The detailing of these collapsed Japanese buildings wassomewhat different than that found in typical US practice, however, much of thefracture damage that occurred was similar to that discovered following theNorthridge event.

Because of a lack of data and experience with the effects of larger, longerduration earthquakes, there is considerable uncertainty about the performance ofall types of buildings in large magnitude seismic events. It is believed thatseismic risks in such large events are highly dependent on the individual groundmotion at a specific site and the characteristics of the individual buildings.Therefore, generalizations with regard to the probable performance of individualtypes of construction may not be particularly meaningful.

The risks to occupants of WSMF buildings with brittle connections is regardedas less, in most cases, than to occupants of the types of buildings listed below.However, because of the uncertainties involved, the degree of risk in large eventscannot be definitively quantified, nor can it categorically be stated that properlyconstructed WSMF buildings sited in the near field of large events are either



3-7

more or less at risk than many other code designed building systems which do notappear on the following list:

• Concentric braced steel frames with bracing connections that are weaker than thebraces

• Knee braced steel frames

• Unreinforced masonry bearing wall buildings

• Non-ductile reinforced concrete moment frames (infilled or otherwise)

• Reinforced concrete moment frames with gravity load bearing elements that werenot designed to participate in the lateral force resisting system and that do nothave capacity to withstand earthquake-induced deformations

• Tilt-up and reinforced masonry buildings with inadequate anchorage of theirheavy walls to their horizontal wood diaphragms

• Precast concrete structures without adequate interconnection of their structuralelements.

In addition, WSMF structures with brittle connections would appear to havelower inherent seismic risk than structures of any construction type that:

• do not having complete, definable load paths

• have significant weak and/or soft stories

• have major torsional irregularity and insufficient stiffness and strength to resistthe resulting seismic demands

• minimal redundancy and concentrations of lateral stiffness

These are general statements that represent a global view of systemperformance. As with all seismic performance generalizations, there are manysteel moment frame buildings that are more vulnerable to damage than someindividual buildings of the general categories listed, just as there are many thatwill perform better.

3.4 Economic Implications




3-8

This Page Intentionally Left Blank

Interim Guidelines Advisory No. 2 SAC99-01

Post Earthquake Evaluation

4-1

4. POST-EARTHQUAKE EVALUATION

4.1 Scope


4.2 Preliminary Evaluation


4.2.1 Evaluation Process

Preliminary evaluation is the process of determining if a building should be subjected todetailed post-earthquake evaluations. Detailed evaluations should be performed for all buildingsthought to have experienced strong ground motion, as indicated in Section 4.2.1.1 or for whichthe other indicators of Section 4.2.1.2 apply. Detailed post-earthquake evaluations include theentire process of determining if a building has experienced significant damage and if damage isfound, determining appropriate strategies for occupancy, structural repair and/or modification.Except as indicated in Section 4.2.3, detailed evaluation should, as a minimum, includeinspections of a representative sample of moment-resisting (and other type) connections withinthe building.

4.2.1.1 Ground Motion

There are no modifications to the Guidelines or Commentary of Section 4.2.1.1 at this time.

4.2.1.2 Additional Indicators


4.2.2 Evaluation Schedule

There are no modifications to the Guidelines of Section 4.2.2 at this time.

Commentary: It is important to conduct post-earthquake evaluations as soonfollowing the earthquake as is practical. Aftershock activity in the monthsimmediately following an earthquake is likely to produce additional strongground motion at the site of a damaged building. If there is adequate reason toassume that damage has occurred, then such damage should be expeditiouslyuncovered and repaired. However, since adequate resources for post-earthquakeevaluation may be limited, a staggered schedule is presented, with those buildingshaving a greater likelihood of damage recommended for evaluation first.



4-2

Large magnitude earthquakes are often followed by large magnitudeaftershocks. Therefore, it is particularly urgent that post-earthquake evaluationsbe performed expeditiously following such events. If insufficient resources areavailable in the affected region to perform the NDT tests recommended by theGuidelines of Chapter 5, it is recommended that visual inspection, in accordancewith Section 5.2.2, proceed as soon as possible. If visual inspection revealssubstantial damage, consideration should be given to vacating the building untileither an adequate period of time has passed so as to make the likelihood of verylarge aftershocks relatively low (e.g. 4 weeks for magnitude 7 and lower, and 8weeks for magnitudes above this), complete inspections and repairs are made, ora detailed evaluation indicates that the structure retains adequate structuralstiffness and strength to resist additional strong ground shaking. Preliminaryvisual inspections should not be used as an alternative to complete evaluation.

The table Table 4-1relates the urgency for post-earthquake buildingevaluation to both the magnitude of the earthquake and the estimated peakground acceleration experienced by the building site. This is because largemagnitude events are more likely to have large magnitude aftershocks andbecause buildings that experienced stronger ground accelerations are more likelyto have been damaged. Except in regions with extensive strong motioninstrumentation, estimates of ground motion are quite subjective. Followingmajor damaging earthquakes, government agencies usually produce groundmotion maps showing projected acceleration contours. These maps should beused when available. When such maps are not available, ground motions can beestimated using any of several attenuation relationships that have been published.

4.2.3 Connection Inspections

Except as indicated in Sections 4.2.3.1 and 4.2.3.2, below, Ddetailed evaluations shouldinclude inspection of the building’s moment-resisting connections in order to determine theircondition. As a first pass, inspections may be limited to careful visual inspection of the joint ofthe beam bottom flange to the column. When such inspection reveals the presence of connectiondamage, a more thorough inspection of the damaged connection should be conducted. Sincemoment-resisting frame buildings commonly have many connections, inspections can be quitecostly. Therefore, it shall be permissible to limit inspections toof a representative sample ofWSMF (and other) connections, except as indicated in Sections 4.2.3.1 and 4.2.3.2, below.Section 4.3.3 provides three alternative approaches to selecting an appropriate sample ofconnections for inspection.

4.2.3.1 Analytical Evaluation




4-3

4.2.3.2 Buildings with Enhanced Connections


4.2.4 Previous Evaluations and Inspections


4.3 Detailed Evaluation Procedure

Where detailed evaluation is recommended by Section 4.2, assessment of the post earthquakecondition of a building, its ability to resist additional strong ground motion and other loads, anddetermination of appropriate occupancy, structural repair and/or modification strategies should bebased on the results of a detailed inspection and assessment of the extent to which structuralsystems have been damaged.

In order to obtain complete data on a building’s post-earthquake condition, it is necessary toinspect each of the building’s moment-resisting frame elements and their connections. However,such extensive inspections could be very costly. As an alternative to that approach, this Sectionpresents a series of procedures by which a representative sample of beam-column connections isselected and inspected. This Section presents one approach for making such assessments. In thisapproach, the results of the sample inspections are used to calculate a cumulative damage index,D, for the structure as well as the probability that if all of the building’s connections had beeninspected, the damage index at any floor of the structure has would have been found to exceededa value of 1/3. General occupancy, structural repair and modification recommendations are madebased upon the values calculated for these damage indices. In particular, a calculated damageindex of 1/3 is used to indicate, in the absence of more detailed analyses, that a potentiallyhazardous condition may exist.

The structural engineer may use other procedures consistent with the principles of statisticsand structural mechanics to determine the residual strength and stiffness of the structure in the as-damaged state and the acceptability of such characteristics relative to the criteria contained in thebuilding code, or other rational criteria acceptable to the building official.

There are no modifications to the Commentary of Section 4.3 at this time.

4.3.1 Eight Step Evaluation Procedure

Post-earthquake evaluation should be carried out under the direct supervision of a structuralengineer. The following eight-step procedure may be used to determine the condition of thestructure and to develop occupancy, repair and modification strategies. Note that this procedureis written presuming that inspection is limited to a representative sample of the total number ofconnections present in the building. If all connections in the building are to be inspected, steps 1,2, 4 and 6 may be omitted.



4-4

Step 1: The moment-resisting connections in the building are categorized into two or more“groups” (Section 4.3.2 and 4.4) comprised of connections expected to have similarprobabilities of being damaged.

Complete steps 2 through 7 below, for each group of connections.

Step 2: Determine the minimum number of connections in each group that should be inspectedand select the specific sample of connections to be inspected. (Section 4.3.3)

Step 3: Inspect the selected set of connections using the technical guidelines of Section 5.2.and determine connection damage indices, dj, for each inspected connection (Section4.3.4)

Step 4: If inspected connections are found to be seriously damaged, perform additionalinspections of connections adjacent to the damaged connections. (Section 4.3.5)

Step 5: Determine the average damage index (davg) for connections in each group, and then theaverage damage index at a typical floor. (Section 4.3.6)

Step 6: Given the average damage index for connections in the group, determine theprobability, P, that the connection damage index for any group, at a floor level,exceeds 1/3, and determine the maximum estimated damage index for any floor, Dmax.(Section 4.3.7)

Step 7: Based on the calculated damage indices and statistics, determine appropriateoccupancy, structural repair and modification strategies (Section 4.3.8). If deemedappropriate, the structural engineer may conduct detailed structural analyses of thebuilding in the as-damaged state, to obtain improved understanding of its residualcondition and to confirm that the recommended strategies are appropriate or tosuggest alternative strategies.

Step 8: Report the results of the inspection and evaluation process to the building official andbuilding owner. (Section 4.3.9)

Sections 4.3.2 through 4.3.9 indicate how these steps should be performed.

There are no modifications to the Commentary of Section 4.3.1 at this time.

4.3.2 Step 1— Categorize Connections by Groups


4.3.3 Step 2— Select Samples of Connections for Inspection




4-5

4.3.3.1 Method A - Random Selection


4.3.3.2 Method B - Deterministic Selection


4.3.3.3 Method C - Analytical Selection


4.3.4 Step 3— Inspect the Selected Samples of Connections


Commentary: The sample size suggested for inspection in the methods of Section4.3.3 are based on full inspection using both visual (Section 5.3.1) and NDTtechniques (Section 5.3.2) at all connections in the sample. Other methods ofselection and inspection may be used as provided in Section 4.3, with theapproval of the building official. One such approach might be the visual-onlyinspection of the bottom girder flange to column connection, but with theinspection of a large fraction of the total connections in the group, possiblyincluding all of them. If properly performed, such an inspection procedure woulddetect almost all instances of the most severe damage but would not detect welddefects (W1a), or root cracking (W1b), nor lamellar damage in columns (C5).The occurrence of a few of these conditions, randomly scattered through thebuilding would not greatly affect the assessment of the building’s post-earthquakecondition, or the calculation of the damage index. However, if a large number ofsuch defects were present in the building, this would be significant to the overallassessment. Therefore, such an inspection approach should probably includeconfirming NDT investigations of at least a representative sample of the totalconnections investigated. If within that sample, significant incidence of visuallyhidden damage is found, then full NDT investigations should be performed, assuggested by these Interim Guidelines. Similarly, if visual damage is found at thebottom flange, then complete connection inspection should be performed todetermine if other types of damage are also present.

4.3.4.1 Damage Characterization

Characterize the observed damage at each of the inspected connections by assigning aconnection damage index, dj, obtained either from Table 4-3a or Table 4-3b. Table 4-3a presentsdamage indices for individual classes of damage and a rule for combining indices where aconnection has more than one type of damage. Table 4-3b provides combined indices for themore common combinations of damage.



4-6

Table 4-3a - Connection Damage Indices

Type Location Description1 Index2djG1 Girder Buckled Flange 4G2 Girder Yielded Flange 1G3 Girder Top or Bottom Flange fracture in HAZ 8G4 Girder Top or Bottom Flange fracture outside HAZ 8G5 Girder Top and Bottom Flange fracture 10G6 Girder Yielding or Buckling of Web 4G7 Girder Fracture of Web 10G8 Girder Lateral-torsional Buckling 8C1 Column Incipient flange crack (detectable by UT) 4C2 Column Flange tear-out or divot 8C3 Column Full or partial flange crack outside HAZ 8C4 Column Full or partial flange crack in HAZ 8C5 Column Lamellar flange tearing 6C6 Column Buckled Flange 8C7 Column Fractured column splice 8W1a CJP weld Minor root indication - thickness <3/16” or tf/4; width < bf/4 01W1b CJP weld Root indication - thickness > 3/16” or tf/4 or width > bf/4 04W2 CJP weld Crack through weld metal thickness 8W3 CJP weld Fracture at girder interface 8W4 CJP weld Fracture at column interface 8W5 CJP weld Root indication— non-rejectable 0S1a Shear tab Partial crack at weld to column (beam flanges sound) 4S1b Shear tab Partial crack at weld to column (beam flange cracked) 8S2a Shear tab Crack in Supplemental Weld (beam flanges sound) 1S2b Shear tab Crack in Supplemental Weld (beam flange cracked) 8S3 Shear tab Fracture through tab at bolt holes 10S4 Shear tab Yielding or buckling of tab 6S5 Shear tab Damaged, or missing bolts4 6S6 Shear tab Full length fracture of weld to column 10P1 Panel Zone Fracture, buckle, or yield of continuity plate3 4P2 Panel Zone Fracture of continuity plate welds3 4P3 Panel Zone Yielding or ductile deformation of web3 1P4 Panel Zone Fracture of doubler plate welds3 4P5 Panel Zone Partial depth fracture in doubler plate3 4P6 Panel Zone Partial depth fracture in web3 8P7 Panel Zone Full (or near full) depth fracture in web or doubler plate3 8P8 Panel Zone Web buckling3 6P9 Panel Zone Fully severed column 10Notes To Table 4-3a:

1. See Figures 3-2 through 3-6 for illustrations of these types of damage.2. Where multiple damage types have occurred in a single connection, then:

a. Sum the damage indices for all types of damage with d=1 and treat as one type. If multiple types stillexist; then:



4-7

b. For two types of damage refer to Table 4-3b. If the combination is not present in Table 4-3b and thedamage indices for both types are greater than or equal to 4, use 10 as the damage index for theconnection. If one is less than 4, use the greater value as the damage index for the connection.

c. If three or more types of damage apply and at least one is greater than 4, use an index value of 10,otherwise use the greatest of the applicable individual indices.

3. Panel zone damage should be reflected in the damage index for all moment connections attached to thedamaged panel zone within the assembly.

4. Missing or loose bolts may be a result of construction error rather than damage. The condition of the metalaround the bolt holes, and the presence of fireproofing or other material in the holes can provide clues to this.Where it is determined that construction error is the cause, the condition should be corrected and a damageindex of “0” assigned.

Table 4-3b - Connection Damage Indices for Common Damage Combinations1

Girder, Columnor Weld Damage

Shear TabDamage

DamageIndex

Girder, Columnor Weld Damage

Shear TabDamage

DamageIndex

G3 or G4 S1a 8 C5 S1a 6S1b 10 S1b 10S2a 8 S2a 6S2b 10 S2b 10S3 10 S3 10S4 10 S4 10S5 10 S5 10S6 10 S6 10

C2 S1a 8 W2, W3, or W4 S1a 8S1b 10 S1b 10S2a 8 S2a 8S2b 10 S2b 10S3 10 S3 10S4 10 S4 10S5 10 S5 10S6 10 S6 10

C3 or C4 S1a 8S1b 10S2a 8S2b 10S3 10S4 10S5 10S6 10

1. See Table 4-3a, footnote 2 for combinations other than those contained in this table.

More complete descriptions (including sketches) of the various types of damage are providedin Section 3.1. When the engineer can show by rational analysis that other values for the relativeseverities of damage are appropriate, these may be substituted for the damage indices provided in



4-8

the tables. A full reporting of the basis for these different values should be provided to thebuilding official, upon request.

Commentary: The connection damage indices provided in Table 4-3 (rangingfrom 0 to 10) represent judgmental estimates of the relative severities of thisdamage. An index of 0 indicates no damage and an index of 10 indicates verysevere damage.

When initially developed, these connection damage indices wereconceptualized as estimates of the connection’s lost capacity to reliablyparticipate in the building’s lateral-force-resisting system in future earthquakes(with 0 indicating no loss of capacity and 10 indicating complete loss ofcapacity). However, due to the limited data available, no direct correlationbetween these damage indices and the actual residual strength and stiffness of adamaged connection was ever made. They do provide a convenient measure,however, of the extent of damage that various connections in a building haveexperienced.

When FEMA-267 was first published, weld root discontinuities, Type W1a anddefects, type W1b, were classified as damage in Table 4-3a with damage indicesof 1 and 4, respectively assigned. Recent evidence and investigations, however,suggest strongly that these W1 conditions are not likely to be damage, and alsoare difficult to reliably detect. As a result, with the publication of InterimGuidelines Advisory No. 2, the damage indices for these conditions has beenreduced to a null value, consistent with classifying them as pre-existingconditions, rather than damage.

It should be noted that the reduced damage index associated with theseconditions is not intended to indicate that these are not a concern with regard tofuture performance of the building. In particular, type W1b conditions can serveas ready initiators for the types of brittle fractures associated with the otherdamage types and connections having such conditions are more susceptible tofuture earthquake-induced damage than connections that do not have theseconditions. Correction of these conditions should generally be considered anupgrade or modification, rather than a damage repair.

4.3.5 Step 4— Inspect Connections Adjacent to Damaged Connections


4.3.6 Step 5— Determine Average Damage Index for Each Group




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4.3.7 Step 6— Determine the Probability that the Connections in a Group at a Floor Level Sustained Excessive Damage


4.3.7.1 Some Connections in Group Not Inspected


4.3.7.2 All Connections in Group Inspected


4.3.8 Step 7— Determine Recommended Recovery Strategies for the Building


4.3.9 Step 8 - Evaluation Report


4.4 Alternative Group Selection for Torsional Response


4.5 Qualified Independent Engineering Review


4.5.1 Timing of Independent Review


4.5.2 Qualifications and Terms of Employment


4.5.3 Scope of Review


4.5.4 Reports




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4.5.5 Responses and Corrective Actions


4.5.6 Distribution of Reports


4.5.7 Engineer of Record


4.5.8 Resolution of Differences



Post-Earthquake Inspection

5-1

5. POST-EARTHQUAKE INSPECTION

When required by the building official, or recommended by the Interim Guidelines in Chapter4, post-earthquake inspections of buildings may be conducted in accordance with the InterimGuidelines of this Chapter. In order to determine, with certainty, the actual post-earthquakecondition of a building, it is necessary to inspect all elements and their connections. However, itis permissible to select An an appropriate sample (or samples) of WSMF connections should beselected for inspection in accordance with the Chapter 4 Guidelines. These connections, andothers deemed appropriate by the engineer, should be subjected to visual inspection (VI) andsupplemented by non-destructive testing (NDT) as required by this Chapter.

Commentary: The only way to be certain that all damage sustained by a buildingis detected is to perform complete inspections of every structural element andconnection. In most cases, such exhaustive post-earthquake inspections would beboth economically impractical and also unnecessary. As recommended by theseguidelines, the purpose of post-earthquake inspections is not to detect all damagethat has been sustained by a building, but rather, to detect with reasonablecertainty, that damage likely to result in a significant degradation in thebuilding’s ability to resist future loading. The connection sampling process,suggested by Chapter 4 of these Interim Guidelines was developed to provide alow probability that damage in buildings that had sustained a substantialreduction in load carrying capacity would be overlooked while avoiding theperformance of exhaustive investigations of buildings that have sustainedrelatively insignificant damage.

Where greater certainty in the detection of damage is desired for a building, amore extensive program of inspection can be conducted. For those cases inwhich it is desired to perform an analytical determination of the residual loadcarrying capacity of the structure, complete inspections of elements andconnections should be performed so that an analytical model of the building canbe developed that reasonably represents its post-earthquake condition.

5.1 Connection Types Requiring Inspection

5.1.1 Welded Steel Moment Frame (WSMF) Connections

The inspection of a WSMF connection should start with visual inspection of the weldedbottom beam flange to column flange joint and the base materials immediately adjacent to thisjoint. If damage to this joint is apparent, or suspected, then inspections of that connection shouldbe extended to include the complete joint penetration (CJP) groove welds connecting both topand bottom beam flanges to the column flange, including the backing bar and the weld accessholes in the beam web; the shear tab connection, including the bolts, supplemental welds and



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beam web; the column's web panel zone, including doubler plates; and the continuity plates andcontinuity plate welds (See Figure 3-1). In addition, where visual inspection indicates potentialconcealed damage, visual inspection should be supplemented with other methods ofnondestructive testing.

Commentary: The largest concentration of reported damage following theNorthridge Earthquake occurred at the welded joint between the bottom girderflange and column, or in the immediate vicinity of this joint. To a much lesserextent, damage was also observed in some buildings at the joint between the topgirder flange and column. If damage at either of these locations is substantial (dj

per Chapter 4 greater than 5), then damage is also commonly found in the panelzone or shear tab areas.

When originally published,These these Interim Guidelines recommendedcomplete inspection, by visual and NDT assisted means, of all of these potentialdamage areas for a small representative sample of connections. This practice iswas consistent with that followed by most engineers in the Los Angeles area,following the Northridge Earthquake. It requires removal of fireproofing from arelatively large surface of the steel framing, which at most connections will beundamaged.

In the time since the Interim Guidelines were first published, extensiveinvestigations have been conducted of the statistical distribution of damagesustained by buildings in the Northridge earthquake, the nature of this damageand the effect of this damage on the future load-carrying capacity of thebuildings. These investigations strongly suggest that the W1a and W1bconditions at the weld root are unlikely to be earthquake damage, but rather,conditions of discontinuity and defects from the original construction. Further,studies have shown that NDT methods are generally unreliable in the detection ofthese conditions. As a result, the current recommendation is not to conductexhaustive NDT investigations of connections in order to discover hiddendamage, as was originally recommended.

In a series of analytical investigations of the effect of moment-resistingconnection damage on building behavior, it was determined that even if a largenumber of connections experience fracture at one beam flange to column joint,there is relatively little increase in the probability of global collapse in a futureearthquake. Similarly, these investigations indicate that if both the top andbottom beam flange to column joints fracture in a large a number of connections,a very significant increase in the probability of global building collapse occurs. Therefore, to reduce the costs associated with post-earthquake inspections, withthe publication of Interim Guidelines Advisory No.2 it is recommended that post-earthquake inspections initially be limited to visual inspection of the beam bottom



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flange to column joint region. If there is evidence of potential damage in thisregion that is not directly observable by visual means, for example, a gap betweenthe weld backing and column flange, then supplemental investigations of this jointshould be conducted using NDT. Similarly, if it is determined that fractures haveoccurred at the beam bottom flange joint, then inspections of that connectionshould be extended to encompass the entire connection including the top beamflange joint, the shear tab and column panel zone. This approach was permittedas an alternate, in the original publication of the Interim Guidelines.

Some engineers have suggested an alternative approach consisting of visual -only inspections, limited to the girder bottom flange to column joint, but for avery large percentage of the total connections in the building. These bottomflange joint connections can be visually inspected with much less fireproofingremoved from the framing surfaces. When significant damage is found at theexposed bottom connection, then additional fireproofing is removed to allow fullexposure of the connection and inspection of the remaining surfaces. Theseengineers feel that by inspecting more connections, albeit to a lesser scope thanrecommended in these Interim Guidelines, their ability to locate the most severeoccurrences of damage in a building is enhanced. These engineers use NDTassisted inspection on a very small sample of the total connections exposed toobtain an indication of the likelihood of hidden problems including damage types.

If properly executed, such an approach can provide sufficient information toevaluate the post-earthquake condition of a building and to make appropriateoccupancy, structural repair and/or modification decisions. It is important thatthe visual inspector be highly trained and that visual inspections be carefullyperformed, preferably by a structural engineer. Casual observation may missclues that hidden damage exists. If, as a result of the partial visual inspection,there is any reason to believe that damage exists at a connection (such as smallgaps between the CJP weld backing and column face), then complete inspectionof the suspected connection, in accordance with the recommendations of theseInterim Guidelines should be performed. If this approach is followed, it isrecommended that a significantly larger sample of connections than otherwiserecommended by these Interim Guidelines, perhaps nearly all of the connections,be inspected.

5.1.2 Gravity Connections


5.1.3 Other Connection Types




5-4

5.2 Preparation

5.2.1 Preliminary Document Review and Evaluation

5.2.1.1 Document Collection and Review


5.2.1.2 Preliminary Building Walk-Through.


5.2.1.3 Structural Analysis


5.2.1.4 Vertical Plumbness Check


5.2.2 Connection Exposure

Pre-inspection activities to expose and prepare a connection for inspection should include thelocal removal of suspended ceiling panels or (as applicable) local demolition of permanent ceilingfinish to access the connection; and cleaning of sufficient fireproofing from the beam and columnsurfaces to allow visual observation of the area to be inspected. If initial inspections are to belimited to the beam bottom flange to column joint and the surrounding material, fireproofingshould be removed from the connection as indicated in Figure 5-1a. Removal of fireproofing needonly be sufficient to permit observation of the surfaces of base and weld metals. Wire brushingand cleaning to remove all particles of fireproofing material is not necessary unless ultrasonictesting of the joint area is to be conducted. In the event that damage is found at the bottom beamflange to column joint, then additional fireproofing should be removed, as indicated in Figure 5-1b, to expose the column panel zone, the column flange, continuity plates, beam web and flanges. The extent of the removal of fireproofing should be sufficient to allow adequate inspection of thesurfaces to be inspected. Figure 5-1b suggests a pattern that will allow both visual and NDTinspection of the top and bottom beam flange to column joints, the beam web and shearconnection, column panel zone and continuity plates, and column flanges in the areas of highestexpected demands. The maximum extent of the removal of fireproofing need not be greater thana distance equal to the beam depth "d" into the beam span to expose evidence of any yielding.



5-5

6”

6”6”

Fireproofing

Exposed surfaces

Figure 5-1a Recommended Zone for Fireproofing Removal for Initial Inspections

6”

6”12”

Fireproofing

Figure 5-1b Recommended Zone for Removal of Fireproofing for Complete Inspections

Commentary: If inspection is to be limited to visual observation of the surfaces ofthe base metal and welds, cleaning of fireproofing need only be sufficient toexpose these surfaces. However, if ultrasonic testing is to be performed, thesurface over which the scanning will be performed must be free Cleaning of weldareas and removal of mill scale and weld spatter. Such cleaning should be donewith care, preferably using a power wire brush, to ensure a clean surface thatdoes not affect the accuracy of ultrasonic testing. The resulting surface finishshould be clean, free of mill scale, rust and foreign matter. The use of a chiselshould be avoided to preclude scratching the steel surfaces which could bemistaken for yield lines. Sprayed-on fireproofing on WSMFs erected prior toabout 19801970 is likely to contain asbestos and should be handled according to



5-6

applicable standards for the removal of hazardous materials. Health hazardsassociated with asbestos were recognized by industry in the late 1960s and by1969, most commercial production of asbestos containing materials had ceased. In April, 1973, the federal government formally prohibited the production ofasbestos containing materials with the adoption of the National EmissionStandards for Hazardous Air Pollutants. Allowing for shelf life of materialsproduced prior to that date, it should be considered possible that buildingsconstructed prior to 1975 contain some asbestos hazards. To preclude physicalexposure to hazardous materials and working conditions in such buildings, thestructural engineer should require by contractual agreement with the buildingowner, prior to the start of the inspection program, that the building ownerdeliver to the structural engineer for his/her review and files a laboratorycertificate that confirms the absence of asbestos in structural steel fireproofing,local pipe insulation, ceiling tiles, and drywall joint compound.

The pattern of fireproofing removal indicated in Figure 5-1 is adequate toallow visual and UT inspection of the top and bottom girder flange to columnjoints, the beam web and shear connection and the column panel zone. Asdiscussed in the commentary to Section 5.1.1, some engineers prefer to initiallyinspect only the bottom beam flange to column joint. In such cases, the initialremoval of fireproofing can be more limited than indicated in the figure. If afterinitial inspection, damage at a connection is suspected, then full removal, asindicated in the figure, should be performed to allow inspection of all areas of theconnection.

5.3 Inspection Program

5.3.1 Visual Inspection (VI)


5.3.1.1 Top Flange


5.3.1.2 Bottom Flange


5.3.1.3 Column and Continuity Plates




5-7

5.3.1.4 Beam Web Shear Connection


5.3.2 Nondestructive Testing (NDT)

NDT should may be used to supplement the visual inspection of connections selected inaccordance with the Interim Guidelines of Chapter 4. The testing agency and NDT personnelperforming this work should conform to the qualifications indicated in Chapter 11 of these InterimGuidelines. The following NDT techniques should may be used at the top and bottom of eachconnection, where accessible, to supplement visual inspection: These techniques should be usedwhenever visual inspection indicates the potential for damage that is not directly observable.

a) Magnetic particle testing (MT) of the beam flange to column flange weld surfaces may beused to confirm the presence of suspected surface cracks based on visual evidence. Wherefractures are evident from visual inspection, MT should be used to confirm the lateralextent of the fracture.All surfaces which were visually inspected should be tested using themagnetic particle technique.

Commentary: The color of powder should be selected to achieve maximumcontrast to the base and weld metal under examination. The test may be furtherenhanced by applying a white coating made specifically for MT or by applyingpenetrant developer prior to the MT examination. This background coatingshould be allowed to thoroughly dry before performing the MT.

b) Ultrasonic testing (UT) may be used to detect the presence of hidden fractures, wherevisual inspection reveals the potential for such fractures. of all faces at the beam flangewelds and adjacent column flanges (extending at least 3 inches above and below thelocation of the CJP weld, along the face of the column, but not less than 1-1/2 times thecolumn flange thickness).

Commentary: The purpose of UT is to 1) locate and describe the extent ofinternal defects not visible on the surface and 2) to determine the extent of cracksobserved visually and by MT. These guidelines recommend the use of visualinspection as the primary tool for detecting earthquake damage (See commentaryto Sec. 5..1.1). UT can be a useful technique for confirmation of the presence ofsuspected fractures at the beam flange to column flange joints. Visual evidencethat may suggest the need for such testing could include apparent separation ofthe base of the weld backing from the face of the column.

Requirements and acceptance criteria for NDT should be as given in AWS D1.1-98 Sections 6and 8. Acceptance or rejection of planar weld discontinuity (cracks, slag inclusion, or lack offusion), including root indications, should, as a minimum, be consistent with AWS DiscontinuitiesSeverity Class designations of cracks and defects per Table 8.26.2 of AWS D1.1-98 for Static



5-8

Structures. Beam flange welds should be tested as "tension welds" per AWS D1.1 Table 8.15.3,Note 3. Backing bars need not be removed prior to performing UT.

Commentary: The value of UT for locating small discontinuities at the root ofbeam flange to column flange welds when the backing is left in place is notuniversally accepted. The reliability of this technique is particularly questionableat the center of the joint, where the beam web obscures the signal. There havebeen a number of reported instances of UT detected indications which were notfound upon removal of the backing, and similarly, there have been reportedinstances of defects which were missed by UT examination but were evident uponremoval of the backing. The smaller the defect, the less likely it is that UT alonewill reliably detect its presence.

Despite the potential inaccuracies of this technique, it is the only methodcurrently available, short of removal of the backing, to find subsurface damage inthe welds. It is also the most reliable method for finding lamellar problems in thecolumn flange (type C4 and C5 damage) opposite the girder flange. Removal ofweld backing at these connections results in a significant cost increase that isprobably not warranted unless UT indicates widespread, significant defectsand/or damage in the building.

The proper scanning techniques, beam angle(s) and transducer sizes should be used asspecified in the written UT procedure contained in the Written Practice, prepared in accordancewith Section 5.3.3 of these Interim Guidelines. The acceptance standard should be that specifiedin the original contract documents, but in no case should it be less than the acceptance criteria ofAWS D1.1, Chapter 8, for Statically Loaded Structures.

The base metal should be scanned with UT for cracks. Cracks which have propagated to thesurface of the weld or beam and column base metal will probably have been detected by visualinspection and magnetic particle tests performed earlier. The purpose of ultrasonic testing of thebase metal is to:

1. Locate and describe the extent of internal indications not apparent on the surface and,

2. Determine the extent of cracks found visually and by magnetic particle test.

Commentary: Liquid dye penetrant testing (PT) may be used where MT isprecluded due to geometrical conditions or restricted access. Note that morestringent requirements for surface preparation are required for PT than MT, perAWS D1.1.

If practical, NDT should be performed across the full width of the bottombeam flange joint. However, if there are no discontinuity signals from UT of



5-9

accessible faces on one side of the bottom flange weld, obstructions on the otherside of the connection need not be removed for testing of the bottom flange weld.

Slabs, flooring and roofing need not be removed to permit NDT of the topflange joint unless there is significant visible damage at the bottom beam flange,adjacent column flange, column web, or shear connection. Unless such damageis present, NDT of the top flange should be performed as permitted, without localremoval of the diaphragms or perimeter wall obstructions.

It should be noted that UT is not 100% effective in locating discontinuitiesand defects in CJP beam flange to column flange welds. The ability of UT toreliably detect such defects is very dependent on the skill of the operator and thecare taken in the inspection. Even under perfect conditions, it is difficult toobtain reliable readings of conditions at the center of the beam flange to columnflange connection as return signals are obscured by the presence of the beamweb. If backing is left in place on the welds, UT becomes even less reliable. There have been a number of reported instances in which UT indicated apparentdefects, that were found not to exist upon removal of the backing. Similarly, UThas failed in some cases to locate defects that were later discovered upon removalof the backing. Additional information on UT may be found in AWS B1.10.

5.3.3 Inspector Qualification

5.3.4 Post-Earthquake Field Inspection Report


5.3.5 Written Report




5-10



Post-earthquake Repair and Modification

6-1

6. POST-EARTHQUAKE REPAIR AND MODIFICATION

6.1 Scope


6.2 Shoring


6.3 Repair Details


6.4 Preparation


6.5 Execution


6.6 STRUCTURAL MODIFICATION

6.6.1 Definition of Modification


6.6.2 Damaged vs. Undamaged Connections


6.6.3 Criteria

Connection modification intended to permit inelastic frame behavior should be proportionedso that the required plastic deformation of the frame may be accommodated through thedevelopment of plastic hinges at pre-determined locations within the girder spans, as indicated inFigure 6-12 Figure 6.6.3-1. Beam-column connections should be designed with sufficientstrength (through the use of cover plates, haunches, side plates, etc.) to force development of theplastic hinge away from the column face. This condition may also be attained through localweakening of the beam section, at the desired location for plastic hinge formation. All elementsof the connection should have adequate strength to develop the forces resulting from the



6-2

formation of the plastic hinge at the predetermined location, together with forces resulting fromgravity loads.

Plastic Hinges

Deformed frame shapeUndeformedframe

L’

L

h

drift angle - θ

Figure 6-12 Figure 6.6.3-1 - Desired Plastic Frame Behavior

Commentary: Nonlinear deformation of frame structures is typicallyaccommodated through the development of inelastic flexural or shear strainswithin discrete regions of the structure. At large inelastic strains these regionscan develop into plastic hinges, which can accommodate significant concentratedrotations at constant (or nearly constant) load through yielding at tensile fibersand buckling at compressive fibers. If a sufficient number of plastic hingesdevelop in a frame, a mechanism is formed and the frame can deform laterally ina plastic manner. This behavior is accompanied by significant energydissipation, particularly if a number of members are involved in the plasticbehavior, as well as substantial local damage to the highly strained elements.The formation of hinges in columns, as opposed to beams, is undesirable, as thisresults in the formation of weak story mechanisms with relatively few elementsparticipating, and consequently little energy dissipation occurring. In addition,such mechanisms also result in local damage to critical gravity load bearingelements.

The prescriptive connection contained in the UBC and NEHRP RecommendedProvisions prior to the Northridge Earthquake was based on the assumeddevelopment of plastic hinge zones within the beams at adjacent to the face of thecolumn, or within the column panel zone itself. If the plastic hinge develops inthe column panel zone, the resulting column deformation results in very largesecondary stresses on the beam flange to column flange joint, a condition whichcan contribute to brittle failure. If the plastic hinge forms in the beam, at the faceof the column, this can result in very large through-thickness strain demands on



6-3

the column flange material and large inelastic strain demands on the weld metaland surrounding heat affected zones stress and strain demands on the weldedbeam flange to column flange joint. These conditions can also lead to brittlejoint failure. Although ongoing research may reveal conditions of materialproperties, design and detailing configurations that permit connections withyielding occurring at the column face to perform reliably, for the present, it isrecommended In order to achieve more reliable performance, it is recommendedthat the connection of the beam to the column be modified to be sufficientlystrong to force the inelastic action (plastic hinge) away from the column face.Plastic hinges in steel beams have finite length, typically on the order of half thebeam depth. Therefore, the location for the plastic hinge should be shifted atleast that distance away from the face of the column. When this is done, theflexural demands on the columns are increased. Care must be taken to assurethat weak column conditions are not inadvertently created by local strengtheningof the connections.

It should be noted that connection modifications of the type described above,while believed to be effective in preventing brittle connection fractures, will notprevent structural damage from occurring. Brittle connection fractures areundesirable because they result in a substantial reduction in the lateral-force-resisting strength of the structure which, in extreme cases, can result in instabilityand collapse. Connections modified as described in these Interim Guidelinesshould experience many fewer such brittle fractures than unmodified connections.However, the formation of a plastic hinge within the span of a beam is not acompletely benign event. Beams which have formed such hinges may exhibitlarge buckling and yielding deformation, damage which typically must berepaired. The cost of such repairs could be comparable to the costs incurred inrepairing fracture damage experienced in the Northridge Earthquake. Theprimary difference is that life safety protection will be significantly enhanced andmost structures that have experienced such plastic deformation damage shouldcontinue to be safe for occupancy while repairs are made.

If the types of damage described above are unacceptable for a given building,then alternative methods of structural modification should be considered that willreduce the plastic deformation demands on the structure during a strongearthquake. Appropriate methods of achieving such goals include the installationof supplemental braced frames, energy dissipation systems, and similarsystematic modifications of the building’s basic lateral force resisting system.

It is important to recognize that in frames with relatively short bays, theflexural hinging indicated in Figure 6.6.3-1 may not be able to form. If theeffective flexural length (L’ in the figure) of beams in a frame becomes too short,then the beams or girders will yield in shear before zones of flexural plasticity



6-4

can form, resulting in an inelastic behavior that is more like that of aneccentrically braced frame than that of a moment frame. This behavior mayinadvertently occur in frames in which relatively large strengthened connections,such as haunches, cover plates or side plates have been used on beams withrelatively short spans. This behavior is illustrated in Figure 6.6.3-2.

The guidelines contained in this section are intended to address the design offlexurally dominated moment resisting frames. When utilizing these guidelines, itis important to confirm that the configuration of the structure is such that thepresumed flexural hinging can actually occur. It is possible that shear yielding offrame beams, such as that schematically illustrated in Figure 6.6.3-2 may be adesirable behavior mode. However, to date, there has not been enough researchconducted into the behavior of such frames to develop recommended designguidelines. If modifications to an existing frame result in such a configurationdesigners should consider referring to the code requirements for eccentricallybraced frames. Particular care should be taken to brace the shear link of suchbeams against lateral-torsional buckling and also to adequately stiffen the websto avoid local buckling following shear plastification.

Shear Link

Shear Link

Figure 6.6.3-2 Shear Yielding Dominated Behavior of Short Bay Frames

6.6.4 Strength and Stiffness

6.6.4.1 Strength

When these Interim Guidelines require determination of the strength of a framing element orcomponent, this shall be calculated in accordance with the criteria contained in UBC-94, Section



6-5

2211.4.2 {NEHRP-91 Section 10.2, except that the factor φ should be taken as 1.0}, restated asfollows:

2211.4.1 Member strength. Where this section requires that the strength of the member bedeveloped, the following shall be used:

Flexure Ms = Z Fy

Shear Vs = 0.55 Fy d tAxial compression Psc = 1.7 Fa AAxial tension Pst = Fy AConnectors Full Penetration welds Fy A Partial Penetration welds 1.7 allowable (see commentary) Bolts and fillet welds 1.7 allowable

Alternatively, the criteria contained in the 1997 edition of the AISC Seismic Provisions forStructural Steel Buildings (AISC, 1997) may be used.

Commentary: At the time the Interim Guidelines were first published, they werebased on the 1994 edition of the Uniform Building Code and the 1994 edition ofthe NEHRP Provisions. In the time since that initial publication, more recenteditions of both documents have been published, and codes based on thesedocuments have been adopted by some jurisdictions. In addition, the AmericanInstitute of Steel Construction has adopted a major revision to its SeismicProvisions for Structural Steel Buildings (AISC Seismic Provisions), largelyincorporating, with some modification, the recommendations contained in theInterim Guidelines. This updated edition of the AISC Seismic Provisions hasbeen incorporated by reference into the 1997 edition of the NEHRP Provisionsand has also been adopted by some jurisdictions as an amendment to the modelbuilding codes. Structural upgrades designed to comply with the requirements ofthe 1997 AISC Seismic Provisions may be deemed to comply with the intent ofthese Interim Guidelines. Where reference is made herein to the requirements ofthe 1994 Uniform Building Code or 1994 NERHP Provisions, the parallelprovisions of the 1997 editions may be used instead, and should be used in thosejurisdictions that have adopted codes based on these updated standards.

Partial penetration welds are not recommended for tension applications incritical connections resisting seismic induced stresses. The geometry of partialpenetration welds creates a notch-like condition that can initiate brittle fractureunder conditions of high tensile strain.

Many WSMF structures are constructed with concrete floor slabs that areprovided with positive shear attachment between the slab and the top flanges ofthe girders of the moment-resisting frames. Although not generally accounted forin the design of the frames, the resulting composite action can increase the



6-6

effective strength of the girder significantly, particularly at sections wherecurvature of the girder places the top flange into compression. Although thiseffect is directly accounted for in the design of composite systems, it is typicallyneglected in the design of systems classified as moment resisting steel frames.The increased girder flexural strength caused by this composite action can resultin a number of effects including the unintentional creation of weak column -strong beam and weak panel zone conditions. In addition, this composite effecthas the potential to reduce the effectiveness of reduced section or “dog-bone”type connection assemblies. Unfortunately, very little laboratory testing of largescale connection assemblies with slabs in place has been performed to date andas a result, these effects are not well quantified. In keeping with typicalcontemporary design practice, the design formulae provided in these Guidelinesneglect the strengthening effects of composite action. Designers should, however,be alert to the fact that these composite effects do exist. Similar, and perhapsmore severe, effects may also exist where steel beams support masonry orconcrete walls.

6.6.4.2 Stiffness

Calculation of frame stiffness for the purpose of determining interstory drift under theinfluence of the design lateral forces should be based on the properties of the bare steel frame,neglecting the effects of composite action with floor slabs. The stiffening effects of connectionreinforcements (e.g.: haunches, side plates, etc.) may be considered in the calculation of overallframe stiffness and drift demands. When reduced beam section connections are utilized, thereduction in overall frame stiffness, due to local reductions in girder cross section, should beconsidered.

Commentary: For design purposes, frame stiffness is typically calculatedconsidering only the behavior of the bare frame, neglecting the stiffening effectsof slabs, gravity framing, and architectural elements such as walls. The resultingcalculation of building stiffness and period typically underestimates the actualproperties, substantially. Although this approach can result in unconservativeestimates of design force levels, it typically produces conservative estimates ofinterstory drift demands. Since the design of most moment-resisting frames arecontrolled by considerations of drift, this approach is considered preferable tomethods that would have the potential to over-estimate building stiffness. Also,many of the elements that provide additional stiffness may be subject to rapiddegradation under severe cyclic lateral deformation, so that the bare framestiffness may provide a reasonable estimate of the effective stiffness under longduration ground shaking response.

Notwithstanding the above, designers should be alert to the fact thatunintentional stiffness introduced by walls and other non-structural elements can



6-7

significantly alter the behavior of the structure in response to ground shaking. Ofparticular concern, if these elements are not uniformly distributed throughout thestructure, or isolated from its response, they can cause soft stories and torsionalirregularities, conditions known to result in poor behavior.

6.6.5 Plastic Rotation Capacity

The plastic rotation capacity of modified connections should reflect realistic estimates of therequired level of plastic rotation demand. In the absence of detailed calculations of rotationdemand, connections should be shown to be capable of developing a minimum plastic rotationcapacity on the order of 0.025 to 0.030 radian. The demand may be lower when braced frames,supplemental damping, base isolation, or other elements are introduced into the moment framesystem, to control its lateral deformation; when the design ground motion is relatively low in therange of predominant periods for the structure; and when the frame is sufficiently strong and stiff.

As used in these Guidelines, plastic rotation is defined as the plastic chord rotation angle. Theplastic chord rotation angle is calculated using the rotated coordinate system shown in Fig. 6.6.5-1 as the plastic deflection of the beam or girder, at the point of inflection (usually at the center ofits span,) ∆CI, divided by the distance between the center of the beam span and the centerline ofthe panel zone of the beam column connection, LCL. This convention is illustrated in Figure 6.6.5-1.

It is important to note that this definition of plastic rotation is somewhat different than theplastic rotation that would actually occur within a discrete plastic hinge in a frame model similarto that shown in Figure 6.6.3-1. These two quantities are related to each other, however, and ifone of them is known, the other may be calculated from Eq. 6.6.5-1.

cL

cL

Beam span center line

∆CL

LCL

θ pCL

CLL= ∆

Plastic hinge

lh

Figure 6.6.5-1 Calculation of Plastic Rotation Angle



6-8

( )θ θp ph

CL h

CL

L lL

= − (6.6.5-1)

where: θp is the plastic chord angle rotation, as used in these Guidelines θph is the plastic rotation, at the location of a discrete hinge LCL is the distance from the center of the beam span to the center of the beam-column assembly panel zone lh is the assumed location of the discrete plastic hinge relative to the center of the beam-column assembly panel zone

If calculations are performed to determine the required connection plastic rotation capacity,the capacity should be taken somewhat greater than the calculated deformation demand, due tothe high variability and uncertainty inherent in predictions of inelastic seismic response. Untilbetter guidelines become available, a required plastic rotation capacity on the order of 0.005radians greater than the demand calculated for the design basis earthquake (or if greaterconservatism is desired - the maximum capable considered earthquake) is recommended.Rotation demand calculations should consider the effect of plastic hinge location within the beamspan, as indicated in Figure 6-12 Figure 6.6.3-1, on plastic rotation demand. Calculations shouldbe performed to the same level of detail specified for nonlinear dynamic analysis for base isolatedstructures in UBC-94 Section 1655 {NEHRP-94 Section 2.6.4.4}. Ground motion time historiesutilized for these nonlinear analyses should satisfy the scaling requirements of UBC-94 Section1655.4.2 {NEHRP-94 Section 2.6.4.4} except that instead of the base isolated period, TI, thestructure period, T, calculated in accordance with UBC-94 Section 1628 {NEHRP-94 Section2.3.3.1} should be used.

Commentary. When the Interim Guidelines were first published, the plasticrotation was defined as that rotation that would occur at a discrete plastic hinge,similar to the definition of θph. in Eq. 6.6.5-1, above. In subsequent testing ofprototype connection assemblies, it was found that it is often very difficult todetermine the value of this rotation parameter from test data, since actual plastichinges do not occur at discrete points in the assembly and because some amountof plasticity also occurs in the panel zone of many assemblies. The plastic chordangle rotation, introduced in Interim Guidelines Advisory No. 1, may morereadily be obtained from test data and also more closely relates to the driftexperienced by a frame during earthquake response.

Traditionally, structural engineers have calculated demand in momentframes by sizing the members for strength and drift using code forces (eitherequivalent static or reduced dynamic forces) and then "developing the strength ofthe members." Since 1988, "developing the strength" has been accomplished byprescriptive means. It was assumed that the prescribed connections would bestrong enough so that the girder would yield (in bending), or the panel zone



6-9

would yield (in shear) in a nearly perfectly plastic manner producing the plasticrotations necessary to dissipate the energy of the earthquake. It is now knownthat the prescriptive connection is often incapable of behaving in this manner.

In the 1994 Northridge earthquake, many moment-frame connectionsfractured with little evidence of plastic hinging of the girders or yielding of thecolumn panel zones. Testing of moment frame connections both prior to andsubsequent to the earthquake suggests that the standard welded flange-bolted webconnection is unable to reliably provide plastic rotations beyond about 0.005radian for all ranges of girder depths and often fails below that level. Thus, forframes designed for code forces and for the code drift limits, new connectionconfigurations must be developed to reliably accommodate such rotation withoutbrittle fracture.

In order to develop reasonable estimates of the plastic rotation demands on aframe’s connections, it is necessary to perform inelastic time history analyses.For regular structures, approximations of the plastic rotation demands can beobtained from linear elastic analyses. Analytical research (Newmark and Hall -1982) suggests that for structures having the dynamic characteristics of mostWSMF buildings, and for the ground motions typical of western US earthquakes,the total frame deflections obtained from an unreduced (no R or Rw factor)dynamic analysis provide an approximate estimate of those which would beexperienced by the inelastic structure. For the typical spectra contained in thebuilding code, this would indicate expected drift ratios on the order of 1%. Thedrift demands in a real structure, responding inelastically, tend to concentrate ina few stories, rather than being uniformly distributed throughout the structure’sheight. Therefore, it is reasonable to expect typical drift demands in individualstories on the order of 1.5% to 2% of the story height. As a roughapproximation, the drift demand may be equated to the joint rotation demand,yielding expected rotation demands on the order of perhaps 2%. Since there isconsiderable variation in ground motion intensity and spectra, as well as theinelastic response of buildings to these ground motions, conservatism in selectionof an appropriate connection rotation demand is warranted.

In recent testing of large scale subassemblies incorporating modifiedconnection details, conducted by SAC and others, when the connection design wasable to achieve a plastic rotation demand of 0.025 radians or more for severalcycles, the ultimate failure of the subassembly generally did not occur in theconnection, but rather in the members themselves. Therefore, the statedconnection capacity criteria would appear to result in connections capable ofproviding reliable performance.

It should be noted that the connection assembly capacity criteria for themodification of existing buildings, recommended by these Interim Guidelines, is



6-10

somewhat reduced compared to that recommended for new buildings (Chapter 7).This is typical of approaches normally taken for existing structures. For newbuildings, these Interim Guidelines discourage building-specific calculation ofrequired plastic rotation capacity for connections and instead, encourage thedevelopment of highly ductile connection designs. For existing buildings, such anapproach may lead to modification designs that are excessively costly, as well asthe modification of structures which do not require such modification.Consequently, an approach which permits the development of semi-ductileconnection designs, with sufficient plastic rotation capacity to withstand theexpected demands from a design earthquake is adopted. It should be understoodthat buildings modified to this reduced criteria will not have the same reliabilityas new buildings, designed in accordance with the recommendations of Chapter7. The criteria of Chapter 7 could be applied to existing buildings, if superiorreliability is desired.

When performing inelastic frame analysis, in order to determine the requiredconnection plastic rotation capacity, it is important to accurately account for thelocations at which the plastic hinges will occur. Simplified models, whichrepresent the hinge as occurring at the face of the column, maywill underestimatethe plastic rotation demand. This problem becomes more severe as the columnspacing, L, becomes shorter and the distance between plastic hinges, L’, agreater portion of the total beam span. Eq. 6.6.5-1 may be used to convertcalculated values of plastic rotation at a hinge remotely located from the column,to the chord angle rotation, used for the definition of acceptance criteriacontained in these Guidelines. In extreme cases, the girder will not form plastichinges at all, but instead, will develop a shear yield, similar to an eccentricbraced frame.

6.6.6 Connection Qualification and Design

Modified girder-column connections may be qualified by testing or designed usingcalculations. Qualification by testing is the preferred approach. Preliminary designs ofconnections to be qualified by test may be obtained using the calculation procedures of Section6.6.6.3. The procedures of that section may also be used to calibrate previous tests of similarconnection configurations to slightly different applications, by extrapolation. Extrapolation of testresults should be limited to connections of elements having similar geometries and materialspecifications as the tested connections. Designs based on calculation alone should be subject toqualified independent third party review.

Commentary: Because of the cost of testing, use of calculations for interpolationor extrapolation of test results is desirable. How much extrapolation should beaccepted is a difficult decision. As additional testing is done, more informationmay be available on what constitutes "conservative" testing conditions, thereby



6-11

allowing easier decisions relative to extrapolating tests to actual conditions whichare likely to be less demanding than the tests. For example, it is hypothesizedthat connections of shallower, thinner flanged members are likely to be morereliable than similar connections consisting of deeper, thicker flanged members.Thus, it may be possible to test the largest assemblages of similar details andextrapolate to the smaller member sizes? - at least within comparable membergroup families. However, there is evidence to suggest that extrapolation of testresults to assemblies using members of reduced size is not always conservative.In a recent series of tests of cover plated connections, conducted at the Universityof California at San Diego, a connection assembly that produced acceptableresults for one family of beam sizes, W24, did not behave acceptably when thebeam depth was reduced significantly to W18. In that project, the change inrelative flexibilities of the members and connection elements resulted in a shift inthe basic behavior of the assembly and initiation of a failure mode that was notobserved in the specimens with larger member sizes. In order to minimize thepossibility of such occurrences, when extrapolation of test results is performed, itshould be done with a basic understanding of the behavior of the assembly, andthe likely effects of changes to the assembly configuration on this behavior. Testresults should not be extrapolated to assembly configurations that are expected tobehave differently than the tested configuration. Extrapolation or interpolationof results with differences in welding procedures, details or material properties iseven more difficult.

6.6.6.1 Qualification Test Protocol


6.6.6.2 Acceptance Criteria

The minimum acceptance criteria for connection qualification for specimens tested inaccordance with these Interim Guidelines should be as follows:

a) The connection should develop beam plastic rotations as indicated in Section 6.6.5, forat least one complete cycle.

b) The connection should develop a minimum strength equal to 80% of the plasticstrength of the girder, calculated using minimum specified yield strength Fy,throughout the loading history required to achieve the required plastic rotationcapacity, as indicated in a), above.

c) The connection should exhibit ductile behavior throughout the loading history. Aspecimen that exhibits a brittle limit state (e.g. complete flange fracture, columncracking, through-thickness failures of the column flange, fractures in welds subject to



6-12

tension, shear tab cracking, etc. ) prior to reaching the required plastic rotation shall beconsidered unsuccessful.

d) Throughout the loading history, until the required plastic rotation is achieved, theconnection should be judged capable of supporting dead and live loads required by thebuilding code. In those specimens where axial load is applied during the testing, thespecimen should be capable of supporting the applied load throughout the loadinghistory.

The evaluation of the test specimen’s performance should consistently reflect the relevant limitstates. For example, the maximum reported moment and the moment at the maximum plasticrotation are unlikely to be the same. It would be inappropriate to evaluate the connection usingthe maximum moment and the maximum plastic rotation in a way that implies that they occurredsimultaneously. In a similar fashion, the maximum demand on the connection should beevaluated using the maximum moment, not the moment at the maximum plastic rotation unless thebehavior of the connection indicated that this limit state produced a more critical condition in theconnection.

Commentary: Many connection configurations will be able to withstandplastic rotations on the order of 0.025 radians or more, but will have sustainedsignificant damage and degradation of stiffness and strength in achieving thisdeformation. The intent of the acceptance criteria presented in this Section is toassure that when connections experience the required plastic rotation demand,they will still have significant remaining ability to participate in the structure’slateral load resisting system.

In evaluating the performance of specimens during testing, it is important todistinguish between brittle behavior and ductile behavior. It is not uncommon forsmall cracks to develop in specimens after relatively few cycles of inelasticdeformation. In some cases these initial cracks will rapidly lead to ultimatefailure of the specimen and in other cases they will remain stable, perhapsgrowing slowly with repeated cycles, and may or may not participate in theultimate failure mode. The development of minor cracks in a specimen, prior toachievement of the target plastic rotation demand should not be cause forrejection of the design if the cracks remain stable during repeated cycling.Similarly, the occurrence of brittle fracture at inelastic rotations significantly inexcess of the target plastic rotation should not be cause for rejection of thedesign.

6.6.6.3 Calculations




6-13

6.6.6.3.1 Material Strength Properties

In the absence of project specific material property information (for example, mill testreports), the values listed in Table 6-3 Table 6.6.6.3.1-1 should be used to determine the strengthof steel shape and plate for purposes of calculation. The permissible strength for weld metalshould be taken in accordance with the building code.

Table 6-3Table 6.6.6.3-1 - Properties for Use in Connection Modification Design

Material Fy (ksi) Fy m (ksi) Fu (ksi)A36 Beam 36 1 1

Dual Certified Beam Axial, Flexural Shape Group 1 Shape Group 2 Shape Group 3 Shape Group 4 Through-Thickness

50

-

552

582

572

542

-

65 min.

Note 3A572 Column/Beam Axial, Flexural Shape Group 1 Shape Group 2 Shape Group 3 Shape Group 4 Shape Group 5 Through-Thickness

50

-

582

582

572

572

552

-

65 min.

Note 3A992 Structural Shape1 Use same values as for A572, Gr. 50Notes:1. See Commentary2. Based on coupons from web. For thick flanges,

the Fy flange is approximately 0.95 Fy web.3. See Commentary

Commentary: Table 6-3, Note 1 - The material properties for steel nominallydesignated on the construction documents as ASTM A36 can be highly variableand in recent years, steel meeting the specified requirements for both ASTM A36and A572 has routinely been incorporated in projects calling for A36 steel.Consequently, unless project specific data is available to indicate the actualstrength of material incorporated into the project, the properties for ASTM A572steel should be assumed when ASTM A36 is indicated on the drawings, and theassumption of a higher yield stress results in a more severe design condition.

The ASTM A992 specification was specifically developed by the steel industryin response to expressed concerns of the design community with regard to thepermissible variation in chemistry and mechanical properties of structural steelunder the A36 and A572 specifications. This new specification, which wasadopted in late 1998, is very similar to ASTM A572, except that it includessomewhat more restrictive limits on chemistry and on the permissible variation in



6-14

yield and ultimate tensile stress, as well as the ratio of yield to tensile strength.At this time, no statistical data base is available to estimate the actualdistribution of properties of material produced to this specification. However, theproperties are likely to be very similar, albeit with less statistical scatter, to thoseof material recently produced under ASTM A572, Grade 50.

Table 6-3Table 6.6.6.3-1, Note 3 - In the period immediately following theNorthridge earthquake, the Seismology Committee of the Structural EngineersAssociation of California and the International Conference of Building Officialsissued Interim Recommendation No. 2 (SEAOC-1995) to provide guidance on thedesign of moment resisting steel frame connections. Interim RecommendationNo. 2 included a recommendation that the through-thickness stress demand oncolumn flanges be limited to a value of 40 ksi, applied to the projected area ofbeam flange attachment. This value was selected somewhat arbitrarily, to ensurethat through-thickness yielding did not initiate in the column flanges of moment-resisting connections and because it was consistent with the successful tests ofassemblies with cover plates conducted at the University of Texas at Austin(Engelhardt and Sabol - 1994), rather than being the result of a demonstratedthrough-thickness capacity of typical column flange material. Despite thesomewhat arbitrary nature of the selection of this value, its use often controls theoverall design of a connection assembly including the selection of cover platethickness, haunch depth, and similar parameters.

It would seem to be important to prevent the inelastic behavior of connectionsfrom being controlled by through-thickness yielding of the column flanges. Thisis because it would be necessary to develop very large local ductilities in thecolumn flange material in order to accommodate even modest plastic rotationdemands on the assembly. However, extensive investigation of the through-thickness behavior of column flanges in a “T” joint configuration reveals thatneither yielding, nor through-thickness failure are likely to occur in theseconnections. Barsom and Korvink (1997) conducted a statistical survey ofavailable data on the tensile strength of rolled shape material in the through-thickness direction. These tests were generally conducted on small diametercoupons, extracted from flange material of heavy shapes. The data indicates thatboth the yield stress and ultimate tensile strength of this material in the through-thickness direction is comparable to that of the material in the direction parallelto rolling. However, it does indicate somewhat greater scatter, with a number ofreported values where the through-thickness strength was higher, as well as lowerthan that in the longitudinal direction. Review of this data indicates with highconfidence that for small diameter coupons, the yield and ultimate tensile valuesof the material in a through-thickness direction will exceed 90% and 80%respectively of the comparable values in the longitudinal direction. theThe causesfor through-thickness failures of column flanges (types C2, C4, and C5), observed



6-15

both in buildings damaged by the Northridge Earthquake and in some testspecimens, are not well understood. They are thought to be a function of themetallurgy and “purity” of the steel; conditions of loading including the presenceof axial load and rate of loading application; conditions of tri-axial restraint;conditions of local hardening and embrittlement within the weld’s heat affectedzone; stress concentrations induced by the presence of backing bars and defectsat the root of beam flange to column flange welds; and by the relationship of theconnection components as they may affect flange bending stresses and flangecurvature induced by panel zone yielding. Given the many complex factors whichcan affect the through-thickness strength of the column flange, determination of areliable basis upon which to set permissible design stresses will requiresignificant research. Such research is currently being conducted under the SACphase II program.

While this statistical distribution suggests the likelihood that the through-thickness strength of column flanges could be less than the flexural strength ofattached beam elements, testing of more than 40 specimens at Lehigh Universityindicates that this is not the case. In these tests, high strength plates,representing beam flanges and having a yield strength of 100 ksi were welded tothe face of A572, Grade 50 and A913, Grade 50 column shapes, to simulate theportion of a beam-column assembly at the beam flange. These specimens wereplaced in a universal testing machine and loaded to produce high through-thickness tensile stresses in the column flange material. The tests simulated awide range of conditions, representing different weld metals as well and also toinclude eccentrically applied loading. In 40 of 41 specimens tested, the assemblystrength was limited by tensile failure of the high strength beam flange plate asopposed to the column flange material. In the one failure that occurred withinthe column flange material, fracture initiated at the root of a low-toughness weld,at root defects that were intentionally introduced to initiate such a fracture.

The behavior illustrated by this test series is consistent with mechanics ofmaterials theory. At the joint of the beam flange to column flange, the material isvery highly restrained. As a result of this, both the yield strength and ultimatetensile strength of the material in this region is significantly elevated. Underthese conditions, failure is unlikely to occur unless a large flaw is present that canlead to unstable crack propagation and brittle fracture. In light of this evidence,Interim Guidelines Advisory No. 2 deletes any requirement for evaluation ofthrough-thickness flange stress in columns.

Interim Recommendation No. 2 (SEAOC-1995) included a value of 40 ksi,applied to the projected area of beam flange attachment, for the through-thickness strength to be used in calculations. This value was selected because itwas consistent with the successful tests of cover plated assemblies conducted at



6-16

the University of Texas at Austin (Engelhardt and Sabol - 1994). However,because of the probable influence of all the factors noted above, this value canonly be considered to reflect the specific conditions of those tests and specimens.

Although reduced stresses at the column face produced acceptable results inthe University of Texas tests, the key to that success was more likely the result offorcing the plastic hinge away from the column than reduction of the through-thickness stress by the cover plates. Reduction of through-thickness columnflange stress to ever lower levels by the use of thicker cover plates is notrecommended, since such cover plates will result in ever higher forces on the faceof the column flange as well as larger weldments with potential for enlarged heataffected zones, higher residual stresses and conditions of restraint.

Since the initial publication of the Interim Guidelines, a significant number oftests have been performed on reduced beam section connections (See section7.5.3), most of which employed beam flanges which were welded directly to thecolumn flanges using improved welding techniques, but without reinforcementplates. No through-thickness failures occurred in these tests despite the fact thatcalculated through-thickness stresses at the root of the beam flange to columnflange joint ranged as high as 58 ksi. The successful performance of these weldedjoints is most probably due to the shifting of the yield area of the assembly awayfrom the column flange and into the beam span. Based on the indications of theabove described tests, and noting the undesirability of over reinforcingconnections, it is now suggested that a maximum through-thickness stress of0.9Fyc may be appropriate for use with connections that shift the hinging awayfrom the column face. Notwithstanding this recommendation, engineers are stillcautioned to carefully consider the through-thickness issue when these otherpreviously listed conditions which are thought to be involved in this type offailure are prevalent.

Notwithstanding all of the above, successful tests using cover plates and othermeasures of moving hinges (and coincidentally reducing through-thickness stress)continue to be performed. In the interim, structural engineers choosing to utilizeconnections relying on through-thickness strength should recognize that despitethe successful testing, connections relying on through-thickness strength can notbe considered to be fully reliable until the influence of the other parametersdiscussed above can be fully understood. A high amount of structuralredundancy is recommended for frames employing connections which rely onthrough-thickness strength of the column flange.

6.6.6.3.2 Determine Plastic Hinge Location

The desired location for the formation of plastic hinges should be determined as a basicparameter for the calculations. For beams with gravity loads representing a small portion of the



6-17

total flexural demand, the location of the plastic hinge may be assumed to occur as indicated inTable 6.6.6.3.2-1 and illustrated in Figure 6.6.6.3.2-1, at a distance equal to 1/3 of the beam depthfrom the edge of the reinforced connection (or start of the weakened beam section), unlessspecific test data for the connection indicates that a different value is appropriate. Refer to Figure6-13.

Table 6.6.6.3.2-1 Plastic Hinge Location - Strengthened Connections

Connection Type Reference Section Hinge Location “sh”

Cover plates Sect. 7.9.1 d/4 beyond end of cover plates

Haunches Sect. 7.9.3, 7.9.4 d/3 beyond toe of haunch

Vertical Ribs Sect. 7.9.2 d/3 beyond toe of ribs

L

Bea

m d

epth

- d

Edge

of r

einf

orce

dco

nnec

tion

Edge

of r

einf

orce

dco

nnec

tion

sh=d/3

L’

Plastichinge

Connectionreinforcementsh=

d/4

Figure 6-13 Figure 6.6.6.3.2-1 - Location of Plastic Hinge

Commentary: The suggested locations for the plastic hinge, at a distance d/3away from the end of the reinforced section indicated in Table 6.6.6.3.2-1 andFigure 6.6.6.3.2-1 are is based on the observed behavior of test specimens, withno significant gravity load present. If significant gravity load is present, this canshift the locations of the plastic hinges, and in the extreme case, even change theform of the collapse mechanism. If flexural demand on the girder due to gravityload is less than about 30% of the girder plastic capacity, this effect can safely beneglected, and the plastic hinge locations taken as indicated. If gravity demandssignificantly exceed this level then plastic analysis of the girder should be



6-18

performed to determine the appropriate hinge locations. Note that in zones ofhigh seismicity (UBC Zones 3 and 4, and NEHRP Map Areas 6 and 7) gravityloading on the girders of earthquake resisting frames typically has a very smalleffect.

6.6.6.3.3 Determine Probable Plastic Moment at Hinges

The probable value of the plastic moment, Mpr, at the location of the plastic hinges should bedetermined from the equation:

M 0.95 Z Fpr b ya= a (6-1)

M 1.1Z Fpr b ya= (6.6.6.3.3-1)

where: α is a coefficient that accounts for the effects of strain hardening and modelinguncertainty, taken as:

1.1 when qualification testing is performed or calculations are correlated with previous qualification testing

1.3 when design is based on calculations, alone.

Fya is the actual yield stress of the material, as identified from mill test reports. Wheremill test data for the project is not traceable to specific framing elements, theaverage of mill test data for the project for the given shape may be used. Whenmill test data for the project is not available, the value of Fym, fromtable 6-3Table 6.6.6.3-1 may be used.

Zb is the plastic modulus of the section

Commentary: The 1.10.95 factor, in equation 6.6.6.3.3-1, is used to adjustaccount for two effects. First, it is intended to account for the typical differencebetween the yield stress in the beam web, where coupons for mill certificationtests are normally extracted, andto the value in the beam flange. Beam flanges,being comprised of thicker material, typically have somewhat lower yieldstrengths than do beam web material. Second, it is intended toThe factor of 1.1recommended to account for strain hardening, or other sources of strength aboveyield, and agrees fairly well with available test results. It should be noted that the1.1 factor could underestimate the over-strength where significant flangebuckling does not act as the gradual limit on the connection. Nevertheless, the1.1 factor seems a reasonable expectation of over-strength considering thecomplexities involved.



6-19

Connection designs that result in excessive strength in the girder connectionrelative to the column or excessive demands on the column panel zone are notexpected to produce superior performance. There is a careful balance that mustbe maintained between developing connections that provide for an appropriateallowance for girder overstrength and those that arbitrarily increase connectiondemand in the quest for a “conservative” connection design. The factorssuggested above were chosen in an attempt to achieve this balance, and arbitraryincreases in these values are not recommended.

When the Interim Guidelines were first published, Eq. 6.6.6.3.3-1 included acoefficient, α, intended to account both for the effects of strain hardening andalso for modeling uncertainty when connection designs were based oncalculations as opposed to a specific program of qualification testing. The intentof this modeling uncertainty factor was twofold: to provide additionalconservatism in the design when specific test data for a representative connectionwas not available, and also as an inducement to encourage projects to undertakeconnection qualification testing programs. After the Interim Guidelines had beenin use for some time, it became apparent that this approach was not an effectiveinducement for projects to perform qualification testing, and also that the use ofan overly large value for the α coefficient often resulted in excessively largeconnection reinforcing elements (cover plates, e.g.) and other design features thatdid not appear conducive to good connection behavior. Consequently, it wasdecided to remove this modeling uncertainty factor from the calculation of theprobable strength of an assembly.

6.6.6.3.4 Determine Beam Shear

The shear in the beam at the location of the plastic hinge should be determined. A free bodydiagram of that portion of the beam located between plastic hinges is a useful tool for obtainingthe shear at each plastic hinge. Figure 6-14Figure 6.6.3.4-1 provides an example of such acalculation.



6-20

L

sh

L’

Plastichinge P

L/2

P

Mpr MprL’Vp

taking the sum of moments about “A” = 0Vp ={Mpr + Mpr + P L’/2 + wL’2/2}/L’

“A”

VAw

Note: if 2Mpr /L’ is less then the gravity shear in the free body (in thiscase P/2 + wL’/2),then the plastic hinge location will shift and L’must be adjusted, accordingly

Figure 6-14 Figure 6.6.3.4-1 - Sample Calculation of Shear at Plastic Hinge

6.6.6.3.5 Determine Strength Demands on Connection

In order to complete the design of the connection, including sizing the various plates andjoining welds which make up the connection, it is necessary to determine the shear and flexuralstrength demands at each critical section. These demands may be calculated by taking a free bodyof that portion of the connection assembly located between the critical section and the plastichinge. Figure 6-15 Figure 6.6.3.5-1 demonstrates this procedure for two critical sections, for thebeam shown in Figure 6-14Figure 6.6.3.4-1.

Plastichinge

Vp

Mpr

Plastichinge

Vp

Mpr

x

Mf

x+dc/2

dc

Mf=Mpr +Vpx

Mc

Mc=Mpr +Vp(x+dc/2)

Critical Section at Column Face Critical Section at Column Centerline

Figure 6-15 Figure 6.6.3.5-1 - Calculation of Demands at Critical Sections



6-21

Commentary: Each unique connection configuration may have different criticalsections. The vertical plane that passes through the joint between the beamflanges and column (if such joining occurs) will typically define at least one suchcritical section, used for designing the joint of the beam flanges to the column, aswell as evaluating shear demands on the column panel zone. A second criticalsection occurs at the center line of the column. Moments calculated at this pointare used to check weak beam - strong column conditions. Other critical sectionsshould be selected as appropriate to the connection configuration.

6.6.6.3.6 Check for Strong Column - Weak Beam Condition

Buildings which form sidesway mechanisms through the formation of plastic hinges in thebeams can dissipate more energy than buildings that develop mechanisms consisting primarily ofplastic hinges in the columns. Therefore, if an existing building’s original design was such thathinging would occur in the beams rather than the columns, care should be taken not to alter thisbehavior with the addition of connection reinforcement. To determine if the desired strongcolumn - weak beam condition exists, the connection assembly should be checked to determine ifthe following equation is satisfied:

Z (F f ) M 1.0c yc a c− >∑ ∑ (6.6.6.3.6-12)

where: Zc is the plastic modulus of the column section above and below the connectionFyc is the minimum specified yield stress for the column above and belowfa is the axial load in the column above and belowΣMc is the moment calculated at the center of the column in accordance with

Section 6.6.6.3.5 sum of the column moments at the top and bottom of the panel zone, respectively, resulting from the development of the probable beam plastic moments, Mpr, within each beam in the connection.

Commentary: Equation 6.6.6.3.6-12 is based on the building code provisions forstrong column - weak beam design. The building code provisions for evaluatingstrong column - weak beam conditions presume that the flexural stiffness of thecolumns above and below the beam are approximately equal, that the beams willyield at the face of the column, and that the depth of the columns and beams aresmall relative to their respective span lengths. This permits the code to use arelatively simple equation to evaluate strong column - weak beam conditions inwhich the sum of the flexural capacities of columns at a connection are comparedto the sums of the flexural capacities in the beams. The first publication of theInterim Guidelines took this same approach, except that the definition of ΣMc wasmodified to explicitly recognize that because flexural hinging of the beams wouldoccur at a location removed from the face of the column, the moments deliveredby the beams to the connection would be larger than the plastic moment strengthof the beam. In this equation, ΣMc was taken as the sum of the moments at the



6-22

center of the column, calculated in accordance with the procedures of Sect.6.6.3.5.

This simplified approach is not always appropriate. If non-symmetricalconnection configurations are used, such as a haunch on only the bottom side ofthe beam, this can result in an uneven distribution of stiffness between the twocolumn segments, and premature yielding of the column, either above, or below,the beam-column connection. Also, it was determined that for connectionconfigurations in which the panel zone depth represents a significant fraction ofthe total column height, such as can occur in some haunched and side-platedconnections, the definition of ΣMc contained in the initial printing of theGuidelines could lead to excessive conservatism in determining whether or not astrong column - weak beam condition exists in a structure. Consequently, InterimGuidelines Advisory No. 1 adopted the current definition of ΣMc for use in thisevaluation. This definition requires that the moments in the column, at the topand bottom of the panel zone be determined for the condition when a plastichinge has formed at all beams in the connection. Figure 6.6.6.3.6-1 illustrates amethod for determining this quantity. In such cases, When evaluation indicatesthat a strong column - weak beam condition does not exist, a plastic analysisshould be considered to determine if an undesirable story mechanism is likely toform in the building.

(L-L’)/2

d ph t

h b

Mpr

Vp

Vp

Mpr

Vc

Vc+Vf

Mct

Mcb

assumed point of zero moment

Note:The quantities Mpr, Vp, L, and L’ areas previously identified. Vf is the incremental shear distributedto the column at the floor level.Other quantities are as shown.

Vf

( )[ ] ( )

( )

VM V L L V h d

h d h

M V h

M V V h

M M M

cpr p f b p

b p t

ct c t

cb c f b

c ct cb

=+ − − +

+ +== +

= +

∑

∑

' ) / /2 2

Figure 6.6.6.3.6-1 Calculation of Column Moment for Strong Column Evaluation



6-23

6.6.6.3.7 Check Column Panel Zone

The adequacy of the shear strength of the column panel zone should be checked. For thispurpose, the term 0.8ΣMf should be substituted for the term 0.8ΣMs in UBC-94 Section2211.7.2.1 {0.9ΣφbMp in NEHRP-91 Section 10.10.3.1} repeated below for convenience ofreference. Mf is the calculated moment at the face of the column, when the beam mechanismforms, calculated as indicated in Section 6.6.6.3.5, above. In addition, it is recommended not touse the alternative design criteria indicated in UBC-94 Section 2211.7.2.1 (NEHRP-91 Sect.10.10.3.1), permitting panel zone shear strength to be proportioned for the shear induced bybending moments from gravity loads plus 1.85 times the prescribed seismic forces. Forconvenience of reference, UBC-94 Section 2211.7.2.1 is reproduced below, edited, to indicate therecommended application:

2211.7.2.1 Strength (edited). The panel zone of the joint shall be capable of resisting theshear induced by beam bending moments due to gravity loads plus 1.85 times theprescribed seismic forces, but the shear strength need not exceed that required to develop0.8ΣMs 0.8ΣMf of the girders framing into the column flanges at the joint. The joint panelzone shear strength may be obtained from the following formula:

V 0.55F d t 13b td d ty c

c c f2

b c

= +

(11-1)

where: bc = width of column flangedb = the depth of the beam (including haunches or cover plates)dc = the depth of the columnt = the total thickness of the panel zone including doubler platestcf = the thickness of the column flange

Commentary: The effect of panel zone shear yielding on connection behavior isnot well understood. In the past, panel zone shear yielding has been viewed as abenign mechanism that permits overall frame ductility demands to beaccommodated while minimizing the extent of inelastic behavior required of thebeam and beam flange to column flange joint. The criteria permitting panel zoneshear strength to be proportioned for the shears resulting from moments due togravity loads plus 1.85 times the design seismic forces was adopted by the codespecifically to encourage designs with weak panel zones. However, during recenttesting of large scale connection assemblies with weak panel zones, it has beennoted that in order to accommodate the large shear deformations that occur inthe panel zone, extreme “kinking” deformations were induced into the columnflanges at the beam flange to column flange welded joint. While this did not leadto premature joint failure in all cases, it is believed to have contributed to suchpremature failures in at least some of the specimens. The recommendations ofthis section are intended to result in stronger panel zones than previously



6-24

permitted by the code, thereby avoiding potential failures due to this kinkingaction on the column flanges.

6.6.7 Modification Details


6.6.7.1 Haunch at Bottom Flange

Figure 6-166.6.7.1-1 illustrates the basic configuration for a connection modificationconsisting of the addition of a welded haunch at the bottom beam flange. Several tests of such amodification were conducted by Uang under the SAC phase I project (Uang, 1995). Followingthat work, additional research on the feasibility of improving connection performance with weldedhaunches was conducted under a project that was jointly sponsored by NIST and AISC (NIST,1998). As indicated in the report of that work, the haunched modification improves connectionperformance by altering the basic behavior of the connection. In essence, the haunch creates aprop type support, beneath the beam bottom flange. This both reduces the effective flexuralstresses in the beam at the face of the support, and also greatly reduces the shear that must betransmitted to the column through the beam. Laboratory tests indicate this modification can beeffective when the existing low-toughness welds between the beam bottom flange and column areleft in place, however, more reliable performance is obtained when the top welds are modified. Acomplete procedure for the design of this modification may be found in NIST, 1998. twoalternative configurations of this detail that have been tested (Uang - 1995). The basic concept isto reinforce the connection with the provision of a triangular haunch at the bottom flange. Theintended behavior of both configurations is to shift the plastic hinge from the face of the columnand to reduce the demand on the CJP weld by increasing the effective depth of the section. In onetest, shown on the left of Figure 6-16, the joint between the girder bottom flange and column wascut free, to simulate a condition which might occur if the bottom joint had been damaged, but notrepaired. In a second tested configuration, the bottom flange joint was repaired and the top flangewas replaced with a locally thickened plate, similar to the detail shown in Figure 6-9.

Design Issues: This approach developed acceptable levels of plastic rotation. Acceptable levelsof connection strength were also maintained during large inelastic deformations of the plastichinge. This approach does not require that the top flange be modified, or slab disturbed, unlessother conditions require repair of the top flange, as in the detail on the left of Figure 6-16. Thebottom flange is generally far more accessible than the top flange because a slab does not haveto be removed. In addition, the haunch can be installed at perimeter frames without removal ofthe exterior building cladding. There did not appear to be any appreciable degradation inperformance when the bottom beam flange was not re-welded to the face of the column.Eliminating this additional welding should help reduce the cost of the repair.

Performance is dependent on properly executed complete joint penetration welds at the columnface and at the attachment of the haunch to the girder bottom flange. The joint can be subject tothrough-thickness flaws in the column flange; however, this connection may not be as sensitive



6-25

to this potential problem because of the significant increase in the effective depth of the beamsection which can be achieved. Welding of the bottom haunch requires overhead welding. Theskewed groove welds of the haunch flanges to the girder and column flanges may be difficult toexecute.

Experimental Results: This approach developed excellent levels of plastic rotation. In Specimen1, the bottom flange CJP weld was damaged in a prior test but was not repaired: only the bottomhaunch was added. During the test of specimen 1, a slowly growing crack developed at theunderside of the top flange-web intersection, perhaps exacerbated by significant local bucklingof the top flange. Some of the buckling may be attributed to lateral torsional buckling thatoccurred because the bottom flange was not restrained by a CJP weld. A significant portion ofthe flexural strength was lost during the cycles of large plastic rotation. In the second specimen,the bottom girder flange weld was intact during the haunch testing, and its performance wassignificantly improved compared with the first specimen. The test was stopped when significantlocal buckling led to a slowly growing crack at the beam flange and web intersection. At thistime, it appears that repairing damaged bottom flange welds in this configuration can producebetter performance. Acceptable levels of flexural strength were maintained during largeinelastic deformations of the plastic hinge for both specimens. As reported in NIST, 1998, a totalof 9 beam-column connection tests incorporating bottom haunch modifications of pre-Northridge connections have been tested in the laboratory, including two dynamic tests. Most ofthe connection assemblies tested resisted in excess of 0.02 radians of imposed plastic rotation.However, for those specimens in which the existing low-toughness weld was left in place at thebeam top flange, without modification, connection behavior was generally limited by fracturesgenerating at these welds at relatively low plastic rotations. It may be expected that enhancedperformance can be obtained by replacing or reinforcing these welds as part of the modification.

Figure 6-166.6.7.1-1 - Bottom Haunch Connection Modification



6-26

Quantitative Results: No. of specimens tested: 29Girder Size: W30 x 99Column Size: W14 x 176Plastic Rotation achieved-

Specimen 1 UCSD-1R: 0.04 radian (w/o bottom flange weld)Specimen 2 UCSD-3R:0.05 radian (with bottom flange weld)

Specimen UCSD-4R: 0.014 radian (dynamic- limited by test setup)Speciemn UCSD-5R: 0.015 radian (dynamic- limited by test setup)

Girder Size: W36x150Column Size: W14x257Plastic Rotation achieved -

Specimen UCB-RN2: 0.014 radian (no modification of top weld)Specimen UTA-1R: 0.019 radian (partial modification of top weld)Specimen UTA-1RB: 0.028 radian (modified top weld)

Girder Size: W36x150Column Size: W14x455Plastic Rotation achieved-

Specment UTA-NSF4: 0.015 radian (no modification of top weld)


Specimen SFCCC-8: 0.035 radian (cover plated top flange)

6.6.7.2 Top and Bottom Haunch


6.6.7.3 Cover Plate Sections

Figure 6.6.7.3-1 Figure 6-18 illustrates the basic configurations of cover plate connections.The assumption behind the cover plate is that it reduces the applied stress demand on the weld atthe column flange and shifts the plastic hinge away from the column face. Only the connectionwith cover plates on the top of the top flange has been tested. There are no quantitative resultsfor cover plates on the bottom side of the top flange, such as might be used in repair. It is likelythat thicker plates would be required where the plates are installed on the underside of the topflange. The implications of this deviation from the tested configuration should be considered.



6-27

Top &Bottom

Near and Far Sides

Top &Bottom

d

d/2, typical

Figure 6-18 Figure 6.6.7.3-1 - Cover Plate Connection Modification

Design Issues: Following the Northridge earthquake, the University of Texas at Austinconducted a program of research, under private funding, to develop a modified connectionconfiguration for a specific project. Following a series of unsuccessful tests on various types ofconnections,Approximately eight connections similar to that shown in Figure 6-18Figure 6.7.3-1have been were tested (Engelhardt & Sabol - 1994), and have demonstrated the ability toachieve acceptable levels of plastic rotation provided that the beam flange to column flangewelding wasis correctly executed and through-thickness problems in the column flange wereareavoided. Due to the significant publicity that followed these successful tests, as well as theeconomy of these connections relative to some other alternatives, cover plated connectionsquickly became the predominant configuration used in the design of new buildings. As a result,a number of qualification tests have now been performed on different variations of cover platedconnections, covering a wide range of member sizes ranging from W16 to W36 beams, as part ofthe design process for individual building projects. The results of these tests have beensomewhat mixed, with a significant number of failures reported. Although this connection typeappears to be significantly more reliable than the typical pre-Northridge connection, it should beexpected that some connections in buildings incorporating this detail may still be subjected toearthquake initiated fracture damage. Designers should consider using alternative connectiontypes, unless highly redundant framing systems are employed.

The option with the top flange cover plate located on top of the flange can be used onperimeter frames where access to the outer side of the beam is restricted by existing buildingcladding. The option with the cover plate for the top flange located beneath the flange can beinstalled without requiring modification of the slab. In the figures shown, the bottom cover plateis rectangular, and sized slightly wider than the beam flange to allow downhand fillet welding ofthe joint between the two plates. Some configurations using triangular plates at the bottomflange, similar to the top flange have also been tested.



6-28

Designers using this detail are cautioned to be mindful of not making cover plates so thickthat excessively large welds of the beam flange combination to column flange result. As thecover plates increase in size, the weld size must also increase. Larger welds invariably result ingreater shrinkage stresses and increased potential for cracking prior to actual loading. Inaddition, larger welds will lead to larger heat affected zones in the column flange, a potentiallybrittle area.

Performance is dependent on properly executed girder flange welds. The joint can be subjectto through-thickness failures in the column flange. Access to the top of the top flange requiresdemolition of the existing slab. Access to the bottom of the top flange requires overhead weldingand may be problematic for perimeter frames. Costs are greater than those associated withapproaches that concentrate modifications on the bottom flange

Experimental Results: Six of eight connections tested by the University of Texas at Austin wereable to achieve plastic rotations of at least 0.025 radians, or better. These tests were performedusing heavy column sections which forced nearly all of the plastic deformation into the beamplastic hinge; very little column panel zone deformation occurred. Strength loss at the extremelevels of plastic rotation did not reduce the flexural capacity to less than the plastic momentcapacity of the section based on minimum specified yield strength. One specimen achievedplastic rotations of 0.015 radians when a brittle fracture of the CJP weld (type W2 failure)occurred. This may partially be the result of a weld that was not executed in conformance withthe specified welding procedure specification. The second unsuccessful test specimen achievedplastic rotations of 0.005 radian when a section of the column flange pulled out (type C2failure). The successful tests were terminated either when twisting of the specimen threatened todamage the test setup or the maximum stroke of the loading ram was achieved. Since thecompletion of that testing, a number of additional tests have been performed. Data for 18 tests,including those performed by Engelhardt and referenced above, are in the public domain. Atleast 12 other tests have been performed on behalf of private parties, however, the data fromthese tests are not available. Some of those tests exhibited premature fractures.

Quantitative Results: No. of specimens tested: 18Girder Size: W21 x 68 to W36 x 150Column Size: W12 x 106 to W14 x 455, and 426Plastic Rotation achieved-

6 13 Specimens : >.025 radian to 0.05 radian13 Specimens: 0.005 < θp < 0.0250.015 radian (W2 failure)12 Specimens: 0.005 radian (C2 failure)

6.6.7.4 Upstanding Ribs




6-29

6.6.7.5 Side-Plate Connections


6.6.7.6 Bolted Brackets

Heavy bolted brackets, incorporating high strength bolts, may be added to existing weldedconnections to provide an alternative load path for transfer of stress between the beams andcolumns. To be compatible with existing welded connections, the brackets must have sufficientstrength and rigidity to transfer beam stresses with negligible deformation. Pre-tensioning of thebolts or threaded rods attaching the brackets to the column flanges and use of welds or slip-critical connections between the brackets and beam flanges can help to minimize deformationunder load. Reinforcement of the column flanges may be required to prevent local yielding andexcessive deformation of these elements. Two alternative configurations, which may be usedeither to repair an existing damaged, welded connection or to reinforce an existing undamagedconnection are illustrated in Figure 6.6.7.6-1. The developer of these connections offers thebrackets in the form of proprietary steel castings. Several tests of these alternative connectionshave been performed on specimens with beams ranging in size from W16 to W36 sections andwith large plastic rotations successfully achieved. Under a project jointly funded by NIST andAISC, the use of a single bracket at the bottom flange of the beam was investigated. It wasdetermined that significant improvement in connection behavior could be obtained by placing abracket at the bottom beam flange and by replacing existing low-toughness welds at the top flangewith tougher material. NIST, 1998 provides a recommended design procedure for suchconnection modifications.

Design Issues: The concept of bolted bracket connections is similar to that of the riveted “windconnections” commonly installed in steel frame buildings in the early twentieth century. Theprimary difference is that the riveted wind connections were typically limited in strength eitherby flexural yielding of outstanding flanges of the brackets, or shear and tension on the rivets,rather than by flexural hinging of the connected framing. Since the old-style wind connectionscould not typically develop the flexural strength of the girders and also could be quite flexible,they would be classified either as partial strength or partially restrained connections. Followingthe Northridge earthquake, the concept of designing such connections with high strength boltsand heavy plates, to behave as fully restrained connections, was developed and tested by aprivate party who has applied for patents on the concept of using steel castings for this purpose.

Bolted bracket connections can be installed in an existing building without field welding. Sincethis reduces the risk of construction-induced fire, brackets may be installed with somewhat lessdemolition of existing architectural features than with welded connections. In addition, thequality assurance issues related to field welding are eliminated. However, the fabrication of thebrackets themselves does require quality assurance. Quality assurance is also required foroperations related to the drilling of bolt holes for installation of bolts, surface preparation offaying surfaces and for installation and tensioning of the bolts themselves.



6-30

PipePlate

Bolts

High tensilethreaded rod

Bolts

Steelcasting

WARNING: The information presented in this figure is PROPRIETARY. US and ForeignPatents have been applied for. Use of this information is strictly prohibited except as authorizedin writing by the developer. Violators shall be prosecuted in accordance with US and ForeignPatent Intellectual Property Laws.

Figure 6.6.7.6-1 Bolted Bracket Modification

Bolted brackets can be used to repair damaged connections. If damage is limited to the beamflange to column flange welds, the damaged welds should be dressed out by grinding. Anyexisting fractures in base metal should be repaired as indicated in Section 6.3, in order torestore the strength of the damaged members and also to prevent growth of the fractures underapplied stress. Since repairs to base metal typically require cutting and welding, this reducessomewhat the advantages cited above, with regard to avoidance of field welding.

Experimental Results: A series of tests on several different configurations of proprietary heavybolted bracket connections have been performed at Lehigh University (Ksai & Bleiman, 1996) toqualify these connections for use in repair and modification applications. To test repairapplications, brackets were placed only on the bottom beam flange to simulate installations on aconnection where the bottom flange weld in the original connection had failed. In thesespecimens, bottom flange welds were not installed, to approximate the condition of a fullyfractured weld. The top flange welds of these specimens were made with electrodes rated fornotch toughness, to preclude premature failure of the specimens at the top flange. Forspecimens in which brackets were placed at both the top and bottom beam flanges, both weldswere omitted. Acceptable plastic rotations were achieved for each of the specimens tested. Notesting has yet been performed to determine the effectiveness of bolted brackets when applied toan existing undamaged connection with full penetration beam flange to column flange welds withlow toughness or significant defects or discontinuities.

Quantitative Results: No. of specimens tested: 8Girder Size: W16x40 and W36x150Column Size: W12x65 and W14x425Plastic Rotation achieved - 0.05 radians - 0.07 radians



6-31



New Construction

7-1

7. NEW CONSTRUCTION

7.1 Scope

This Chapter presents interim design guidelines for new welded steel moment frames(WSMFs) intended to resist seismic demands through inelastic behavior. The criteria apply to allSMRF structures designed for earthquake resistance and those IMRF and OMRF structureslocated in Uniform Building Code (UBC) Seismic Zones 3 and 4 {National Earthquake HazardsReduction Program (NEHRP) Map Areas 6 and 7} or assigned to 1997 NEHRP Seismic DesignCategories D, E, or F. Light, single-story buildings, the design of which is governed by wind,need not consider these Interim Guidelines. Frames with bolted connections, either fullyrestrained (type FR) or partially restrained (type PR), are beyond the scope of this document. However, the acceptance criteria for connections may be applied to type FR bolted connections aswell.

Commentary: Observation of damage experienced by WSMF buildings in theNorthridge Earthquake and subsequent laboratory testing of large scale beam-column assemblies has demonstrated that the standard details for WSMFconnections commonly used in the past are not capable of providing reliableservice in the post-elastic range. Therefore, structures which are expected toexperience significant post-elastic demands from design earthquakes, or forwhich highly reliable seismic performance is desired, should be designed usingthe Interim Guidelines presented herein.

In order to determine if a structure will experience significant inelasticbehavior in a design earthquake, it is necessary to perform strength checks of theframe components for the combination of dead and live loads expected to bepresent, together with the full earthquake load. Except for structures with specialperformance goals, or structures located within the near field (within 10kilometers) of known active earthquake faults, the full earthquake load may betaken as the minimum design earthquake load specified in the building code, butcalculated using a lateral force reduction coefficient (Rw or R) of unity. If allcomponents of the structure and its connections have adequate strength to resistthese loads, or nearly so, then the structure may be considered to be able to resistthe design earthquake, elastically.

Design of frames to remain elastic under unreduced (Rw {R} taken as unity)earthquake forces may not be an overly oppressive requirement, particularly inmore moderate seismic zones. Most frame designs are currently controlled bydrift considerations and have substantially more strength than the minimumspecified for design by the building code. As part of the SAC Phase 1 research, anumber of modern frame buildings designed with large lateral force reductioncoefficients (Rw = 12, {R = 8}) were evaluated for unreduced forces calculated


New Construction

7-2

using the standard building code spectra. It was determined that despite thenominally large lateral force reduction coefficients used in the original design,the maximum computed demands from the dynamic analyses were only on theorder of 2 to 3 times those which would cause yielding of the real structures(Krawinkler, et. al. - 1995; Uang, et. al. - 1995; Engelhardt, et. al. - 1995, Hart,et. al. - 1995; Kariotis and Eimani - 1995). Therefore, it is not unreasonable toexpect that OMRF structures (nominally designed with a lateral force reductioncoefficient Rw = 6 {R = 4.5}) could resist the design earthquakes with near elasticbehavior. Regardless of these considerations, better seismic performance can beexpected by designing structures with greater ductility rather than less, andengineers are not encouraged to design structures for elastic behavior usingbrittle or unreliable details..

For structures designed to meet special performance goals, and buildingslocated within the near field of major active faults, full earthquake loadscalculated in accordance with the above procedure may not be adequate. Forsuch structures, the full earthquake load should be determined using a sitespecific ground motion characterization and a suitable analysis procedure. Recent research (Heaton, et. al. - 1995) suggests that the elastic responsespectrum technique, typically used for determining seismic forces for structuraldesign, may not provide an adequate indication of the true earthquake demandsproduced by the large impulsive ground motions common in the near field oflarge earthquake events. Further, this research indicates that frame structures,subjected to such impulsive ground motions can experience very large drifts, andpotential collapse. In an attempt to address this, both the 1997 edition of theUniform Building Code and the 1997 edition of the NEHRP Provisions specifydesign ground motions for structures located close to major active faults that aresubstantially more severe than those contained in earlier codes. While the moresevere ground motion criteria contained in these newer provisions are probablyadequate for the design of most structures, analytical studies conducted by SACconfirm that even structures designed to these criteria can experience very largedrift demands, and potentially collapse, if the dynamic characteristics of theimpulsive loading and those of the structure are matched. Direct nonlinear timehistory analysis, using an appropriate ground motion representation would beone method of more accurately determining the demands on structures located inthe near field. Additional research on these effects is required.

As an alternative to use of the criteria contained in these Interim Guidelines,OMRF structures in zones of high seismicity (UBC seismic zones 3 and 4 andNEHRP map areas 6 and 7) and OMRF structures assigned to 1997 NEHRPSeismic Design Categories D, E or F, may be designed for the connections toremain elastic (Rw or R taken as 1.0) while the beams and columns are designedusing the standard lateral force reduction coefficients specified by the building


New Construction

7-3

code. Although this is an acceptable approach, it may result in much largerconnections than would be obtained by following these Interim Guidelines.

The use of partially restrained connections may be an attractive andeconomical alternative to the design of frames with fully restrained connections. However, the design of frames with partially restrained connections is beyond thescope of this document. The AISC is currently working on development ofpractical design guidelines for frames with partially restrained connections.

7.2 General - Welded Steel Frame Design Criteria

7.2.1 Criteria

Welded Steel Moment Frame (WSMF) systems should, as a minimum, be designed for theprovisions of the prevailing building code and these Interim Guidelines. Special Moment-Resisting Frames (SMRF)s and Ordinary Moment-Resisting Frames (OMRF)s with FRconnections, should additionally be designed in accordance with either the 1997 edition of theAISC Seismic Provisions for Structural Steel Buildings (AISC, 1997) or the emergency codechange to the 1994 UBC {NEHRP-1994}, restated as follows:

2211.7.1.1. Required Strength {NEHRP-1994 Section 5.2, revision to Ref. 8.2c of Ref. 5.3}

The girder-to-column connections shall be adequate to develop the lesser of the following:

1. The strength of the girder in flexure.

2. The moment corresponding to development of the panel zone shear strength as determined by Formula (11-1).

2211.7.1.3-2 Connection Strength

Connection configurations utilizing welds and high strength bolts shall demonstrate, by approved cyclic test results orcalculation, the ability to sustain inelastic rotations and to develop the strength criteria in Section 2211.7.1.1considering the effects of steel overstrength and strain hardening.

Commentary: At the time the Interim Guidelines were first published, they werebased on the 1994 edition of the Uniform Building Code and the 1994 edition ofthe NEHRP Provisions. In the time since that initial publication, more recenteditions of both documents have been published, and codes based on thesedocuments have been adopted by some jurisdictions. In addition, the AmericanInstitute of Steel Construction has adopted a major revision to its SeismicProvisions for Structural Steel Buildings (AISC Seismic Provisions), largelyincorporating, with some modification, the recommendations contained in theInterim Guidelines. This updated edition of the AISC Seismic Provisions hasbeen incorporated by reference into the 1997 edition of the NEHRP Provisionsand has also been adopted by some jurisdictions as an amendment to the modelbuilding codes. SMRF and OMRF systems that are designed to comply with therequirements of the 1997 AISC Seismic Provisions may be deemed to comply withthe intent of these Interim Guidelines. Where reference is made herein to the


New Construction

7-4

requirements of the 1994 Uniform Building Code or 1994 NERHP Provisions, theparallel provisions of the 1997 editions may be used instead, and should be usedin those jurisdictions that have adopted codes based on these updated standards.

The 1997 edition NEHRP Provisions and AISC Seismic Provisions introducea new structural system termed an Intermediate Moment Resisting Frame (IMRF). Provisions for IMRF structures include somewhat more restrictive detailing anddesign requirements than those for OMRF structures, and less than those forSMRF structures. The intent was to provide a system that would be moreeconomical than SMRF structures yet have better inelastic response capabilitythan OMRF structures. The SAC project is currently conducting research todetermine if the provisions for the new IMRF system are adequate, but has notdeveloped a position on this at this time.

At this time, no recommendations are made to change the minimum lateralforces, drift limitations or strength calculations which determine member sizingand overall performance of moment frame systems, except as recommended inSections 7.2.2, 7.2.3 and 7.2.4. The design of joints and connections is discussedin Section 7.3. The UBC permits OMRF structures with FR connections,designed for 3/8Rw times the earthquake forces otherwise required, to bedesigned without conforming to Section 2211.7.1. However, this is notrecommended.

7.2.2 Strength and Stiffness

7.2.2.1 Strength

When these Interim Guidelines require determination of the strength of a framing element orcomponent, this shall be calculated in accordance with the criteria contained in UBC-94, Section2211.4.2 {NEHRP-91 Section 10.2, except that the factor φ should be taken as 1.0}, restated asfollows:

2211.4.1 Member strength. Where this section requires that the strength of the member bedeveloped, the following shall be used:

Flexure Ms = Z Fy

Shear Vs = 0.55 Fy d tAxial compression Psc = 1.7 Fa AAxial tension Pst = Fy AConnectors Full Penetration welds Fy A Partial Penetration welds 1.7 allowable (see commentary) Bolts and fillet welds 1.7 allowable


New Construction

7-5

Commentary: Partial penetration welds are not recommended for tensionapplications in critical connections resisting seismic-induced stresses. Thegeometry of partial penetration welds creates a notch-like condition that caninitiate brittle fracture under conditions of high tensile strain.

Many WSMF structures are constructed with concrete floor slabs that areprovided with positive shear attachment between the slab and the top flanges ofthe girders of the moment-resisting frames. Although not generally accounted forin the design of the frames, the resulting composite action can increase theeffective strength of the girder significantly, particularly at sections wherecurvature of the girder places the top flange into compression. Although thiseffect is directly accounted for in the design of composite systems, it is typicallyneglected in the design of systems classified as moment resisting steel frames. The increased girder flexural strength caused by this composite action can resultin a number of effects including the unintentional creation of weak column -strong beam and weak panel zone conditions. In addition, this composite effecthas the potential to reduce the effectiveness of reduced section or “dog-bone”type connection assemblies. Unfortunately, very little laboratory testing of largescale connection assemblies with slabs in place has been performed to date andas a result, these effects are not well quantified. In keeping with typicalcontemporary design practice, the design formulae provided in these Guidelinesneglect the strengthening effects of composite action. Designers should, however,be alert to the fact that these composite effects do exist.

7.2.2.2 Stiffness

Calculation of frame stiffness for the purpose of determining interstory drift under theinfluence of the design lateral forces should be based on the properties of the bare steel frame,neglecting the effects of composite action with floor slabs. The stiffening effects of connectionreinforcements (e.g.: haunches, side plates, etc.) may be considered in the calculation of overallframe stiffness and drift demands. When reduced beam section connections are utilized, thereduction in overall frame stiffness, due to local reductions in girder cross section, should beconsidered.

Commentary: For design purposes, frame stiffness is typically calculatedconsidering only the behavior of the bare frame, neglecting the stiffening effectsof slabs, gravity framing, and architectural elements. The resulting calculation ofbuilding stiffness and period typically underestimates the actual properties,substantially. Although this approach can result in unconservative estimates ofdesign force levels, it typically produces conservative estimates of interstory driftdemands. Since the design of most moment-resisting frames are controlled byconsiderations of drift, this approach is considered preferable to methods thatwould have the potential to over-estimate building stiffness. Also, many of the


New Construction

7-6

elements that provide additional stiffness may be subject to rapid degradationunder severe cyclic lateral deformation, so that the bare frame stiffness mayprovide a reasonable estimate of the effective stiffness under long durationground shaking response.

Notwithstanding the above, designers should be alert to the fact thatunintentional stiffness introduced by walls and other non-structural elements cansignificantly alter the behavior of the structure in response to ground shaking. Ofparticular concern, if these elements are not uniformly distributed throughout thestructure, or isolated from its response, they can cause soft stories and torsionalirregularities, conditions known to result in poor behavior.

7.2.3 Configuration

Frames should be proportioned so that the required plastic deformation of the frame can maybe accommodated through the development of plastic hinges at pre-determined locations withinthe girder spans, as indicated in Figure 7-1Figure 7.2.3-1. Beam-column connections should bedesigned with sufficient strength (through the use of cover plates, haunches, side plates, etc.) toforce development of the plastic hinge away from the column face. This condition may also beattained through local weakening of the beam section at the desired location for plastic hingeformation.

Plastic Hinges

Deformed frame shapeUndeformedframe

L’

L

h

drift angle - θ

Figure 7-1 Figure 7.2.3-1 - Desired Plastic Frame Behavior

Commentary: Nonlinear deformation of frame structures is typicallyaccommodated through the development of inelastic flexural or shear strainswithin discrete regions of the structure. At large inelastic strains these regionscan develop into plastic hinges, which can accommodate significant concentratedrotations at constant (or nearly constant) load through yielding at tensile andcompressive fibers and by buckling at compressive fibers. If a sufficient numberof plastic hinges develop in a frame, a mechanism is formed and the frame can


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deform laterally in a plastic manner. This behavior is accompanied by significantenergy dissipation, particularly if a number of members are involved in theplastic behavior, as well as substantial local damage to the highly strainedelements. The formation of hinges in columns, as opposed to beams, isundesirable, as this results in the formation of weak story mechanisms withrelatively few elements participating, so called “story mechanisms” andconsequently little energy dissipation occurring. In addition, such mechanismsalso result in local damage to critical gravity load bearing elements.

The prescriptive connection contained in the UBC and NEHRP RecommendedProvisions prior to the Northridge Earthquake was based on the assumeddevelopment of plastic hinge zones within the beams at adjacent to the face of thecolumn, or within the column panel zone itself. If the plastic hinge develops inthe column panel zone, the resulting column deformation results in very largesecondary stresses on the beam flange to column flange joint, a condition whichcan contribute to brittle failure. If the plastic hinge forms in the beam, at the faceof the column, this can result in very large through-thickness strain demands onthe column flange material and large inelastic strain demands on the weld metaland surrounding heat affected zones. These conditions can also lead to brittlejoint failure. Although ongoing research may reveal conditions of materialproperties, design and detailing configurations that permit connections withyielding occurring at the column face to perform reliably, for the present it isrecommended In order to achieve more reliable performance, it is recommendedthat the connection of the beam to the column be configured to force the inelasticaction (plastic hinge) away from the column face. This can be done either bylocal reinforcement of the connection, or locally reducing the cross section of thebeam at a distance away from the connection. Plastic hinges in steel beams havefinite length, typically on the order of half the beam depth. Therefore, thelocation for the plastic hinge should be shifted at least that distance away fromthe face of the column. When this is done through reinforcement of theconnection, the flexural demands on the columns, for a given beam size, areincreased. Care must be taken to assure that weak column conditions are notinadvertently created by local strengthening of the connections.

It should be noted that some professionals and researchers believe thatconfigurations which permit plastic hinging to occur adjacent to the column facemay still provide reliable service under some conditions. These conditions mayinclude limitations on the size of the connected sections, the use of base and weldmetals with adequate notch toughness, joint detailing that minimizes notch effects,and appropriate control of the relative strength of the beam and columnmaterials. Sufficient research has not been performed to date either to confirmthese suggestions or define the conditions in which they are valid. Researchhowever does indicate that reliable performance can be attained if the plastic


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hinge is shifted away from the column face, as suggested above. Consequently,these Interim Guidelines make a general recommendation that this approach betaken. Additional research should be performed to determine the acceptability ofother approaches.

It should also be noted that reinforced connection (or reduced beam section)configurations of the type described above, while believed to be effective inpreventing brittle connection fractures, will not prevent structural damage fromoccurring. Brittle connection fractures are undesirable because they result in asubstantial reduction in the lateral-force-resisting strength of the structure which,in extreme cases, can result in instability and collapse. Connections configuredas described in these Interim Guidelines should experience many fewer suchbrittle fractures than unmodified connections. However, the formation of aplastic hinge within the span of a beam is not a completely benign event. Beamswhich have formed such hinges may, if plastic rotations are large, exhibitsignificantlarge buckling and yielding deformation, damage which typically mustbe repaired. The cost of such repairs could be comparable to the costs incurredin repairing fracture damage experienced in the Northridge Earthquake. Theprimary difference is that life safety protection will be significantly enhanced andmost structures that have experienced such plastic deformation damage shouldcontinue to be safe for occupancy while repairs are made.

If the types of damage described above are unacceptable for a given building,then alternative structural systems should be considered that will reduce theplastic deformation demands on the structure during a strong earthquake. Appropriate methods of achieving such goals include the installation ofsupplemental braced frames, energy dissipation systems, base isolation systemsand similar structural systems. Framing systems incorporating partiallyrestrained connections may also be quite effective in resisting large earthquakeinduced deformation with limited damage.

It is important to recognize that in frames with relatively short bays, theflexural hinging indicated in Figure 7.2.3-1 may not be able to form. If theeffective flexural length (L’ in the figure) of beams in a frame becomes too short,then the beams or girders will yield in shear before zones of flexural plasticitycan form, resulting in an inelastic behavior that is more like that of aneccentrically braced frame than that of a moment frame. This behavior mayinadvertently occur in frames in which relatively large strengthened connections,such as haunches, cover plates or side plates have been used on beams withrelatively short spans. This behavior is illustrated in Figure 7.2.3-2.

The guidelines contained in this section are intended to address the design offlexurally dominated moment resisting frames. When utilizing these guidelines, itis important to confirm that the configuration of the structure is such that the


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presumed flexural hinging can actually occur. It is possible that shear yielding offrame beams, such as that schematically illustrated in Figure 7.2.3-2 may be adesirable behavior mode. However, to date, there has not been enough researchconducted into the behavior of such frames to develop recommended designguidelines. Designers wishing to utilize such configurations should refer to thecode requirements for eccentrically braced frames. Particular care should betaken to brace the shear link of such beams against lateral-torsional buckling andalso to adequately stiffen the webs to avoid local buckling following shearplastification.

Shear Link

Shear Link

Figure 7.2.3-2 Shear Yielding Dominated Behavior of Short Bay Frames

7.2.4 Plastic Rotation Capacity

The plastic rotation capacity of tested connection assemblies should reflect realistic estimatesof the total (elastic and plastic) drift likely to be induced in the frame by earthquake groundshaking, and the geometric configuration of the frame. For frames of typical configuration, andfor ground shaking of the levels anticipated by the building code, a minimum plastic rotationcapacity of 0.03 radian is recommended. As used in these Guidelines, plastic rotation is definedas the plastic chord rotation angle. The plastic chord rotation angle is calculated using the rotatedcoordinate system shown in Fig. 7.2.4-1 as the plastic deflection of the beam or girder, at its pointof inflection (usually the mid-span,) ∆CI, divided by the distance between this mid-span point andthe centerline of the panel zone of the beam column connection, LCL. This convention isillustrated in Figure 7.2.4-1.


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cL

cL

Beam span center line

∆CL

LCL

θ pCL

CLL= ∆

Plastic hinge

lh

Figure 7.2.4-1 Calculation of Plastic Rotation Angle

It is important to note that this definition of plastic rotation is somewhat different than theplastic rotation that would actually occur within a discrete plastic hinge in a frame model similarto that shown in Figure 7.2.3-1. These two quantities are related to each other, however, and ifone of them is known, the other may be calculated from Eq. 7.2.4-1.When the configuration of aframe is such that the ratio L/L’ is greater than 1.25, the plastic rotation demand should be takenas follows:

( )θ θp phCL h

CL

L lL

= −(7.2.4-1)

where: θp is the plastic chord angle rotation, as used in these Guidelines θph is the plastic rotation, at the location of a discrete hinge LCL is the distance from the center of the beam-column assembly panel zone to the center of the beam span lh is the location of the discrete plastic hinge relative to the center of the beam-column assembly panel zone

( )( )θ = + −0.025 1 L L' L' (7-1)

where: L is the center to center spacing of columns, and L’ is the center to center spacing of plastic hinges in the bay under consideration

The indicated rotation demands may be reduced when positive means, such as the use of baseisolation or energy dissipation devices, are introduced into the design to control the building’sresponse. When such measures are taken, nonlinear dynamic analyses should be performed andthe connection demands taken as 0.005 radians greater than the plastic rotation demandscalculated in the analyses. The nonlinear analyses should conform to the criteria specified inUBC-94 Section 1655 {NEHRP-94 Section 2.6.4.2} for nonlinear dynamic analysis of base


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isolated structures. Ground motion time histories utilized for these nonlinear analyses shouldsatisfy the scaling requirements of UBC-94 Section 1655.4.2 {NEHRP-94 Section 2.6.4.4},except that if the building is not base isolated, the structure period T, calculated in accordancewith UBC-94 Section 1628 {NEHRP-94 Section 2.3.3.1} should be substituted for TI. Whenusing methods of nonlinear analysis to establish the plastic rotation demands on frameconnections, the analysis results should not be scaled by the factor Rw (R) or RWi (Ri), asotherwise permitted by the building code.

Commentary: When the Interim Guidelines were first published, the plasticrotation was defined as that rotation that would occur at a discrete plastic hinge,similar to the definition of θph. in Eq. 7.2.4-1, above. In subsequent testing ofprototype connection assemblies, it was found that it is often very difficult todetermine the value of this rotation parameter from test data, since actual plastichinges do not occur at discrete points in the assembly and because some amountof plasticity also occurs in the panel zone of many assemblies. The plastic chordangle rotation, introduced in this advisory, may more readily be obtained fromtest data and also more closely relates to the drift experienced by a frame duringearthquake response.

This change in the definition of plastic rotation does not result in anysignificant change in the acceptance criteria for beam-column assemblyqualification testing. When the Interim Guidelines were first published, theyrecommended an acceptance criteria given by Eq. 7.2.4-2, below:

θ pL L

L= + −

0 025 1.'

'(7.2.4-2)

For typical beam-column assemblies in which the plastic hinge forms relativelyclose to the face of the column, perhaps within a length of 1/2 the beam depth,this typically resulted in a plastic rotation demand of 0.03 radians, as currentlymeasured.

Traditionally, engineers have calculated demand in moment frames by sizingthe members for strength and drift using code forces (either equivalent static orreduced dynamic forces) and then "developing the strength of the members." Since 1988, "developing the strength" has been accomplished by prescriptivemeans based on a review of testing of moment frame connections to that date. Itwas assumed that the prescribed connections would be strong enough that thebeam or girder would yield (in bending), or the panel zone would yield (in shear)in a nearly perfectly plastic manner producing the plastic rotations necessary todissipate the energy of the earthquake.


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A realistic estimate of the interstory drift demand for most structures and mostearthquakes is on the order of 0.015 to 0.025 times the story height for WSMFstructures designed to code allowable drift limits. In such frames, a portion ofthe drift will be due to elastic deformations of the frame, while the balance mustbe provided by inelastic rotations of the beam plastic hinges, by yielding of thecolumn panel zone, or by a combination of the two.

In the 1994 Northridge Earthquake, many moment-frame connectionsfractured with little evidence of plastic hinging of the beams or yielding of thecolumn panel zones. Testing of moment frame connections both prior to andsubsequent to the earthquake suggests that the standard, pre-Northridge, weldedflange-bolted web connection is unable to reliably provide plastic rotationsbeyond about 0.005 radian for all ranges of beam depths and often fails belowthat level. Since the elastic contribution to drift may approach 0.01 radian, thenecessary inelastic contributions will exceed the capability of the standardconnection in many cases. For frames designed for code forces and for the codedrift, the necessary plastic rotational demand may be expected to be on the orderof 0.02 radian or more and new connection configurations should be developed toaccommodate such rotation without brittle fracture.

The recommended plastic rotation connection demand of 0.03 radians wasselected both to provide a comfortable margin against the demands actuallyexpected in most cases and because in recent testing of connection assemblies,specimens capable of achieving this demand behaved in a ductile manner throughthe formation of plastic hinges.

For a given building design, and known earthquake hazard, it is possible tomore accurately estimate plastic rotation demands on frame connections. Thisrequires the use of nonlinear analysis techniques. Analysis software capable ofperforming such analyses is becoming more available and many design officeswill have the ability to perform such analyses and develop more accurateestimates of inelastic demands for specific building designs. However, whenperforming such analyses, care should be taken to evaluate building response formultiple earthquake time histories, representative of realistic ground motions forsites having similar geologic characteristics and proximity to faults as the actualbuilding site. Relatively minor differences in the ground motion time history usedas input in such an analysis can significantly alter the results. Since there issignificant uncertainty involved in any ground motion estimate, it isrecommended that analysis not be used to justify the design of structures withnon-ductile connections, unless positive measures such as the use of baseisolation or energy dissipation devices are taken to provide reliable behavior ofthe structure.


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It has been pointed out that it is not only the total plastic rotation demand thatis important to connection and frame performance, but also the connectionmechanism (for example - panel zone yielding, girder flange yielding/buckling,etc.) and hysteretic loading history. These are matters for further study in thecontinuing research on connection and joint performance.

7.2.5 Redundancy

The frame system should be designed and arranged to incorporate as many moment-resistingconnections as is reasonable into the moment frame.

Commentary: Early moment frame designs were highly redundant and nearlyevery column was designed to participate in the lateral-force-resisting system. Inan attempt to produce economical designs, recent practice often yieldedproduceddesigns which utilized only a few large columns and beams in a small proportionof the building’s frames for lateral resistance, with the balance of the buildingcolumns designed not considered or designed to participate in lateral resistance. This practice led to the need for large welds at the connections and to reliance ononly a few connections for the lateral stability of the building. The resultinglarge framing elements and connections are believed to have exacerbated thepoor performance of the pre-Northridge connection. Further, if only a fewframing elements are available to resist lateral demands, then failure of only afew connections has the potential to result in a significant loss of earthquake-resisting strength. Together, these effects are not beneficial to buildingperformance.

The importance of redundancy to building performance can not be over-emphasized. Even connections designed and constructed according to theimproved procedures recommended by these Interim Guidelines will have somepotential, albeit greatly reduced, for brittle failures. As the number of individualbeams and columns incorporated into the lateral-force-resisting system isincreased, the consequences of isolated connection failures are significantlyreduceds. Further, as more framing elements are activated in the building’sresponse to earthquake ground motion, the building develops greater potential forenergy absorption and dissipation, and greater ability to limit controlearthquake-induced deformations to acceptable levels.

Incorporation of more of the building framing into the lateral-force-resistingsystem will lead to smaller members and therefore an anticipated increase in the reliability of individual connections. It will almost certainly lead to improvedoverall system reliability. Further, recent studies conducted by designers indicatethat under some conditions, redundant framing systems can be constructed aseconomically as non-redundant systems. In these studies, the additional costsincurred in making a greater number of field-welded moment-resisting


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connections in the more redundant frame were balanced by a reduced totaltonnage of steel in the lateral-force-resisting systems, and sometimes reducedfoundation costs as well.

In order to codify the need for more redundant structural systems, the 1997Uniform Building Code has specifically adopted a reliability coefficient, ρx, tiedto the redundancy of framing present in the building. This coefficient, with valuesvarying from 1.0 for highly redundant structures to 1.5 for non-redundantstructures, is applied to the design earthquake forces, E, in the load combinationequations, and has the effect of requiring more conservative design force levelsfor structures with nonredundant systems. The Building Seismic Safety Council’sProvisions Update Committee has also approved a proposal to include such acoefficient in the1997 NEHRP Provisions also includes such a coefficient. Theformulation of this coefficient and its application are very similar in both the1997 Uniform Building Code and 1997 NEHRP Provisions.

As proposed contained in the 1997 NEHRP Provisions, the reliabilitycoefficient is given by the equation:

xArmax

202 −=ρ (7.25-1)

where:

r xmax = the ratio of the design story shear resisted by the single element

carrying the most shear force in the story to the total story shear, for agiven direction of loading. For moment frames, r xmax is taken as the

maximum of the sum of the shears in any two adjacent columns in amoment frame divided by the story shear. For columns common to twobays with moment resisting connections on opposite sides at the levelunder consideration, 70% of the shear in that column may be used in thecolumn shear summation.

Ax = the floor area in square feet of the diaphragm level immediately above thestory.

The 1997 UBC and NEHRP Provisions also require that structures utilizingmoment resisting frames as the primary lateral force resisting system beproportioned such that they qualify for a maximum value of ρx of 1.25. Structureslocated within a few kilometers of major active faults must be configured so as toqualify for a maximum value of ρx of 1.1.


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The most redundant moment-resisting frame systems are distributed frames inwhich all beam-column connections are detailed to be moment resisting. In thesetypes of structures, half of the moment-resisting connections will be to the minoraxis of the column which will typically result in weak column/strong beamframing. The 1994 UBC requirements limit the portion of the building designlateral forces that can be resisted by relative number of weak column/strong beamconnections in the moment frame system. This limitation was adopted to avoidthe design of frames likely to develop story mechanisms as opposed to concernabout the adequacy of moment-resisting connections to the minor axis ofcolumns. However, the limited research data available on such connectionssuggests that they do not behave well.

There is a divergence of opinion among structural engineers on thedesirability of frames in which all beam-column connections are made moment-resisting, including those of beams framing to the minor axis of columns. Use ofsuch systems as a means of satisfying the redundancy recommendations of theseInterim Guidelines requires careful consideration by the structural engineer. Limited testing in the past has indicated that moment connections made to theminor axis of wide flange columns are subject to the same types of fracturedamage experienced by major axis connections. As of this time, there has notbeen sufficient research to suggest methods of making reliable connections to thecolumn minor axis.

7.2.6 System Performance


7.2.7 Special Systems


7.3 Connection Design & Qualification Procedures - General

7.3.1 Connection Performance Intent

The intent of connection design should be to force the plastic hinge away from the face of thecolumn to a pre-determined location within the beam span. This may be accomplished by localreinforcement of the connection itself (cover plates, haunches, side plates, etc.) or by localreductions of the beam section (drilled holes, trimmed flanges, etc.). All elements of theconnection should have adequate strength to develop the forces resulting from the formation ofthe plastic hinge at the predetermined location, together with forces resulting from gravity loads. Section 7.5.2 outlines a design procedure for reinforced connection designs. Section 7.5.3provides a design procedure for reduced section connections.


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7.3.2 Qualification by Testing


7.3.3 Design by Calculation


7.4 Guidelines for Connection Qualification by Testing

7.4.1 Testing Protocol


7.4.2 Acceptance Criteria

The minimum acceptance criteria for connection qualification for specimens tested inaccordance with these Interim Guidelines should be as follows:

a) The connection should develop beam plastic rotations as indicated in Section7.2.4, for at least one complete cycle.

b) The connection should develop a minimum strength equal to the plastic strength ofthe girder, calculated using minimum specified yield strength Fy, tThroughout theloading history required to achieve the required plastic rotation capacity, asindicated in a), above, the connection should develop a minimum moment at thecolumn face as follows:

i) For strengthened connections, the minimum moment at the column faceshould be equal to the plastic moment of the girder, calculated using theminimum specified yield strength, Fy of the girder. If the load limitingmechanism in the test is buckling of the girder flanges, the engineer, uponconsideration of the effect of strength degradation on the structure, mayconsider a minimum of 80% of the nominal strength as acceptable.

ii) For reduced section connection designs, the minimum moment at thecolumn face should be equal to the moment corresponding to developmentof the nominal plastic moment of the reduced section at the reducedsection, calculated using the minimum specified yield strength, Fy of thegirder, and the plastic section modulus for the reduced section. Themoment at the column face should not be less than 80% of the nominalplastic moment capacity of the unreduced girder section.


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c) The connection should exhibit ductile behavior throughout the loading history. Aspecimen that exhibits a brittle limit state (e.g. complete flange fracture, columncracking, through-thickness failures of the column flange, fractures in weldssubject to tension, shear tab cracking, etc. ) prior to reaching the required plasticrotation should be considered unsuccessful.

d) Throughout the loading history, until the required plastic rotation is achieved, theconnection should be judged capable of supporting dead and live loads required bythe building code. In those specimens where axial load is applied during thetesting, the specimen should be capable of supporting the applied load throughoutthe loading history.

The evaluation of the test specimen’s performance should consistently reflect the relevant limitstates. For example, the maximum reported moment and the moment at the maximum plasticrotation are unlikely to be the same. It would be inappropriate to evaluate the connection usingthe maximum moment and the maximum plastic rotation in a way that implies that they occurredsimultaneously. In a similar fashion, the maximum demand on the connection should beevaluated using the maximum moment, not the moment at the maximum plastic rotation unless thebehavior of the connection indicated that this limit state produced a more critical condition in theconnection.

Commentary: While the testing of all connection geometries and membercombinations in any given building might be desirable, it would not be verypractical nor necessary. Test specimens should replicate, within the limitationsassociated with test specimen simplification, the fabrication and weldingprocedures, connection geometry and member size, and potential modes offailure. If the testing is done in a manner consistent with other testing programs,reasonable comparisons can be made. On the other hand, testing is expensiveand it is difficult to realistically test the beam-column connection using actualboundary conditions and earthquake loading histories and rates.

It was suggested in Interim Recommendation No. 2 by the SEAOC SeismologyCommittee that three tested specimens be the minimum for qualification of aconnection. Further consideration has led to the recognition that while three testsmay be desirable, the actual testing program selected should consider theconditions of the project. Since the purpose of the testing program is to "qualifythe connection", and since it is not practical for a given project to do enough teststo be statistically meaningful considering random factors such as material,welder skills, and other variables, arguments can be made for fewer tests ofidentical specimens, and concentration on testing specimens which represent therange of different properties which may occur in the project. Once a connectionis qualified, that is, once it has been confirmed that the connection can work,monitoring of actual materials and quality control to assure emulation of thetested design becomes most important.


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Because of the cost of testing, use of calculations for interpolation orextrapolation of test results is desirable. How much extrapolation should beaccepted is a difficult decision. As additional testing is done, more informationmay be available on what constitutes "conservative" testing conditions, therebyallowing easier decisions relative to extrapolating tests to actual conditions whichare likely to be less demanding than the tests. For example, it is hypothesizedthat connections of shallower, thinner flanged members are likely to be morereliable than similar connections consisting of deeper, thicker flanged members. Thus, it may be possible to test the largest assemblages of similar details andextrapolate to the smaller member sizes - at least within comparable membergroup families. However, there is evidence to suggest that extrapolation of testresults to assemblies using members of reduced size is not always conservative. In a recent series of tests of cover plated connections, conducted at the Universityof California at San Diego, a connection assembly that produced acceptableresults for one family of beam sizes, W24, did not behave acceptably when thebeam depth was reduced significantly, to W18. In that project, the change inrelative flexibilities of the members and connection elements resulted in a shift inthe basic behavior of the assembly and initiation of a failure mode that was notobserved in the specimens with larger member sizes. In order to minimize thepossibility of such occurrences, when extrapolation of test results is performed, itshould be done with a basic understanding of the behavior of the assembly, andthe likely effects of changes to the assembly configuration on this behavior. Testresults should not be extrapolated to assembly configurations that are expected tobehave differently than the tested configuration. Extrapolation or interpolationof results with differences in welding procedures, details or material properties iseven more difficult.

7.5 Guidelines for Connection Design by Calculation

In conditions where it has been determined that design of connections by calculation issufficient, or when calculations are used for interpolation or extrapolation, the followingguidelines should be used.

7.5.1 Material Strength Properties

In the absence of project specific material property information, the values listed in Table 7-1Table 7.5.1-1 should be used to determine the strength of steel shape and plate for purposes ofcalculation. The permissible strength for weld metal should be taken in accordance with thebuilding code. Additional information on material properties may be found in the InterimGuidelines of Chapter 8.


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Table 7-1Table 7.5.1-1 - Properties for Use in Connection Design

Material Fy (ksi) Fy m (ksi) Fu (ksi)A36 36 use values for

Dual Certified58

Dual Certified Beam Axial, Flexural3

Shape Group 1 Shape Group 2 Shape Group 3 Shape Group 4 Through-Thickness

50

-

551

581

571

541

-

65 min.

Note 5A572 Column/Beam Axial, Flexural3

Shape Group 1 Shape Group 2 Shape Group 3 Shape Group 4 Shape Group 5 Through-Thickness

50

-

581

581

571

571

551

-

65 min.

, Note 5A9922 Use same values as ASTM A572A913-50 Axial, Flexural Through-thickness

50-

581

-65 min., Note 5

A913--— 65 Axial, Flexural Through-thickness

65 751(4) 80 min.Note 5

Notes:1. Based on coupons from web. For thick flanges,

the Fy flange is approximately 0.95 Fy web.2. See Commentary3. Values based on (SSPC-1994)4. ASTM A913, Grade 65 material is not recommended for use in the beams of moment resisting frames5. See Commentary

Commentary: Table7.5.1-1 Note 2 - The ASTM A992 specification wasspecifically developed by the steel industry in response to expressed concerns ofthe design community with regard to the permissible variation in chemistry andmechanical properties of structural steel under the A36 and A572 specifications. This new specification, which was adopted in late 1998, is very similar to ASTMA572, except that it includes somewhat more restrictive limits on chemistry andon the permissible variation in yield and ultimate tensile stress, as well as theratio of yield to tensile strength. At this time, no statistical data base is availableto estimate the actual distribution of properties of material produced to thisspecification. However, the properties are likely to be very similar, albeit withless statistical scatter, to those of material recently produced under ASTM A572,Grade 50.


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Table 7.5.1-1 Note 5 -In the period immediately following the Northridgeearthquake, the Seismology Committee of the Structural Engineers Association ofCalifornia and the International Conference of Building Officials issued InterimRecommendation No. 2 (SEAOC-1995) to provide guidance on the design ofmoment resisting steel frame connections. Interim Recommendation No. 2included a recommendation that the through-thickness stress demand on columnflanges be limited to a value of 40 ksi, applied to the projected area of beamflange attachment. This value was selected somewhat arbitrarily, to ensure thatthrough-thickness yielding did not initiate in the column flanges of moment-resisting connections and because it was consistent with the successful tests ofassemblies with cover plates conducted at the University of Texas at Austin(Engelhardt and Sabol - 1994), rather than being the result of a demonstratedthrough-thickness capacity of typical column flange material. Despite thesomewhat arbitrary nature of the selection of this value, its use often controls theoverall design of a connection assembly including the selection of cover platethickness, haunch depth, and similar parameters.

It would seem to be important to prevent the inelastic behavior of connectionsfrom being controlled by through-thickness yielding of the column flanges. Thisis because it would be necessary to develop very large local ductilities in thecolumn flange material in order to accommodate even modest plastic rotationdemands on the assembly. However, extensive investigation of the through-thickness behavior of column flanges in a “T” joint configuration reveals thatneither yielding, nor through-thickness failure are likely to occur in theseconnections. Barsom and Korvink (1997) conducted a statistical survey ofavailable data on the tensile strength of rolled shape material in the through-thickness direction. These tests were generally conducted on small diametercoupons, extracted from flange material of heavy shapes. The data indicates thatboth the yield stress and ultimate tensile strength of this material in the through-thickness direction is comparable to that of the material in the direction parallelto rolling. However, it does indicate somewhat greater scatter, with a number ofreported values where the through-thickness strength was higher, as well as lowerthan that in the longitudinal direction. Review of this data indicates with highconfidence that for small diameter coupons, the yield and ultimate tensile valuesof the material in a through-thickness direction will exceed 90% and 80%respectively of the comparable values in the longitudinal direction. the actualThe causes for through-thickness failures of column flanges (types C2, C4, andC5), observed both in buildings damaged by the Northridge Earthquake and insome test specimens, are not well understood. They are thought to be a functionof the metallurgy and “purity” of the steel; conditions of loading including thepresence of axial load and rate of loading application; conditions of tri-axialrestraint; conditions of local hardening and embrittlement within the weld’s heataffected zone; stress concentrations induced by the presence of backing bars and


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defects at the root of beam flange to column flange welds; and by the relationshipof the connection components as they may affect flange bending stresses andflange curvature induced by panel zone yielding. Given the many complex factorswhich can affect the through-thickness strength of the column flange,determination of a reliable basis upon which to set permissible design stresseswill require significant research. Such research is currently being conductedunder the SAC phase II program.

While this statistical distribution suggests the likelihood that the through-thickness strength of column flanges could be less than the flexural strength ofattached beam elements, testing of more than 40 specimens at Lehigh Universityindicates that this is not the case. In these tests, high strength plates,representing beam flanges and having a yield strength of 100 ksi were welded tothe face of A572, Grade 50 and A913, Grade 50 and 65 column shapes, tosimulate the portion of a beam-column assembly at the beam flange. Thesespecimens were placed in a universal testing machine and loaded to produce highthrough-thickness tensile stresses in the column flange material. The testssimulated a wide range of conditions, representing different weld metals as welland also to include eccentrically applied loading. In 40 of 41 specimens tested,the assembly strength was limited by tensile failure of the high strength beamflange plate as opposed to the column flange material. In the one failure thatoccurred within the column flange material, fracture initiated at the root of a low-toughness weld, at root defects that were intentionally introduced to initiate sucha fracture.

The behavior illustrated by this test series is consistent with mechanics ofmaterials theory. At the joint of the beam flange to column flange, the material isvery highly restrained. As a result of this, both the yield strength and ultimatetensile strength of the material in this region is significantly elevated. Underthese conditions, failure is unlikely to occur unless a large flaw is present that canlead to unstable crack propagation and brittle fracture. In light of this evidence,Interim Guidelines Advisory No. 2 deletes any requirement for evaluation ofthrough-thickness flange stress in columns.

Interim Recommendation No. 2 (SEAOC-1995) included a value of 40 ksi,applied to the projected area of beam flange attachment, for the through-thickness strength to be used in calculations. This value was selected because itwas consistent with the successful tests of assemblies with cover plates conductedat the University of Texas at Austin (Engelhardt and Sabol - 1994). However,because of the probable influence of all the factors noted above, this value canonly be considered to reflect the specific conditions of those tests and specimens.

Although reduced stresses at the column face produced acceptable results inthe University of Texas tests, the key to that success was more likely the result of


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forcing the plastic hinge away from the column than reduction of the through-thickness stress by the cover plates. Reduction of through-thickness columnflange stress to ever lower levels by the use of thicker cover plates is notrecommended, since such cover plates will result in ever higher forces on the faceof the column flange as well as larger weldments with potential for enlarged heataffected zones, higher residual stresses and conditions of restraint.

Since the initial publication of the Interim Guidelines, a significant number oftests have been performed on reduced beam section connections (See section7.5.3), most of which employed beam flanges which were welded directly to thecolumn flanges using improved welding techniques, but without reinforcementplates. No through-thickness failures occurred in these tests despite the fact thatcalculated through-thickness stresses at the root of the beam flange to columnflange joint ranged as high as 58 ksi. The successful performance of these weldedjoints is most probably due to the shifting of the yield area of the assembly awayfrom the column flange and into the beam span. Based on the indications of theabove described tests, and noting the undesirability of over reinforcingconnections, it is now suggested that a maximum through-thickness stress of0.9Fyc may be appropriate for use with connections that shift the hinging awayfrom the column face. Notwithstanding this recommendation, engineers are stillcautioned to carefully consider the through-thickness issue when these otherpreviously listed conditions which are thought to be involved in this type offailure are prevalent. Connections relying on through-thickness strength can notbe considered to be fully reliable until the influence of the other parametersdiscussed above can be fully understood. A high amount of structuralredundancy is recommended for frames employing connections which rely onthrough-thickness strength of the column flange.

Notwithstanding all of the above, successful tests using cover plates and othermeasures of moving hinges (and coincidentally reducing through-thickness stress)continue to be performed. In the interim, engineers choosing to utilizeconnections relying on through-thickness strength should recognize that despitethe successful testing, connections relying on through-thickness strength can notbe considered to be fully reliable until the influence of the other parametersdiscussed above can be fully understood. A high amount of structuralredundancy is recommended for frames employing connections which rely onthrough-thickness strength of the column flange.

7.5.2 Design Procedure - Strengthened Connections

The following procedure may be followed to size the various elements of strengthenedconnection assemblies that are intended to promote formation of plastic hinges within the beamspan by providing a reinforced beam section at the face of the column. Section 7.5.3 provides a


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modified procedure recommended for use in the design of connection assemblies using reducedbeam sections to promote similar inelastic behavior. Begin by selecting Select a connectionconfiguration, such as one of those indicated in Sections 7.9.1, 7.9.2, 7.9.3, 7.9.4, or 7.9.5, thatwill permit the formation of a plastic hinge within the beam span, away from the face of thecolumn, when the frame is subjected to gravity and lateral loads. Then proceed as described inthe following sections. The following procedure should be followed to size the various elementsof the connection assembly:

7.5.2.1 Determine Plastic Hinge Locations

For beams with gravity loads representing a small portion of the total flexural demand, thelocation of the plastic hinge may be assumed to occur as indicated in Table 7.5.2.1-1 at a distanceequal to 1/3 of the beam depth from the edge of the reinforced connection (or start of the reducedbeam section), unless specific test data for the connection indicates that a different location valueis more appropriate. Refer to Figure 7-2Figure 7.5.2.1-1.

Table 7.5.2.1-1 Plastic Hinge Location - Strengthened Connections

Connection Type Reference Section Hinge Location “sh”

Cover plates Sect. 7.9.1 d/4 beyond end of cover plates

Haunches Sect. 7.9.3, 7.9.4 d/3 beyond toe of haunch

Vertical Ribs Sect. 7.9.2 d/3 beyond toe of ribs

L

Bea

m d

epth

- d

Edge

of r

einf

orce

dco

nnec

tion

Edge

of r

einf

orce

dco

nnec

tion

sh=d/3

L’

Plastichinge

Connectionreinforcementsh=

d/4

Figure 7-2 Figure 7.5.2.1-1 - Location of Plastic Hinge

Commentary: The suggested locations for the plastic hinge, at a distance d/3away from the end of the reinforced section (or beginning of reduced section)indicated in Table 7.5.2.1-1 and Figure 7.5.2.1-1 are is based on the observedbehavior of test specimens, with no significant gravity load present. If significant


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gravity load is present, this can shift the locations of the plastic hinges, and in theextreme case, even change the form of the collapse mechanism. If flexuraldemand on the girder due to gravity load is less than about 30% of the girderplastic capacity, this effect can safely be neglected, and the plastic hingelocations taken as indicated. If gravity demands significantly exceed this level,then plastic analysis of the girder should be performed to determine theappropriate hinge locations. In zones of high seismicity (UBC Zones 3 and 4,and NEHRP Map Areas 6 and 7), gravity loading on the girders of earthquakeresisting frames typically has a very small effect, unless tributary areas forgravity loads are large.

7.5.2.2 Determine Probable Plastic Moment at Hinges

Determine the probable value of the plastic moment, Mpr, at the location of the plastic hingesas:

M M Z Fpr p b y= =β β (7.5.2.2-12)

where: ß is a coefficient that adjusts the nominal plastic moment to the estimated hingemoment based on the mean yield stress of the beam material and the estimatedstrain hardening. A value of 1.2 should be taken for β for ASTM A572, A992 andA913 steels. When designs are based upon calculations alone, an additional factoris recommended to account for uncertainty. In the absence of adequate testing ofthe type described above, ß should be taken as 1.4 for ASTM A572 and for A913,Grades 50 and 65 steels. Where adequate testing has been performed ß should bepermitted to be taken as 1.2 for these materials.

Zb is the plastic modulus of the section

Commentary: In order to compute β, the expected yield strength, strainhardening and an appropriate uncertainty factor need to be determined. Thefollowing assumed strengths are recommended:

Expected Yield: The expected yield strength, for purposes of computing (Mpr) may be taken as:

Fye = 0.95 Fym (7.5.2.2-2-3)

The 0.95 factor is used to adjust the yield stress in the beam web, wherecoupons for mill certification tests are normally extracted, to the value in thebeam flange. Beam flanges, being comprised of thicker material, typically havesomewhat lower yield strengths than do beam web material.


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Fy m for various steels are as shown in Table 7-1 Table 7.5.1-1, based on asurvey of web coupon tensile tests (Steel Shape Producers Council - 1994). Theengineer is cautioned that there is no upper limit on the yield point for ASTM A36steel and consequently, dual-certification steel having properties consistent withASTM A572, Grade 50 is routinely supplied when ASTM A36 is specified. Consequently, it is the recommendation here that the design of connections bebased on an assumption of Grade 50 properties, even when A36 steel is specifiedfor beams. It should be noted that at least one producer offers A36 steel with amaximum yield point of 50 ksi in shape sizes ranging up to W 24x62. Refer to thecommentary to Section 8.1.3 for further discussion of steel strength issues.

Strain Hardening: A factor of 1.1 is recommended for use with the mean yieldstress in the foregoing table when calculating the probable plastic momentcapacity Mpr.. The 1.1 factor for strain hardening, or other sources of strengthabove yield, agrees fairly well with available test results. The 1.1 factor couldunderestimate the over-strength where significant flange buckling does not act asa gradual limit on the beam strength. Nevertheless, the 1.1 factor seems areasonable expectation of over-strength considering the complexities involved.

Modeling Uncertainty: Where a design is based on approved cyclic testing, themodeling uncertainty may be taken as 1.0, otherwise the recommended value is1.2. When the Interim Guidelines were first published, the β coefficient includeda 1.2 factor to account for modeling uncertainty when connection designs werebased on calculations as opposed to a specific program of qualification testing. The intent of this factor was twofold: to provide additional conservatism in thedesign when specific test data for a representative connection was not availableand also as an inducement to encourage projects to undertake connectionqualification testing programs. After the Interim Guidelines had been in use forsome time, it became apparent that this approach was not an effective inducementfor projects to perform qualification testing, and also that the use of an overlylarge value for the β coefficient often resulted in excessively large connectionreinforcing elements (cover plates, e.g.) and other design features that did notappear conducive to good connection behavior. Consequently, it was decided toremove this modeling uncertainty factor from the calculation of β.

In summary, for Grade 50 steel, we have:

β = [0.95 (54 ksi to 58 ksi)/50 ksi] (1.1) 1.2) = 1.35 t0 1.45, say 1.4

β = [0.95 (54 ksi to 59 ksi)/50 ksi] (1.1) = 1.13 to 1.21, say 1.2


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7.5.2.3 Determine Shear at the Plastic Hinge

The shear at the plastic hinge should be determined by statics, considering gravity loads actingon the beam. A free body diagram of that portion of the beam between plastic hinges, is a usefultool for obtaining the shear at each plastic hinge. Figure 7-3 Figure 7.5.2.3-1 provides anexample of such a calculation. For the purposes of such calculations, gravity load should be basedon the load combinations required by the building code in use.

L

sh

L’

Plastichinge P

L/2

P

Mpr MprL’Vp

taking the sum of moments about “A” = 0Vp ={Mpr + Mpr + P L’/2 + wL’2/2}/L’

“A”

VAw

Note: Gravity loads can effect the location of the plastic hinges. If 2Mpr /L’is less then the gravity shear in the free body (in this case P/2 + wL’/2), then the plastichinge location will shiftsignificantly and L’ must beadjusted, accordingly

w

Figure 7-3 Figure 7.5.2.3-1- Sample Calculation of Shear at Plastic Hinge

Commentary: The UBC gives no specific guidance on the load combinations touse with strength level calculations while the NEHRP Recommended Provisionsdo specify load factors for the various dead, live and earthquake components ofload. For designs performed in accordance with the UBC, it is customary to useunfactored gravity loads when checking the strength of elements.

7.5.2.4 Determine Strength Demands at Each Critical Section

In order to complete the design of the connection, including sizing the various plates andjoining welds which make up the connection, it is necessary to determine the shear and flexuralstrength demands at each critical section. These demands may be calculated by taking a free bodyof that portion of the connection assembly located between the critical section and the plastichinge. Figure 7-4 Figure 7.5.2.4-1 demonstrates this procedure for two critical sections, for thebeam shown in Figure 7-3 Figure 7.5.2.3-1.


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Plastichinge

Vp

Mpr

Plastichinge

Vp

Mpr

x

Mf

x+dc/2

dc

Mf=Mpr +Vpx

Mc

Mc=Mpr +Vp(x+dc/2)

Critical Section at Column Face Critical Section at Column Centerline

Figure 7-4 Figure 7.5.2.4-1 - Calculation of Demands at Critical Sections

Commentary: Each unique connection configuration may have different criticalsections. The vertical plane that passes through the joint between the beamflanges and column (if such joining occurs) will typically define at least one suchcritical section, used for designing the joint of the beam flanges to the column, aswell as evaluating shear demands on the column panel zone. A second criticalsection occurs at the center line of the column. Moments calculated at this pointare used to check strong column - weak beam conditions. Other critical sectionsshould be selected as appropriate.

7.5.2.5 Check for Strong Column - Weak Beam Condition

When required by the building code, the connection assembly should be checked to determineif strong column - weak beam conditions are satisfied. In lieu of UBC-94 equation 11-3.1{NEHRP-91 equation 10-3}, the following equation should be used:

Z (F f ) M 1.0c yc a c− >∑ ∑ (7.5.2.5-1-4)

where: Zc is the plastic modulus of the column section above and below the connectionFyc is the minimum specified yield stress for the column above and belowfa is the axial load in the column above and belowΣMc is the moment calculated at the center of the column in accordance with

Section 7.5.2.4 sum of the column moments at the top and bottom of the panel zone, respectively, resulting from the development of the probable beam plastic moments, Mpr, within each beam in the connection.


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Commentary: The building code provisions for evaluating strong column - weakbeam conditions presume that the flexural stiffness of the columns above andbelow the beam are approximately equal, that the beams will yield at the face ofthe column, and that the depth of the columns and beams are small relative totheir respective span lengths. This permits the code to use a relatively simpleequation to evaluate strong column - weak beam conditions in which the sum ofthe flexural capacities of columns at a connection are compared against the sumsof the flexural capacities in the beams. The first publication of the InterimGuidelines took this same approach, except that the definition of ΣMc wasmodified to explicitly recognize that because flexural hinging of the beams wouldoccur at a location removed from the face of the column, the moments deliveredby the beams to the connection would be larger than the plastic moment strengthof the beam. In this equation, ΣMc was taken as the sum of the moments at thecenter of the column, calculated in accordance with the procedures of Sect.7.5.2.4.

(L-L’)/2

d ph t

h b

Mpr

Vp

Vp

Mpr

Vc

Vc+Vf

Mct

Mcb

assumed point of zero moment

Note:The quantities Mpr, Vp, L, and L’ areas previously identified. Vf is the incremental shear distributedto the column at the floor level.Other quantities are as shown.

Vf

( )[ ] ( )

( )

VM V L L V h d

h d h

M V h

M V V h

M M M

cpr p f b p

b p t

ct c t

cb c f b

c ct cb

=+ − − +

+ +== +

= +

∑

∑

' ) / /2 2

Figure 7.5.2.5-1 Calculation of Column Moment for Strong ColumnEvaluation

This simplified approach is not always appropriate. If non-symmetricalconnection configurations are used, such as a haunch on only the bottom side ofthe beam, this can result in an uneven distribution of stiffness between the twocolumn segments, and premature yielding of the column, either above, or below,the beam-column connection. Also, it was determined that for connectionconfigurations in which the panel zone depth represents a significant fraction of


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the total column height, such as can occur in some haunched and side-platedconnections, the definition of ΣMc contained in the initial printing of theGuidelines could lead to excessive conservatism in determining whether or not astrong column - weak beam condition exists in a structure. Consequently, InterimGuidelines Advisory No. 1 adopted the current definition of ΣMc for use in thisevaluation. This definition requires that the moments in the column, at the topand bottom of the panel zone be determined for the condition when a plastichinge has formed at all beams in the connection. Figure 7.5.2.5-1 illustrates amethod for estimating this quantity.

7.5.2.6 Check Column Panel Zone

The adequacy of the shear strength of the column panel zone should be checked. For thispurpose, the term 0.8ΣMf should be substituted for the term 0.8ΣMs in UBC-94 Section2211.7.2.1 {0.9ΣφbMp in NEHRP-91 Section 10.10.3.1}, repeated below for convenience ofreference. Mf is the calculated moment at the face of the column, when the beam mechanismforms, calculated as indicated in Section 7.5.2.4 above. In addition, it is recommended that thealternative design criteria indicated in UBC-94 Section 2211.7.2.1 (NEHRP-91 Sect. 10.10.3.1),permitting panel zone shear strength to be proportioned for the shear induced by bendingmoments from gravity loads plus 1.85 times the prescribed seismic forces, not be used. Forconvenience of reference, UBC-94 Section 2211.7.2.1 is reproduced below, edited, to indicate therecommended application.

2211.7.2.1 Strength (edited). The panel zone of the joint shall be capable of resisting theshear induced by beam bending moments due to gravity loads plus 1.85 times theprescribed seismic forces, but the shear strength need not exceed that required to develop0.8ΣMs 0.8ΣMf of the girders framing into the column flanges at the joint. The joint panelzone shear strength may be obtained from the following formula:

V 0.55F d t3b td d ty c

c c f2

b c

= +

1 (11-1)

where: bc = width of column flangedb = the depth of the beam (including any haunches or cover plates)dc = the depth of the columnt = the total thickness of the panel zone including doubler platestcf = the thickness of the column flange

Commentary: The effect of panel zone shear yielding on connection behavior isnot well understood. In the past, panel zone shear yielding has been viewed as abenign, or even beneficial mechanism that permits overall frame ductilitydemands to be accommodated while minimizing the extent of inelastic behaviorrequired of the beam and beam flange to column flange joint. The criteria


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permitting panel zone shear strength to be proportioned for the shears resultingfrom moments due to gravity loads plus 1.85 times the design seismic forces wasadopted by the code specifically to permit designs with somewhat weak panelzones. However, during recent testing of large scale connection assemblies withweak panel zones, it has been noted that in order to accommodate the large sheardeformations that occur in the panel zone, extreme “kinking” deformations wereinduced into the column flanges at the beam flange to column flange welded joint. While this did not lead to premature joint failure in all cases, it is believed tohave contributed to such premature failures in at least some of the specimens. The recommendations of this section are intended to result in stronger panelzones than previously permitted by the code, thereby avoiding potential failuresdue to this kinking action on the column flanges.

7.5.3 Design Procedure - Reduced Beam Section Connections

The following procedure may be followed to size the various elements of reduced beamsection (RBS) assemblies with circular curved reductions in beam flanges, such as shown inFigure 7.5.3-1., such as those indicated in Section 7.9.6 indicates other configurations for suchconnections, however, the circular curved configuration shown in Figure 7.5.3-1 is currentlypreferred. RBS assemblies are intended to promote the formation of plastic hinges within thebeam span by developing a segment of the beam with locally reduced section properties andstrength. Begin by selecting an RBS configuration, such as one of those indicated in Figure 7.5.3-1, that will permit the formation of a plastic hinge within the reduced section of the beam. Of theconfigurations shown in the figure, the circular curved configuration is preferred.

c l

a

a

R = radius of cut =4a + l8a

2 2

bf

Figure 7.5.3-1 Geometry of Reduced Beam Section

Commentary: Connection assemblies in which inelastic behavior is shifted awayfrom the column face through development of a segment of the beam withintentionally reduced properties, so-called reduced beam section (RBS) or“dogbone” connections, appear to have the potential to provide an economical


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solution to the WMSF connection problem. These recommendations are based onlimited design configurations that have successfully been tested ing that has beenconducted of these types of connections to date. While a A large number of RBStests have been conducted, these tests have not included the effects of floor slabsor loading rates approximating those that would be produced by a building’sresponse to earthquake ground motionsincluding some tests of assemblies withfloor slabs present. Extensive additional testing of RBS connections, intended toexplore these and other factors relevant to connection performance, are currentlyplanned under funding provided by NIST and the SAC phase II program. In theinterim, designers specifying RBS connections may wish to consider provision ofdetails to minimize the participation of the slab in the flexural behavior of thebeam at the reduced section. The criteria presented in this section are partiallybased on a draft procedure developed by AISC (Iwankiw, 1996).

ReducedSection Drilled Constant Drilled Tapered

Circular

Straight Tapered

Figure 7.5.3-2 Alternative Reduced-Beam Section PatternsFigure 7.5.3-1 Reduced Beam Section Patterns

Several alternative configurations of RBS connections have also been testedto date. As indicated in Figures 7.5.3-21 and 7.9.6-1, these include constantsection, tapered section, curved section, and drilled hole patterns. It appears thatseveral of these configurations are more desirable than others. In particular, thedrilled hole section patterns have been subject to tensile failure across thereduced net section of the flange through the drill holes. A few RBS tests utilizingstraight or tapered cuts have failed within the reduced section at plastic rotationdemands less than recommended by these Guidelines. In all of these cases, thefailure occurred at locations at which there was a change in direction of the cutsin the beam flange, resulting in a geometric stress riser or notch effect. It is alsoreported that one of these tests failed at the beam flange continuity plate - tocolumn flange joint. There have been no reported failures of RBS connection


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assemblies employing the circular curved flange cuts, and therefore, this is thepattern recommended in these Guidelines. This would appear, therefore, to be amore desirable configuration, although some successful tests have beenperformed using the straight and tapered configurations.

It is important that the pattern of any cuts made in the flange be proportionedso as to avoid sharp cut corners. All corners should be rounded to minimizenotch effects and in addition, cut edges should be cut or ground in the directionof the flange length to have a surface roughness meeting the requirements of AWSC4.1-77 class 4, or smootherroughness value less than or equal to 1,000, asdefined in ANSI/ASME B46.1.

Concerns have been raised by some engineers over the strength reductioninherent in the RBS. Clearly, code requirements for strength, considering gravityloads and gravity loads in combination with wind, seismic and other loads mustbe met. For higher seismic zones, beam sizes are typically governed by elasticstiffness considerations (drift control) and this must be addressed. Also, forseismic loads, the Building Codes typically require that connections for SpecialMoment Resisting Frames must develop the “strength” or the “plastic bendingmoment” of the beam. There may be a problem of semantics where theserequirements are applied to a system using RBS connections. Is the RBS part ofthe connection or is it part of the beam, the strength of which must be developedby the connection? Clearly, the latter interpretation should be applied.

Notwithstanding the above, it must be kept in mind that, although unstated,and typically not quantified, there is inherent in design practice an impliedrelationship between the elastic behavior that we analyze and the inelasticbehavior which the building is expected to experience. Elastic drift limitationscommonly used are considered to be related to the anticipated inelastic drifts andultimate lateral stability of the framed structure in at least an intuitivelypredictable manner. It can be shown that RBS’s such as those that have beentested will reduce the elastic stiffness (increase the drift) on the order of 5%.However, because of the reduction in strength, the effect on the inelastic drift maybe more significant. Thus, it seems prudent to require that the RBS maintain areasonably high proportion of the frame inelastic strength. For the connectionstested to date, the inelastic strength of the RBS section has been in the range of70% of that of the full section. However, the moment demand at the face of thecolumn, corresponding to development of this reduced section strength, is likelyto be in the range of 85% to 90% of the strength of the full beam. This seems tobe quite reasonably high considering the accuracy of other seismic designassumptions.

Although the use of RBS designs tends to reduce the total strength demand onthe beam flange - to - column flange connection, relative to strengthened


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connections, designs utilizing RBS configurations should continue to follow therecommendations for beam flange continuity plates, weld metal and base metalnotch toughness recommended by the Interim Guidelines for strengthenedconnections.

7.5.3.1 Determine Reduced Section and Plastic Hinge Locations

The reduced beam section should be located at a sufficient distance from the face of thecolumn flange (dimension “c” in Figures 7.5.3-1 and 7.5.3.1-1) to avoid significant inelasticbehavior of the material at the beam flange - to - column flange joint. Based on testing performedto date, it appears that a value of “c” on the order of ½ to ¾ of the beam width, bf, is sufficient.d/4 (where “d” is the beam depth) is sufficient. The total length of the reduced section of beamflange (dimension “l” in Figures 7.5.3-1 and 7.5.3.1-1) should be on the order of 0.65d to 0.85d,where d is the beam depth.3d/4 to d. The location of the plastic hinge, sh,, may be taken as ½ thelength of the cut-out, l.indicated in Table 7.5.3.1-1, unless test data indicates a more appropriatevalue should be used. When tapered configurations are utilized, the slope of the tapered cut in thebeam flange should be arranged such that the variation of the plastic section modulus, Zx, withinthe reduced section approximates the moment gradient in the beam during the condition whenplastic hinges have formed within the reduced beam sections at both ends.

L

Bea

m d

epth

- d

L’Plastichinge

reducedsection

c

l

sh

Figure 7.5.3.1-1 Critical Dimensions - RBS Assemblies

7.5.3.2 Determine Strength and Probable Plastic Moment in RBS

The RBS may be proportioned to meet the following criteria:


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1. The section at the RBS should be sufficient to satisfy the strength criteria specified bythe building code for Dead, Live, Seismic, Snow, Wind, and other applicable designforces.

2. The elastic stiffness of the frame, considering the effects of the RBS, should besufficient to meet the drift requirements specified by the code, under the design seismicand other forces.

3. The expected stress in the beam flange - to - column flange weld, under the applicationof gravity forces and that seismic force that results in development of the probableplastic moment of the reduced section at both ends of the beam, should be less than orequal to the strength of the weld, as indicated in Section 7.2.2 of the InterimGuidelines.

4. The expected through-thickness stress on the face of the column flange, calculated asMf/Sc, under the application of gravity forces and that seismic force that results indevelopment of the probable plastic moment of the reduced section at both ends of thebeam, should be less than or equal to the values indicated in Section 7.5.1, where Mf isthe moment at the face of the column flange, calculated as indicated in Section 7.5.2.4,and Sc is the elastic section modulus of the beam at the connection considering weldreinforcement, bolt holes, reinforcing plates, etc. The maximum moment at the face ofthe column should be in the range of 85 percent to 100 percent of the beam’s expectedplastic moment capacity. The depth of cut-out, a, should be selected to be less than orequal to bf/4.

The plastic section modulus of the RBS may be calculated from the equation:

( )Z Z b t d tRBS x R f f= − − (7.5.3.2-1)

where:ZRBS is the plastic section modulus of the reduced beam sectionZx is the plastic modulus of the unreduced sectionbR is the total width of material removed from the beam flangetf is the thickness of the beam flanged is the depth of the beam

The probable plastic moment, Mpr, at the RBS shall be calculated from the equation:

M Z Fpr RBS y= β (7.5.3.2-2)

where:ZRBS is the plastic section modulus of the reduced beam sectionβ is as defined in Section 7.5.2.2


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The strength demand on the beam flange - to - column flange weld and on the face of thecolumn may be determined by following the procedures of Section .7.5.2.3 and 7.5.2.4 of theInterim Guidelines, using the value of Mpr determined in accordance with Eq. 7.5.3.2-2.

Commentary: Initial design procedures for RBS connections published by SACrecommended that sufficient reduction of the beam flange be made to maintainflexural stresses in the beam, at the column face, below the anticipated through-thickness yield strength of the column flange material. Since the publication ofthose recommendations, extensive testing of RBS connections has been conducted,both with and without composite slabs. The testing conducted to date on RBSspecimens This testing has typically been for configurations that would result insomewhat larger strength demands at the face of the column flange thansuggested by the criteria originally published by SAC. contained in this Advisory. Typically, the tested specimens had reductions in the beam flange area on theorder of 35% to 45% and produced moments at the face of the column thatresulted in stresses on the weld and column as large as large as 90 to 100% of theexpected material strength of the beam, which is often somewhat in excess of thethrough-thickness yield strength of the column material. The specimens in thesetests all developed acceptable levels of inelastic deformation. Recent studiesconducted for SAC at Lehigh University confirm that the significant conditions ofrestraint that exist at the beam flange to column flange joint results insubstantially elevated column through-thickness strength, negating a need toreduce flexural stresses below the anticipated column yield strength. In view ofthis evidence, SAC has elected to adopt design recommendations consistent withconfigurations that were successfully tested. The criteria contained in thisAdvisory suggest that these demands be reduced to a level which would maintainweld stresses within their normally specified values and through-thickness columnflange stresses at the same levels recommended for strengthened connections. This may require the beam flanges to be reduced by as much as 50% or more forsome frame configurations, or that supplemental reinforcement such as coverplates or vertical ribs be provided in addition to the reduced section. Thisapproach was taken to maintain consistency with the criteria recommended forstrengthened connections and with the knowledge that the factors affecting theperformance of these connections are not yet fully understood.

7.5.3.3 Strong Column - Weak Beam Condition

The adequacy of the design to meet strong column - weak beam conditions should be checkedin accordance with the procedures of Section 7.5.2.5


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7.5.3.4 Column Panel Zone

The adequacy of the column panel zone should be checked in accordance with the proceduresof Section 7.5.2.6.

7.5.3.5 Lateral Bracing

The reduced section of the beam flanges should be provided with adequate lateral support toprevent lateral-torsional buckling of the section. Lateral braces should be located within adistance equal to 1/2 the beam depth from the expected location of plastic hinging, but should notbe located within the reduced section of the flanges.

Commentary: Unbraced compression flanges of beams are subject to lateral-torsional buckling, when subjected to large flexural stresses , such as occur in theplastic hinges of beams reduced sections of RBS connections during response tostrong ground motion. To prevent such behavior lateral-torsional buckling, it isrecommended that both flanges of beams be provided with lateral support.Section 9.8 of the 1997 AISC Seismic Specification requires such bracing ingeneral, and specifically states as follows:

“Both flanges of beams shall be laterally supported directly or indirectly. The unbraced length between lateral supports shall not exceed 2500ry/Fy. Inaddition, lateral supports shall be placed near concentrated forces, changesin cross section and other locations where analysis indicates that a plastichinge will form during inelastic deformations of the SMF.”

Adequate lateral support of the top flanges of beams supporting concretefilled metal deck or formed slabs can usually be obtained through the normalwelded attachments of the deck to the beam or through shear studs. Lateralsupport of beam flanges can also be provided through the connections oftransverse framing members or by provision of special lateral braces, attacheddirectly to the flanges. Such attachments should not be made within the reducedsection of the beam flange as the welding or bolting required to make suchattachments can lead to premature fracturing in these regions of high plasticdemands.

For beams in moment-resisting frames, it has traditionally been assumed thatthe direct attachment of the beam flanges to the columns provided sufficientlateral support of both beam flanges to accommodate the plastic hingesanticipated to develop in these frames at the beam-column connection. However,connection configurations like the RBS, developed following the Northridgeearthquake, are intended to promote formation of these plastic hinges at somedistance from the beam-column interface. This brings to question the adequacyof the beam flange to column flange attachments to provide the necessary lateral


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support at the plastic hinge. While this issue is pertinent for any connectionconfiguration that promotes plastic hinge formation remote from the beam-column interface, RBS connections could be more susceptible to lateral-torsionbuckling at the plastic hinge because the reductions in the beam flange used toachieve plastic hinge formation also locally reduce the torsional resistance of thesection. For that reason, FEMA-267a recommended provision of lateral bracingadjacent to the reduced beam section.

Provision of lateral bracing does result in some additional cost. Therefore,SAC has engaged in specific investigations to evaluate the effect of lateralbracing both on the hysteretic behavior of individual connections as well asoverall frame response to large lateral displacements. Until these investigationshave concluded SAC continues to recommend provision of lateral bracing forRBS connections. It should be noted that Section 9.8 of the 1997 AISC SeismicSpecification states:

“If members with Reduced Beam Sections, tested in accordance withAppendix S are used, the placement of lateral support for the member shall beconsistent with that used in the tests.”

Most testing of RBS specimens performed as part of the SAC project haveconsisted of single beams cantilevered off a column to simulate the exteriorconnection in a multi-bay moment-resisting frame. The beams have generallybeen braced at the end of the cantilever length, typically located about 100 inchesfrom the face of the column. For the ASTM A572, Grade 50, W36x150 sectionstypically tested, this results in a nominal length between lateral supports that iscomparable to 2500ry/Fy.

The appropriate design strength for lateral bracing of compression elementshas long been a matter of debate. Most engineers have applied “rules of thumb”that suggest that the bracing element should be able to resist a small portion,perhaps on the order of 2% to 6% of the compressive force in the element beingbraced, applied normal to the line of action of the compression. A recentsuccessful test of an RBS specimen conducted at the University of Texas at Austinincorporated lateral bracing with a strength equal to 6% of the nominalcompressive yield force in the reduced section.

7.5.3.6 Welded Attachments

Headed studs for composite floor construction should not be placed on the beam flangebetween the face of the column and the extreme end of the RBS, as indicated in Figure 7.5.3.6-1.Other welded attachments should also be excluded from these regions of the beam.


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Reduced beam section

welded attachment permitted

welded attachment prohibited

Figure 7.5.3.6-1 Welded Attachments to RBS Beams

Commentary: There are two basic reasons for omitting headed studs in theregion between the reduced beam section and the column. The first of these isthat composite action of the slab and beam can effectively counteract thereduction in beam section properties achieved by the cutouts in the top beamflange. By omitting shear studs in the end region of the beam, this compositebehavior is neutralized, protecting the effectiveness of the section reduction. Thesecond reason is that the portion of the beam at the reduced section is expected toexperience large cyclic inelastic strains. If welded attachments are made to thebeam in this region, the potential for low-cycle fatigue of the beam, under theselarge cyclic inelastic strains is greatly increased. For this same reason, otherwelded attachments should also be excluded from this region.

7.6 Metallurgy and Welding


7.7 Quality Control/Quality Assurance


7.8 Guidelines on Other Connection Design Issues



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7.8.1 Design of Panel Zones

No current recommendations are made to supplement or modify the UBC-1994 {NEHRP-91}provisions for the design of panel zones, other than as indicated in Section 7.5.2.6, above. Panelzone demands should be calculated in accordance with Section 7.5.2.6. As with other elements ofthe connection, available panel zone strength should be computed using minimum specified yieldstress for the material, except when the panel zone strength is used as a limit on the requiredconnection strength, in which case Fym should be used.

Where connection design for two-sided connection assemblies is relying on test data for one-sided connection assemblies, consideration should be given to maintaining the level of panel zonedeformation in the design to a level consistent with that of the test, or at least assume that thepanel zone must remain elastic, under the maximum expected shear demands.

Commentary: At present, no changes are recommended to the code requirementsgoverning the design of panel zones, other than in the calculation of the demand. As indicated in Section 7.5.2.6, it is recommended that the formulation for panelzone demand contained in the UBC, based on 1.85 times the prescribed seismicforces, not be utilized. This formulation, which is not contained in either theAISC Seismic Provisions or the NEHRP Provisions, is felt to lead to the design ofpanel zones that are excessively flexible and weak in shear. There is evidencethat panel zone yielding may contribute to the plastic rotation capability of aconnection. However, there is also concern and some evidence that if thedeformation is excessive, a kink will develop in the column flange at the joint withthe beam flange and, if the local curvature induced in the beam and columnflanges is significant, can contribute to failure of the joint. This would suggestthat greater conservatism in column panel zone design may be warranted.

In addition to the influence of the deformation of the panel zone on theconnection performance, it should be recognized that the use of doubler platesand especially the welding associated with them is likely to be detrimental to theconnection performance. It is recommended that the Engineer consider use ofcolumn sizes which will not require addition of doubler plates, where practical.

7.8.2 Design of Web Connections to Column Flanges

Specific modifications to the code requirements for design of shear connections are not madeat this time. It should be noted that the emergency code change to the UBC-94 {NEHRP-94}deleted the former requirements for supplemental web welds on shear connections. This is felt tobe appropriate since these welds can apparently contribute to the potential for shear tab failure atlarge induced rotations.

When designing shear connections for moment-resisting assemblies, the designer shouldcalculate shear demands on the web connection in accordance with Section 7.5.2.4, above. For


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connection designs based on tested configurations, the web connection design should beconsistent with the conditions in the tested assemblies.

Commentary: Some engineers consider that it is desirable to develop as muchbending strength in the web as possible. Additionally, it has been observed insome laboratory testing that pre-mature slip of the bolted web connection canresult in large secondary flexural stresses in the beam flanges and the weldedjoints to the column flange. However, there is some evidence to suggest that ifflange connections should fail, welding of shear tabs to the beam web maypromote tearing of the tab weld to the column flange or the tab itself through thebolt holes, and some have suggested that welding be avoided and that webconnections should incorporate horizontally slotted holes to limit the momentwhich can be developed in the shear tab, thereby protecting its ability to resistgravity loads on the beam in the event of flexural connection failure.

Some recent finite element studies of typical connections by Goel, Popov andothers have suggested that even when the shear tab is welded, shear demands atthe connections tend to be resisted by a diagonal tension type behavior in the webthat tends to result in much of the shear being resisted by the flanges. Investigation of these effects is continuing.

7.8.3 Design of Continuity Plates


7.8.4 Design of Weak Column and Weak Way Connections


7.9 Moment Frame Connections for Consideration in New Construction


7.9.1 Cover Plate Connections

Figure 7-5 Figure 7.9.1-1 illustrates the basic configuration of cover plated connections. Short cover plates are added to the top and bottom flanges of the beam with fillet welds adequate totransfer the cover plate forces to the beam flanges. The bottom flange cover plate is shop welded to thecolumn flange and the beam bottom flange is field welded to the column flange and to the cover plate. The top flange and the top flange cover plate are both field welded to the column flange with acommon weld. The web connection may be either welded or high strength (slip critical) bolted. Limited testing of these connections (Engelhardt & Sabol - 1994), (Tsai & Popov -1988) has beenperformed. More than 30 tests of such connections have been performed, with data on at least 18 ofthese tests available in the public domain.


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A variation of this concept which has been tested successfully very recently (Forell/ElsesserEngineers -1995), uses cover plates sized to take the full flange force, without direct welding of thebeam flanges themselves to the column. In this version of the detail, the cover plate provides a crosssectional area at the column face about 1.7 times that of the beam flange area. In the detail which hasbeen tested, a welded shear tab is used, and is designed to resist a significant portion of the plasticbending strength of the beam web.

T&B

Figure 7-5 Figure 7.9.1-1 - Cover Plate Connection

Design Issues: Following the Northridge earthquake, the University of Texas at Austinconducted a program of research, under private funding, to develop a modified connectionconfiguration for a specific project. Following a series of unsuccessful tests on various types ofconnections, approximatelyApproximately eight connections similar to that shown in Figure 7-5 Figure 7.9.1-1 were have been recently tested (Engelhardt & Sabol - 1994), and they havedemonstrated the ability to achieve acceptable levels of plastic rotation provided that the beamflange to column flange welding wasis correctly executed and through-thickness problems in thecolumn flange were are avoided. This configuration is relatively economical, compared to someother reinforced configurations, and has limited architectural impact. As a result of thesefactors, and the significant publicity that followed the first successful tests of these connections,cover plated connections quickly became the predominant configuration used in the design ofnew buildings. As a result, a number of qualification tests have now been performed on differentvariations of cover plated connections, covering a wide range of member sizes ranging fromW16 to W36 beams, as part of the design process for individual building projects. The results ofthese tests have been somewhat mixed, with a significant number of failures reported. Althoughthis connection type appears to be significantly more reliable than the typical pre-Northridgeconnection, it should be expected that some connections in buildings incorporating this detailmay still be subjected to earthquake initiated fracture damage. Designers should consider usingalternative connection types, unless highly redundant framing systems are employed.


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Six of eight connections tested by the University of Texas at Austin were able to achieveplastic rotations of at least 0.025 radians, or better. Strength loss at the extreme levels of plasticrotation did not reduce the flexural capacity to less than the plastic moment capacity of thesection based on minimum specified yield strength. One specimen achieved plastic rotations of0.015 radians when a brittle fracture of the CJP weld (type W2 failure) occurred. This maypartially be the result of a weld that was not executed in conformance with the specified weldingprocedure specification. The second unsuccessful test specimen achieved plastic rotations of0.005 radian when a section of the column flange (type C2 failure) occurred. A similar failureoccurred in recent testing by Popov of a specimen with cover plates having a somewhat modifiedplan shape.

Quantitative Results: No. of specimens tested: 18Girder Size: W21 x 68 to W36 x 150Column Size: W12 x 106 to W14 x 455Plastic Rotation achieved-

6 13 Specimens : >0.025 radian1 3 Specimens: 0.015 0.005 < θp < 0.025 radian1 2 Specimens: 0.005 radian

Although apparently more reliable than the former prescriptive connection, thisconfiguration is subject to some of the same flaws including dependence dependent on properlyexecuted beam flange to column flange welds, and through-thickness behavior of the columnflange. Further these effects are somewhat exacerbated as the added effective thickness of thebeam flange results in a much larger groove weld at the joint, and therefore potentially moresevere problems with brittle heat affected zones and lamellar defects in the column. Indeed, asignificant percentage of connections of this configuration have failed to produce the desiredamount of plastic rotation.

One of the issues that must be faced by designers utilizing cover plated connections is thesequence of operations used to attach the cover plate and beam flange to the column. In oneapproach, the bottom cover plate is shop welded to the column, and then used as the backing forthe weld of the beam bottom flange to the column flange. This approach has the advantage ofproviding an erection seat and also results in a somewhat reduced amount of field welding forthis joint. A second approach is to attach the cover plate to the beam flange, and then weld it tothe column, in the field, as an integral part of the beam flange. There are tradeoffs to bothapproaches. The latter approach results in a relatively large field weld at the bottom flange withlarge heat input required into the column and beam. If this operation is not performed withproper preheat and control of the heat input, it can potentially result in an enlarged and brittleheat affected zone in both members. The first approach results in reduced heat input andtherefore, somewhat minimized potential for this effect. However, proper control of preheat andheat input remains as important in either case, as improper procedures can still result in brittleconditions in the heat affected zone. Further, the detail in which the cover plate is shop weldedto the column can lead to a notch effect for the column flange at the seam between the beam


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flange and cover plate. This is effect is illustrated in Figure 7.9.1-2. At least one specimenemploying this detail developed a premature fracture across the column flange that has beenrelated to this notch effect. This effect has been confirmed by recent fracture mechanicsmodeling of this condition conducted by Deierlein.

When developing cover plated connection details, designers should attempt to minimize thetotal thickness of beam flange and cover plate, so as to reduce the size of the complete jointpenetration weld of these combined elements to the column flange. For some frameconfigurations and member sizes, this combined thickness and the resulting CJP weld size canapproach or even exceed the thickness of the column flange. While there is no specific criteriain the AWS or AISC specifications that would suggest such weldments should not be made,judgementally they would not appear to be desirable from either a constructability orperformance perspective. As a rough guideline, it is recommended that for connections in whichboth the beam flange and cover plate are welded to the column flange, the combined thickness ofthese elements should not exceed twice the thickness of the beam flange nor 100% of thethickness of the column flange. For cover plated connections in which only the cover plate iswelded to the column flange, the same thickness limits should be applied to the cover plate.

columnflange(in tension)

cover plate

beam bottom flange

seam acts as notch

Figure 7.9.1-2 Notch Effect at Cover Plated Connections

7.9.2 Flange Rib Connections



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7.9.3 Bottom Haunch Connections

Figure 7.9.3-1 7-7 indicates the configuration of a connection with a haunch at the bottombeam flange.several potential configurations for single, haunched beam-column connections. Aswith the cover plated and ribbed connections, the intent is to shift the plastic hinge away from thecolumn face and to reduce the demand on the CJP weld by increasing the depth of the section. To date, the configuration incorporating the triangular haunch has been subjected to limitedtesting. Testing of configurations incorporating the straight haunch are currently planned, buthave not yet been performed. Several tests of this connection type were conducted by Uang underthe SAC phase I project (Uang, 1995). Following that work, additional research on the feasibilityof improving connection performance with welded haunches was conducted under a project thatwas jointly sponsored by NIST and AISC (NIST, 1998). That project was primarily focused onthe problem of upgrading connections in existing buildings. As indicated in the report of thatwork, the haunched modification improves connection performance by altering the basic behaviorof the connection. In essence, the haunch creates a prop type support, beneath the beam bottomflange. This both reduces the effective flexural stresses in the beam at the face of the support, andalso greatly reduces the shear that must be transmitted to the column through the beam. Acomplete procedure for the design of this modification may be found in NIST, 1998.

Figure 7-7 - Bottom Haunch Connection Modification

Figure 7.9.3-1 Bottom Haunch Connecction

Two Nine tests are known to have been performed to date, both successfully all intended toreplicate the condition of an existing connection that has been upgraded. Except for thosespecimens in which existing vulnerable welded joints were left in place at the top flange, theseconnections generally achieved large plastic rotations. Several dynamic tests have also been


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successfully conducted, although only moderate plastic deformation demands could be imposeddue to limitations of the laboratory equipment. Both tests were conducted in arepair/modification configuration. In one test, a portion of the girder top flange, adjacent to thecolumn, was replaced with a thicker plate. In addition, the bottom flange and haunch were bothwelded to the column. This specimen developed a plastic hinge within the beam span, outside thehaunched area and behaved acceptably. A second specimen did not have a thickened top flangeand the bottom girder flange was not welded to the column. Plastic behavior in this specimenoccurred outside the haunch at the bottom flange and adjacent to the column face at the topflange. Failure initiated in the girder at the juncture between the top flange and web, possiblycontributed to by buckling of the flange as well as lateral torsional buckling of the section. Fracture progressed slowly along the top fillet of the girder and eventually, traveled into theflange itself.

Design Issues: The haunch can be attached to the girder in the shop, reducing field erectioncosts. Weld sizes are smaller than in cover plated connections. The top flange is free ofobstructions.

Quantitative Results: No. of specimens tested: 92Girder Size: W30 x 99Column Size: W14 x 176Plastic Rotation achieved-

Specimen 1 UCSD-1R:0.04 radian (w/o bottom flange weld and reinforced top flange)

Specimen 2 UCSD-3R:0.05 radian (with bottom flange weld and reinforced top flange)

Specimen UCSD-4R: 0.014 radian (dynamic- limited by test setup)Specimen UCSD-5R: 0.015 radian (dynamic- limited by test setup)

Girder Size: W36x150Column Size: W14x257Plastic Rotation achieved -

Specimen UCB-RN2: 0.014 radian (no modification of top weld)Specimen UTA-1R: 0.019 radian (partial modification of top weld)Specimen UTA-1RB: 0.028 radian (modified top weld)


Specimen UTA-NSF4: 0.015 radian (no modification of top weld)


Specimen SFCCC-8: 0.035 radian (cover plated top flange)


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Performance is dependent on properly executed complete joint penetration welds at thecolumn face. The joint can be subject to through-thickness flaws in the column flange; however,this connection may not be as sensitive to this potential problem because of the significantincrease in the effective depth of the beam section which can be achieved. Welding of the bottomhaunch requires overhead welding when relatively shallow haunches are used. The skewedgroove welds of the haunch flanges to the girder and column flanges may be difficult to execute. The increased depth of the beam, resulting from the haunch may have undesirable impact onarchitectural design. Unless the top flange is prevented from buckling at the face of the column,performance may not be adequate. For configurations incorporating straight haunches, thehaunch must be long, in order to adequately develop stress into the haunch, through the web. This tends to increase demands at the column face. Additional testing of all these configurationsis recommended.

7.9.4 Top and Bottom Haunch Connections


7.9.5 Side-Plate Connections


7.9.6 Reduced Beam Section Connections

In this connection, the cross section of the beam is intentionally reduced within a segment, toproduce an intended plastic hinge zone or fuse, located within the beam span, away from thecolumn face. Several ways of performing this cross section reduction have been proposed. Onemethod includes removal of a portion of the flanges, symmetrical about the beam centerline, in aso-called “dog bone” profile. Care should be taken with this approach to provide for smoothlycontoured transitions to avoid the creation of stress risers which could initiate fracture. It has alsobeen proposed to create the reduced section of beam by drilling a series of holes in the beamflanges. Figure 7-11 Figure 7.9.6-1 illustrates both concepts. The most successful configurationshave used reduced sections formed with circular cuts. Configurations which taper the reducedsection, through the use of unsymmetrical cut-outs, or variable size holes, to balance the crosssection and the flexural demand have also been tested with success.

Testing of this concept was first performed by a private party, and US patents were appliedfor and granted. These patents have now been released. Limited testing of both “dog-bone” anddrilled hole configurations have been performed in Taiwan (Chen and Yeh - 1995). The AmericanInstitute of Steel Construction is currently performing additional tests of this configuration(Smith-Emery - 1995), however the full results of this testing are not yet available. has performedsuccessful testing of 4 linearly tapered RBS connections. In the time since the first publication ofthe Interim Guidelines, a number of tests have been successfully conducted of RBS connectionswith circular curved cut-outs, including investigations and at the University of Texas at Austin,has successfully tested 4 circular curved RBS specimens. Others, including Popov at the


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University of California at Berkeley, and Texas A&M University., have also tested circular curvedRBS connections with success.

When this connection type was first proposed, There is a concern was expressed that thepresence of a concrete slab at the beam top flange would tend to limit the effectiveness of thereduced section of that flange, particularly when loading places the top flange into compression. It may be possible to mitigate this effect with proper detailing of the slab. Limited testing of RBSspecimens with composite slabs has recently been successfully conducted at Ecole Polytechnic, inMontreal, Canada. In these tests, shear studs were omitted from the portion of the top flangehaving a reduced section, in order to minimize the influence of the slab on flexural hinging. Inaddition, a 1 inch wide gap was placed in the slab, around the column, to reduce the influence ofthe slab on the connection at the column face. More recently, both the University of Texas atAustin and Texas A&M University have conducted successful tests of RBS connections withslabs and without such gaps present between the slab and column. This most recent testingsuggests that the presence of the slab actually enhances connection behavior by retarding bucklingof the top flange in compression and delaying strength degradation effects commonly observed inspecimens tested without slabs.

Design Issues: This connection type is potentially the most economical of the several types whichhave been suggested. The reliability of this connection type is dependent on the quality of thecomplete joint penetration weld of the beam to column flange, and the through-thicknessbehavior of the column flange. If the slab is not appropriately detailed, it may inhibit theintended “fuse” behavior of the reduced section beam segment. It is not clear at this timewhether it would be necessary to use larger beams with this detail to attain the same overallsystem strength and stiffness obtained with other configurations. In limited testing conducted todate of the unsymmetrical “dog-bone” configuration (Smith-Emery - 1995), the plastic hingingwhich occurred at the reduced section was less prone to buckling of the flanges than in some ofthe other configurations which have been tested, due to the very compact nature of the flange inthe region of the plastic hinge. However, the tendency for lateral-torsional buckling issignificantly increased suggesting the need for lateral bracing of the beam flanges, near thereduced section.

Experimental Results: A number of researchers have performed tests on RBS specimens to date.Most tests have utilized the ATC-24 loading protocol, which is similar to the protocol describedin Section 7.4.1 of the Interim Guidelines. Testing employed at Ecole Polytechnic, in Montreal,Canada utilized a series of different testing protocols including the ATC-24 procedure and adynamic excitation simulating the response of a connection in a building to an actual earthquakeaccelerogram (Tremblay, et. al., 1997). This research included two tests of connections withcomposite floor slabs. All of the reported tests with circular flange cuts have performedacceptably, however, the dynamic tests at Ecole Polytechnic only imposed 0.025 radians ofplastic rotation on the assembly.


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ReducedSection

Straight Tapered

Circular

Drilled Constant Drilled Tapered

Figure 7-11 7.9.6-1 - Reduced Beam Section Connection

Quantitative Results: No. of specimens tested: 219 published (without slabs)2Girder Size: W21 x 62W30 x 99 thru W 36 x 194Column Size: W14x120W14 x 176 thru W 14 x 426, W24 x 229Plastic Rotation achieved:- 0.03 radian

Straight: - 0.02 radianTapered - 0.027 - 0.045 radianCircular - 0.03 - 0.04 radian

No. of specimens tested: 42published (with slabs)Girder Size: W21 x 44 to W36 x 150Column Size: W14 x 90 to W14x257Plastic Rotation achieved: 0.03-0.05 radians (ATC-24 loading protocol)

0.025 radians (earthquake simulation – limitedby laboratory setup, no failure observed)

7.9.7 Slip - Friction Energy Dissipating Connection


7.9.8 Column-Tree Connection


7.9.9 Proprietary Slotted Web Connections

In the former prescriptive connection, in which the beam flanges were welded directly to thecolumn flanges, beam flexural stress was transferred into the column web through the combined


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action of direct tension across the column flange, opposite the column web, and through flexureof the column flange. This stress transfer mechanism and its resulting beam flange prying momentresults in a large stress concentration at the center of the beam flange, opposite the column web. Recent research (Allen, et. al. - 1995) indicates that the provision of continuity plates within thecolumn panel zone reduces this stress concentration somewhat, but not completely. The intent ofthe proprietary slotted web connections is to further reduce this stress concentration and toachieve a uniform distribution of flexural stress across the beam flange at the connection, and also,to promote local buckling of the beam flanges under compressive loads to limit the amount ofdemand on the beam flange to column flange weld. Claimed assets for this connection includeelimination of the vertical beam shear in the beam flange welds, elimination of the beam lateraltorsional buckling mode, and the participation of the beam web in resisting its portion of the beam moment. A number of different configurations for this connection type have been developed andtested. Figure 7.9.9-17-14 indicates one such configuration for this connection type that has beensuccessfully tested and which has been used in both new and retrofit steel moment-resistingframes. In this configuration, slots are cut into the beam web, extending from the weld accesshole adjacent to the top and bottom flanges, and extending along the beam axis a sufficient lengthto alleviate the stress concentration effects at the beam flange to column flange weld. The beamweb is welded to the column flange. vertical plates are placed between the column flanges,opposite the edges of the top and bottom beam flanges to stiffen the outstanding column flangesand draw flexural stress away from the center of the beam flange. Horizontal plates are placedbetween these vertical plates and the column web to transfer shear stresses to the panel zone. Theweb itself is softened with the cutting of a vertical slot in the column web, opposite the beamflange. High fidelity finite element models were utilized to confirm that a nearly uniformdistribution of stress occurs across the beam flange.

Slot, typ.

NOTICE OF CONFIDENTIAL INFORMATION:WARNING: The information presented in this figure is PROPRIETARY. US patents have been grantedand Foreign Patents have been applied for. Use of this information is strictly prohibited except asauthorized in writing by the developer. Violators shall be prosecuted in accordance with US and ForeignPatent Intellectual Property Laws.

Figure 7.9.9-17-14 - Proprietary Slotted Web Connection


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Design Issues: This detail is potentially quite economical, entailing somewhat more shopfabrication than the former prescriptive connection, but similar levels of field erection work. Contrary to the recommendations contained in these Interim Guidelines, this connection doesnot shift the location of plastic hinging away from the column face. However, two a number ofconnections employing details similar to that shown in Figure 7-147.9.9-1 have recently beentested successfully (Allen. - 1995). The connection detail is sensitive to the quality of weldingemployed in the critical welds, including those between the beam and column flanges., andbetween the vertical and horizontal plates and the column elements. It has been reported thatone specimen, with a known defect in the beam flange to column flange weld was informallytested and failed at low levels of loading.

The detail is also sensitive to the balance in stiffness of the various plates and flanges. Forconfigurations other than those tested, detailed finite element analyses may be necessary toconfirm that the desired uniform stress distribution is achieved. The developer of this detailindicates that for certain column profiles, it may be possible to omit the vertical slots in thecolumn web and still achieve the desired uniform beam flange stress distribution.

This detail may also be sensitive to the toughness of the column base metal at the region ofthe fillet between the web and flanges. In heavy shapes produced by some rolling processes themetal in this region may have substantially reduced toughness properties relative to the balanceof the section. This condition, coupled with local stress concentrations induced by the slot in theweb may have the potential to initiate premature fracture. The developer believes that it isessential to perform detailed analyses of the connection configuration, in order to avoid suchproblems. Popov tested one specimen incorporating a locally softened web, but without thevertical and horizontal stiffener plates contained in the detail shown in Figure 7-14. Thatspecimen failed by brittle fracture through the column flange which progressed into the holes cutinto the web. The stress patterns induced in that specimen, however, were significantly differentthan those which occur in the detail shown in the figure.

Quantitative Results: Number of specimens tested: 2Girder Size: W 27x94Column Size: W 14x176Plastic Rotation Achieved:

Specimen 1: 0.025 radianSpecimen 2: 0.030 radian

Quantitative data on connection testing may be obtained from thelicensor.

7.9.10 Bolted Bracket Connections

Framing connections employing bolted or riveted brackets have been used in structural steelconstruction since its inception. Early connections of this type were often quite flexible, and alsohad limited strength compared to the members they were connecting, resulting in partiallyrestrained type framing. However, it is possible to construct heavy bolted brackets employing


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high strength bolts to develop fully restrained moment connections. Pretensioing of the bolts orthreaded rods attaching the brackets to the column flanges and use of slip-critical connectionsbetween the brackets and beam flanges can help to provide the rigidity required to obtain fullyrestrained behavior. Reinforcement of the column flanges may be required to prevent localyielding and excessive deformation of these elements, as well. Two alternative configurations thathave been tested recently are illustrated in Figure 7.9.10-1. The developer of these configurationsoffers the brackets in the form of proprietary steel castings. Several tests of these alternativeconnections have been performed on specimens with beams ranging in size from W16 to W36sections and with large plastic rotations successfully achieved.

Design Issues: The concept of bolted bracket connections is similar to that of the riveted “windconnections” commonly installed in steel frame buildings in the early twentieth century. Theprimary difference is that the riveted wind connections were typically limited in strength eitherby flexural yielding of outstanding flanges of the brackets, or shear and tension on the rivets,rather than by flexural hinging of the connected framing. Since the old-style wind connectionscould not typically develop the flexural strength of the girders and also could be quite flexible,they would be classified either as partial strength or partially restrained connections. Followingthe Northridge earthquake, the concept of designing such connections with high strength boltsand heavy plates, to behave as fully restrained connections, was developed and tested by aprivate party who has applied for patents on the concept of using steel castings for this purpose.

PipePlate

High tensilethreaded rod

Bolts

Bracket

WARNING: The information presented in this figure is PROPRIETARY. US and ForeignPatents have been applied for. Use of this information is strictly prohibited except as authorizedin writing by the developer. Violators shall be prosecuted in accordance with US and ForeignPatent Intellectual Property Laws.

Figure 7.9.10-1 Bolted Bracket Connections

Bolted connections offer a number of potential advantages over welded connections. Since nofield welding is required for these connections, they are inherently less labor intensive during


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erection, and also less dependent on the technique of individual welders for successfulperformance. However, quality assurance should be provided for installation and tensioning ofthe bolts, as well as correction of any problems with fit-up due to fabrication tolerances.

Experimental Results: A series of tests on several different configurations of proprietary heavybolted bracket connections have been performed at Lehigh University (Ksai & Bleiman, 1996) toqualify these connections for use in repair and modification applications. To test repairapplications, brackets were placed only on the bottom beam flange to simulate installations on aconnection where the bottom flange weld in the original connection had failed. In thesespecimens, bottom flange welds were not installed, to approximate the condition of a fullyfractured weld. The top flange welds of these specimens were made with electrodes rated fornotch toughness, to preclude premature failure of the specimens at the top flange. Forspecimens in which brackets were placed at both the top and bottom beam flanges, both weldswere omitted. Acceptable plastic rotations were achieved for each of the specimens tested.

Quantitative Results: No. of specimens tested: 8Girder Size: W16x40 and W36x150Column Size: W12x65 and W14x425Plastic Rotation achieved - 0.05 radians - 0.07 radians

7.10 Other Types of Welded Connection Structures


7.10.1 Eccentrically Braced Frames (EBF)


7.10.2 Dual Systems


7.10.3 Welded Base Plate Details


7.10.4 Vierendeel Truss Systems


7.10.5 Moment Frame Tubular Systems



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7.10.6 Welded Connections of Collectors, Ties and Diaphragm Chords


7.10.7 Welded Column Splices


7.10.8 Built-up Moment Frame Members



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Metallurgy & Welding

8-1

8. METALLURGY & WELDING

8.1 Parent Materials

8.1.1 Steels

Designers should specify materials which are readily available for building construction and whichwill provide suitable ductility and weldability for seismic applications. Structural steels which may beused in the lateral-force-resisting systems for structures designed for seismic resistance without specialqualification include those contained in Table 8.1.1-1. Refer to the applicable ASTM referencestandard for detailed information.

Table 8.1.1-1 - Structural Steel Prequalified for Use in Seismic Lateral-Force-Resisting Systems

ASTM Specification DescriptionASTM A36 Carbon Structural SteelASTM A283Grade D

Low and Intermediate Tensile Strength Carbon Steel Plates

ASTM A500(Grades B & C)

Cold-Formed Welded & Seamless Carbon Steel Structural Tubing in Rounds &Shapes

ASTM A501 Hot-Formed Welded & Seamless Carbon Steel Structural TubingASTM A572(Grades 42 & 50)

High-Strength Low-Alloy Columbium-Vanadium Steels of Structural Quality

ASTM A588 High-Strength Low-Alloy Structural Steel (weathering steel)ASTM A9921 Steel for Structural Shapes for Use in Building FramingNotes:1- See Commentary

Structural steels which may be used in the lateral-force-resisting systems of structures designed forseismic resistance with special permission of the building official are those listed in Table 8.1.1-2. Steelmeeting these specifications has not been demonstrated to have adequate weldability or ductility forgeneral purpose application in seismic-force-resisting systems, although it may well possess suchcharacteristics. In order to demonstrate the acceptability of these materials for such use in WSMFconstruction it is recommended that connections be qualified by test, in accordance with the guidelinesof Chapter 7. The test specimens should be fabricated out of the steel using those welding proceduresproposed for use in the actual work.

Table 8.1.1-2 - Non-prequalified Structural Steel

ASTM Specification DescriptionASTM A242 High-Strength Low-Alloy Structural SteelASTM A709 Structural Steel for BridgesASTM A913 High-Strength Low-Alloy Steel Shapes of Structural Quality, Produced by

Quenching & Self-Tempering Process



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Commentary: Many WSMF structures designed in the last 10 years incorporatedASTM A36 steel for the beams and ASTM A572 grade 50 steel for the columns. This provided an economical way to design structures for the strong column -weak beam provisions contained in the building code. Recent studies conductedby the Structural Shape Producers Council (SSPC), however, indicate thatmaterial produced to the A36 specification has wide variation in strengthproperties with actual yield strengths that often exceed 50 ksi. This widevariation makes prediction of connection and frame behavior difficult. Somehave postulated that one of the contributing causes to damage experienced in theNorthridge earthquake was inadvertent pairing of overly strong beams withaverage strength columns.

The AISC and SSPC have been working for several years to develop a newspecification for structural steel that would have both minimum and maximumyield values defined and provide for a margin between maximum yield andminimum ultimate tensile stress. AISC recently submitted such a specification,for a material with 50 ksi specified yield strength, to ASTM for development intoa standard specification. ASTM formally adopted the new specification forstructural shapes, with a yield strength of 50 ksi, under designation A992 in 1998and It is anticipated that domestic mills will begin have begun producingstructural wide flange shapes to this specification. within a few years and thateventually, this new material will replace A36 as the standard structural materialfor incorporation into lateral-force-resisting systems.

Since the formal approval of the A992 specification by ASTM occurred afterpublication of the 1997 editions of the building codes and the AISC SeismicSpecification, it is not listed in any of these documents as a prequalified materialfor use in lateral force resisting systems. Neither is it listed as prequalified inAWS D1.1-98. However, all steel that complies with the ASTM-992 specificationwill also meet the requirements of ASTM A572, Grade 50 and should therefore bepermissible for any application for which the A572 material is approved. Seealso, the commentary to Section 8.2.2.

Under certain circumstances it may be desirable to specify steels that are notrecognized under the UBC for use in lateral-force-resisting systems. Forinstance, ASTM A709 might be specified if the designer wanted to place limits ontoughness for fracture-critical applications. In addition, designers may wish tobegin incorporating ASTM A913, Grade 65 steel, as well as other higher strengthmaterials, into projects, in order to again be able to economically design forstrong column - weak beam conditions. Designers should be aware, however, thatthese alternative steel materials may not be readily available. It is alsoimportant when using such non-prequalified steel materials, that precautions betaken to ensure adequate weldability of the material and that it has sufficientductility to perform under the severe loadings produced by earthquakes. The



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cyclic test program recommended by these Interim Guidelines for qualification ofconnection designs, by test, is believed to be an adequate approach to qualifyalternative steel material for such use as well.

Note that ASTM A709 steel, although not listed in the building code asprequalified for use in lateral-force-resisting systems, actually meets all of therequirements for ASTM A36 and ASTM A572. Consequently, specialqualification of the use of this steel should not be required.

Although the 1994 editions of the Uniform Building Code and the NEHRPProvisions do not prequalify the use of ASTM A913 steel in lateral force resistingsystems, the pending 1997 edition of the UBC does prequalify its use. Both the1997 NEHRP Provisions and the AISC Seismic Provisions prequalify the use ofthis steel in elements that do not undergo significant yielding, for example, thecolumns of moment-resisting frames designed to meet strong column - weak beamcriteria. Consequently, special approval of the Building Official should nolonger be required as a pre-condition of the use of material conforming to thisspecification, at least for columns.

8.1.2 Chemistry


Commentary: Some concern has been expressed with respect to the movement inthe steel producing industry of utilizing more recycled steel in its processes. Thisresults in added trace elements not limited by current specifications. Althoughthese have not been shown quantitatively to be detrimental to the performance ofwelding on the above steels, a the new A992specification for structural steelproposed by AISC does place more control on these trace elements. Mill testreports now include elements not limited in some or all of the specifications. They include copper, columbium, chromium, nickel, molybdenum, silicon andvanadium. The analysis and reporting of an expanded set of elements should bepossible, and could be beneficial in the preparation of welding procedurespecifications (WPSs) by the welding engineer if critical welding parameters arerequired. Modern spectrographs used by the mills are capable of automatedanalyses. When required by the engineer, a request for special supplementalrequests should be noted in the contract documents.

8.1.3 Tensile/Elongation Properties

Mechanical property test specimens are taken from rolled shapes or plates at the rolling mill in themanner and location prescribed by ASTM A6 and ASTM A370. Table 8-3 Table 8.1.3-1 gives thebasic mechanical requirements for commonly used structural steels. Properties specified, andcontrolled by the mills, in current practice include minimum yield strength or yield point, ultimate



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tensile strength and minimum elongation. However, there can be considerable variability in the actualproperties of steel meeting these specifications.

SSPC, in cooperation with SEAOC, has collected statistical data on the strength characteristics oftwo grades (ASTM A36 and ASTM A572 Grade 50) of structural steels, based on mill test reportsfrom selected domestic producers for the 1992 production year. Data were also collected for "DualGrade" material that was certified by the producers as complying with both ASTM A36 and ASTMA572 Grade 50. Table 8-4 Table 8.1.3-2 summarizes these results as well as data provided by a singleproducer for ASTM A913 material.

Unless special precautions are taken to limit the actual strength of material incorporated into thework to defined levels, new material specified as ASTM A36 should be assumed to be the dual gradefor connection demand calculations, whenever the assumption of a higher strength will result in a moreconservative design condition.

Table 8-3 Table 8.1.3-1 - Typical Tensile Requirements for Structural Shapes

ASTMMinimum YieldStrength or Yield

Point, Ksi

Ultimate TensileStrength, Ksi

Minimum Elongation%

in 2 inches

Minimum Elongation%

in 8 inchesA36 36 Min. 58-801 212 20A242 424 Min.. 63 MIN. 213 18

A572, Gr. 42 42 Min. 60 Min. 24 20A572, GR50 50 Min. 65 MIN. 212 18

A588 50 Min. 70 MIN. 213 18A709, GR36 36 Min. 58-80 212 20A709, GR50 50 Min. 65 MIN. 21 18A913, GR50 50 Min. 65 MIN. 21 18A913, GR65 65 Min. 80 MIN. 17 15

A992 50 Min. – 65 Max. 65 MIN 21 18Notes: 1- No maximum for shapes greater than 426 lb./ft.

2- Minimum is 19% for shapes greater than 426 lb. /ft.3- No limit for Shape Groups 1, 2 and 3.Minimum is 18% for shapes greater than 426 lb./ft.4. Minimum is 50 ksi for Shape Groups 1 and 2, 46 ksi for Shape Group 3, and 42 ksi for Shape Groups 4

and 5.



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Table 8-4 Table 8.1.3-2 - Statistics for Structural Shapes1,2

Statistic A 36 DualGRADE

A572GR50

A913GR65

Yield Point (ksi) Mean 49.2 55.2 57.6 75.3 Minimum 36.0 50.0 50.0 68.2 Maximum 72.4 71.1 79.5 84.1 Standard Deviation [ s ] 4.9 3.7 5.1 4.0 Mean + 1 s 54.1 58.9 62.7 79.3

Tensile Strength (ksi) Mean 68.5 73.2 75.6 89.7 Minimum 58.0 65.0 65.0 83.4 Maximum 88.5 80.0 104.0 99.6 Standard Deviation [ s ] 4.6 3.3 6.2 3.5 Mean + 1 s 73.1 76.5 81.8 93.2

Yield/Tensile Ratio Mean 0.72 0.75 0.76 0.84 Minimum 0.51 0.65 0.62 0.75 Maximum 0.93 0.92 0.95 0.90 Standard Deviation [ s ] 0.06 0.04 0.05 0.03 Mean + 1 s 0.78 0.79 0.81 0.87 Mean - 1 s 0.66 0.71 0.71 0.81

1: The data presented for ASTM A36, “Dual Grade” and ASTM A572 Grade 50 were included aspart of the SSPC study (SSPC-1994). The data for ASTM A913 were derived from a singleproducer and may not be available from all producers.

2. Statistical Data on the distribution of strength properties for material meeting ASTM A992 are notpresently available. Pending the development of such statistics, it should be assumed that A992material will have similar properties to ASTM A572, Gr. 50 material.

Commentary: The data given in Table 8-4 Table 8.1.3-2 for A36 and A572Grade 50 is somewhat weighted by the lighter, Group 1 shapes that will notordinarily be used in WSMF applications. Excluding Group 1 shapes andcombining the Dual Grade and A572 Grade 50 data results in a mean yieldstrength of 48 ksi for A36 and 57 ksi for A572 Grade 50 steel. It should also benoted that approximately 50% of the material actually incorporated in a projectwill have yield strengths that exceed these mean values. For the design offacilities with stringent requirements for limiting post-earthquake damage,consideration of more conservative estimates of the actual yield strength may bewarranted.

Until recently, In wide flange sections the tensile test coupons in wide flangesections are currently were taken from the web. The amount of reduction rolling,finish rolling temperatures and cooling conditions affect the tensile and impact



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properties in different areas of the member. Typically, the web exhibits about fivepercent higher strength than the flanges due to faster cooling. In 1998 ASTM A6was revised to specify that coupons be taken from the flange of wide flangeshapes.

Design professionals should be aware of the variation in actual propertiespermitted by the ASTM specifications. This is especially important for yieldstrength. Yield strengths for ASTM A36 material have consistently increased overthe last 15 years so that several grades of steel may have the same properties orreversed properties, with respect to beams and columns, from those the designerintended. Investigations of structures damaged by the Northridge earthquakefound some WSMF connections in which beam yield strength exceeded columnyield strength despite the opposite intent of the designer.

As an example of the variations which can be found, Table 8-5 Table 8.1.3-2presents the variation in material properties found within a single buildingaffected by the Northridge earthquake. Properties shown include measured yieldstrength (Fya,), measured tensile strength (Fua ) and Charpy V-Notch energy rating(CVN).

Table 8-5Table 8.1.3-2 - Sample Steel Properties from a Building Affected by the NorthridgeEarthquake

Shape Fya1 ksi Fua, ksi CVN, ft-lb.

W36 X 182 38.0 69.3 18

W36 X 230 49.3 71.7 195Note 1 - ASTM A36 material was specified for both structures.

The practice of dual certification of A36 and A572, Grade 50 can result inmean yield strengths that are fifty percent higher than the specified yield of A36. Since there is no practical way to discern whether dual grade steel will besupplied, unless direct purchase of steel from specific suppliers is made, in theabsence of such procurement practices, the prudent action for determiningconnection requirements, where higher strengths could be detrimental to thedesign, would be to assume the dual grade material whenever A36 or A572 Grade50 is specified.

In the period since the initial publication of the Interim Guidelines, severalresearchers and engineers engaged in connection assembly prototype testing havereported that tensile tests on coupons extracted from steel members used in theprototype tests resulted in lower yield strength than reported on the mill testreport furnished with the material, and in a few cases lower yield strength thanwould be permitted by the applicable ASTM specification. This led to someconfusion and concern, as to how mill test reports should be interpreted.



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The variation of the measured yield strength of coupons reported byresearchers engaged in connection prototype testing, as compared to thatindicated on the mill test reports, is not unusual and should be expected. Thesevariations are the result of a series of factors including inconsistencies betweenthe testing procedures employed as well as normal variation in the material itself. The following paragraphs describe the basis for the strengths reported byproducers on mill certificates, as well as the factors that could cause independentinvestigators to determine different strengths for the same material.

Mill tests of mechanical properties of steel are performed in accordance withthe requirements of ASTM specifications A6 and A370. ASTM A6 hadhistorically required that test specimens for rolled W shapes be taken from thewebs of the shapes, but recently was revised to require testing from the flanges ofwide flange shapes with 6 inch or wider flanges. A minimum of two tests must bemade for each heat of steel, although additional tests are required if shapes ofsignificantly different thickness are cast from the same heat. Coupon size andshape is specified based on the thickness of the material. The size of the couponused to test material strength can effect the indicated value. Under ASTM A6,material that is between 3/4 inches thick and 4 inches thick can either be tested infull thickness “straps” or in smaller 1/2” diameter round specimens. In thickmaterial, the yield strength will vary through the thickness, as a result of coolingrate effects. The material at the core of the section cools most slowly, has largergrain size and consequently lower strength. If full-thickness specimens are used,as is the practice in most mills, the recorded yield strength will be an average ofthe relatively stronger material at the edges of the thickness and the lower yieldmaterial at the center. Many independent laboratories will use the smaller 1/2”round specimens, and sometimes even sub-sized 1/4” round specimens for tensiletesting, due to limitations of their testing equipment. Use of these smallerspecimens for thick material will result in testing only of the lower yield strengthmaterial at the center of the thickness.

ASTM A370 specifies the actual protocol for tensile testing including theloading rate and method of reporting test data. Strain rate can affect the strengthand elongation values obtained for material. High strain rates result in elevatedstrength and reduced ductility. Under ASTM A370, yield values may bedetermined using any convenient strain rate, but not more than 1/16 inch perinch, per minute which corresponds to a maximum loading rate of approximately30 ksi per second. Once the yield value is determined, continued testing to obtainultimate tensile values can proceed at a more rapid rate, not to exceed 1/2 inchper inch per minute.

Under ASTM A370, there are two different ways in which the yield propertyfor structural steel can be measured and reported. These include yield point andyield strength. These are illustrated in Figure 8.1.3-1. The yield point is the peak



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stress that occurs at the limit of the elastic range, while the yield strength is asomewhat lower value, typically measured at a specified offset or elongationunder load. Although a number of methods are available to determine yieldpoint, the so-called “drop of the beam” method is most commonly used forstructural steel. In this method the load at which a momentary drop-off inapplied loading occurs is recorded, and then converted to units of stress to obtainthe yield point. Yield strength may also be determined by several methods, but ismost commonly determined using the offset method. In this method, the stress -strain diagram for the test is drawn, as indicated in Figure 8.1.3-1. A specifiedoffset, typically 0.2% strain for structural steel, is laid off on the abscissa of thecurve and a line is drawn from this offset, parallel to the slope of the elasticportion of the test. The stress at the intersection of this offset line with the stress-strain curve is taken as the yield strength.

ε

σYield Point

Yield Strength

Offset

Figure 8.1.3-1 Typical Stress - Strain Curve for Structural Steel

The material specifications for structural steels typically specify minimumvalues for yield point but do not control yield strength. The SSPC has reportedthat actual practice among the mills varies, with some mills reporting yieldstrength and others reporting yield point. This practice is permissible as yieldstrength will always be a somewhat lower value than yield point, resulting in asomewhat conservative demonstration that the material meets specifiedrequirements. However, this does mean that there is inconsistency between thevalues reported by the various mills on certification reports. Similarly, theprocedures followed by independent testing laboratories may be different thanthose followed by the mill, particularly with regard to strain rate and the locationat which a coupon is obtained.



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Under ASTM A6, coupons for tensile tests had historically been obtained fromthe webs of structural shapes. However, most engineers and researchers engagedin connection testing have preferred to extract material specimens from theflanges of the shape, since this is more representative of the flexural strength ofthe section. Coupons removed from the web of a rolled shape tend to exhibitsomewhat higher strength properties than do coupons removed from the flanges,due to the extra amount of working the thinner web material typically experiencesduring the rolling process and also because the thinner material cools morerapidly after rolling, resulting in finer grain size. Given these differences intesting practice, as well as the normal variation that can occur along the lengthof an individual member and between different members rolled from the sameheat, the reported differences in strength obtained by independent laboratories,as compared to that reported on the mill test reports, should not be surprising. Itis worth noting that following the recognition of these differences in testingprocedure, the SSPC in coordination with AISC and ASTM developed andproposed a revision to the A6 specification to require test specimens to be takenfrom the flanges of rolled shapes when the flanges are 6 inches or more wide. Itis anticipated that mills will begin to alter practice to conform to a revisedspecification in early 1997 This has since become the standard practice.

The discovery of the somewhat varied practice for reporting material strengthcalls into question both the validity of statistics on the yield strength of structuralsteel obtained from the SSPC study, and its relevance to the determination of theexpected strength of the material for use in design calculations. Although theyield point is the quantity controlled by the ASTM material specifications, it haslittle relevance to the plastic moment capacity of a beam section. Plastic sectioncapacity is more closely related to the stress along the lower yield plateau of thetypical stress-strain curve for structural steel. This strength may often besomewhat lower than that determined by the offset drop-of-the-beam method. Since the database of material test reports on which the SSPC study was basedappears to contain test data based on both the offset and drop-of-the-beammethods, it is difficult to place great significance in the statistics derived from itand to draw a direct parallel between this data and the expected flexural strengthof rolled shapes. It would appear that the statistics reported in the SSPC studyprovide estimates of the probable material strength that are somewhat high. Thus, the recommended design strengths presented in Tables 6.6.6.3-1 and 7.5.1-1 of the Interim Guidelines would appear to be conservative with regard to designof welds, panel zones and other elements with demands limited by the beam yield strength.

Under the phase II program of investigation, SAC, together with the shapeproducers, is engaged in additional study of the statistical distribution of yieldstrength of various materials produced by the mills. This study is intended toprovide an improved understanding of the statistical distribution of the lower



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yield plateau strength of material extracted from section flanges, measured in aconsistent manner. In addition, it will provide correlation with yield strengthsdetermined by other methods such that the data provided on mill test certificatescan be properly interpreted and utilized. In addition, the possibility of revisingthe ASTM specifications to provide for more consistent reporting of strength dataas well as the reporting of strength statistics that are directly useful in the designprocess will be evaluated. In the interim period, the data reported in Table 8-1.3-2, extracted from the SSPC study, remain the best currently availableinformation.

8.1.4 Toughness Properties


8.1.5 Lamellar Discontinuities


8.1.6 K-Area Fractures

Recently, there have beenIn the period 1995-96 there were several reports of fractures initiating inthe webs of column sections during the fabrication process, as flange continuity plates and/or doublerplates were welded into the sections. This fracturing typically initiated in the region near the filletbetween the flange and web. This region has been commonly termed the “k-area” because the AISCManual of Steel Construction indicates the dimension of the fillet between the web and flange with thesymbol “k”. The k-area may be considered to extend from mid-point of the radius of the fillet into theweb, approximately 1 to 1-1/2 inches beyond the point of tangency between the fillet and web. Thefractures typically extended into, and sometimes across, the webs of the columns in a characteristic“half-moon” or “smiley face” pattern.

Investigations of materials extracted from fractured members have indicated that the material in thisregion of the shapes had elevated yield strength, high yield/tensile ratio, high hardness and very lowtoughness, on the order of a few foot-pounds at 70oF. Material with these properties can behave in abrittle manner. Fracture can be induced by thermal stresses from the welding process or by subsequentweld shrinkage, as apparently occurred in the reported cases. There have been no reported cases of in-service k-line fracture from externally applied loading, as in beam-column connections, although such apossibility is perceived to exist under large inelastic demand.

It appears that this local embrittling of sections can be attributed to the rotary straightening processused by some mills to bring the rolled shapes within the permissible tolerances under ASTM A6. Thestraightening process results in local cold working of the sections, which strain hardens the material. The amount of cold working that occurs depends on the initial straightness of the section andconsequently, the extent that mechanical properties are effected is likely to vary along the length of amember. The actual process used to straighten the section can also affect the amount of local coldworking that occurs.



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Engineers can reduce the potential for weld-induced fracture in the k-area by avoiding weldingwithin the k-area region. This can be accomplished by detailing doubler plates and continuity platessuch that they do not contact the section in this region. The use of large corner clips on beam flangecontinuity plates can permit this. Selection of column sections with thicker webs, to eliminate the needfor doubler plates; the use of fillet welds rather than full penetration groove welds to attach doublerplates to columns, when acceptable for stress transfer; and detailing of column web doubler plates suchthat they are offset from the face of the column web can also help to avoid these fabrication-inducedfracture problems.

Commentary: It appears that detailing and fabrication practice can be adjustedto reduce the potential for k-area fracture during fabrication. However, theacceptability of having low-toughness material in the k-area region for service isa question that remains. It is not clear at this time what percentage of thematerial incorporated in projects is adversely affected, or even if a problem withregard to serviceability exists. SAC recently placed a public call, asking forreports of fabrication-induced fractures at the k-area, but only received limitedresponse. However, in one of the projects that did report this problem, asignificant number of columns were affected. This may have been contributed toby the detailing and fabrication practices applied on that project.

Other than detailing structures to minimize the use of doubler plates, and toavoid large weldments in the potentially sensitive k-area of the shape, it is notclear at this time, what approach, if any, engineers should take with regard to thisissue. There are several methods available to identify possible low notchtoughness in structural carbon steels, including Charpy V-Notch testing andhardness testing of samples extracted from the members. However, both of theseapproaches are quite costly for application as a routine measure on projects andthe need for such measures has not yet been established.

Following publication of advisories on the k-line problem by AISC, and thepublication of similar advisory information in FEMA-267a,reports on thisproblem diminished. It is not clear whether this is due to revised detailingpractice on the part of engineers and fabricators, revised mill rolling practice, ora combination of both. SAC, AISC and SSPC are continuing to research thisissue in order to identify if a significant problem exists, and if it does, todetermine its basic causes, and to develop appropriate recommendations for mill,design, detailing, and fabrication practices to mitigate the problem.

8.2 Welding

8.2.1 Welding Process




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8.2.2 Welding Procedures

Welding should be performed within the parameters established by the electrode manufacturer andthe Welding Procedure Specification (WPS), required under AWS D1.1.

Commentary: A welding procedure specification identifies all the importantparameters for making a welded joint including the material specifications of thebase and filler metals, joint geometry, welding process, requirements for pre- andpost-weld heat treatment, welding position, electrical characteristics, voltage,amperage, and travel speed. Two types of welding procedure specifications arerecognized by AWS D1.1. These are prequalified procedures and qualified-by-test procedures. Prequalified procedures are those for which the importantparameters are specified within the D1.1 specification. If a prequalifiedprocedure is to be used for a joint, all of the variables for the joint must fallwithin the limits indicated in the D1.1 specification for the specific procedure. Ifone or more variables are outside the limits specified for the prequalifiedprocedures, then the fabricator must demonstrate the adequacy of the proposedprocedure through a series of tests and submit documentation (procedurequalification records) demonstrating that acceptable properties were obtained. Regardless of whether or not a prequalified or qualified-by-test procedure isemployed, the fabricator should prepare a welding procedure specification, whichshould be submitted to the engineer of record for review and be maintained at thework location for reference by the welders and inspectors. The followinginformation is presented to help the engineer understand some of the issuessurrounding the parameters controlled by the welding procedure specification.

For example, the position (if applicable), electrode diameter, amperage orwire feed speed range, voltage range, travel speed range and electrode stickout(e.g. all passes, 0.072 in. diameter, 248 to 302 amps, 19 to 23 volts, 6 to 10inches/minute travel speed, 170 to 245 inches/minute wire feed speed, 1/2" to 1"electrode stickout) should be established. This information is generally submittedby the fabricator as part of the Welding Procedure Specification. Its importancein producing a high quality weld is essential. The following information ispresented to help the engineer understand some of the issues surrounding theseparameters.

The amperage, voltage, travel speed, electrical stickout and wire feed speedare functions of each electrode. If prequalified WPSs are utilized, theseparameters must be in compliance with the AWS D1.1 requirements. For FCAWand SMAW, the parameters required for an individual electrode vary frommanufacturer to manufacturer. Therefore, for these processes, it is essential thatthe fabricator/erector utilize parameters that are within the range ofrecommended operation published by the filler metal manufacturer. Alternately,the fabricator/erector could qualify the welding procedure by test in accordance



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with the provisions of AWS D1.1 and base the WPS parameters on the test results.For submerged arc welding, the AWS D1.1 code provides specific amperagelimitations since the solid steel electrodes used by this process operate essentiallythe same regardless of manufacture. The filler metal manufacturer’s guidelineshould supply data on amperage or wire feed speed, voltage, polarity, andelectrical stickout. The guidelines will not, however, include information ontravel speed which is a function of the joint detail. The contractor should select abalanced combination of parameters, including travel speed, that will ensure thatthe code mandated weld-bead sizes (width and height) are not exceeded.

Recently, ASTM approved a new material specification for structural steelshape, ASTM A992. This specification is very similar to the ASTM A572, Grade50 specification except that it includes additional limitations on yield and tensilestrengths and chemical composition. Although material conforming to A992 isexpected to have very similar welding characteristics to A572 material, it wasadopted too late to be included as a prequalified base material in AWS D1.1-98. Although the D1 committee has evaluated A992 and has taken measures toincorporate it as a prequalified material in AWS D1.1-2000, technically, underAWS D1.1-98, welded joints made with this material should follow qualified-by-test procedures.

In reality, structural steel conforming to ASTM A992 may actually havesomewhat better weldability than material conforming to the A572 specification.This is because A992 includes limits on carbon equivalent, precluding thedelivery of steels where all alloys simultaneously approach the maximumspecified limits. Therefore, it should be permissible to utilize prequalifiedprocedures for joint with base metal conforming to this specification.

8.2.3 Welding Filler Metals


Commentary: Currently, there are no notch toughness requirements for weldmetal used in welding ASTM A 36 or A 572, Grade 50, steel in AWS D1.1. Thistopic has been extensively discussed by the Welding Group at the JointSAC/AISC/AISI/NIST Invitational Workshop on September 8 and 9, 1994, and byall participants of the SAC Invitational Workshop on October 28 and 29, 1994.The topic was also considered by the AWS Presidential Task Group, whichdecided that additional research was required to determine the need fortoughness in weld metal. There is general agreement that adding a toughnessrequirement for filler metal would be desirable and easily achievable. Most fillermetals are fairly tough, but some will not achieve even a modest requirement suchas 5 ft-lb. at + 70? F. What is not in unanimous agreement is what level oftoughness should be required. The recommendation from the Joint Workshop was



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20 ft-lb. at -20? F per Charpy V-Notch [CVN] testing. The recommendationfrom the SAC Workshop was 20 ft-lb. at 30? F lower than the Lowest AmbientService Temperature (LAST) and not above 0? F. The AWS Presidential TaskGroup provided an interim recommendation for different toughness valuesdepending on the climatic zone, referenced to ASTM A709. Specifically, therecommendation was for 20 ft-lb. at temperatures of 70 degrees F for Zone 1, 40degrees F for Zone 2, and 10 degrees F for Zone 3. The AWS also suggestedtoughness values for base metals used in these applications.

Some fractured surfaces in the Northridge and Kobe Earthquakes revealedevidence of improper use of electrodes and welding procedures. Prominentamong the misuses were high production deposition rates. Pass widths of up to 1-1/2 inches and pass heights of 1/2 inch were common. The kind of heat inputassociated with such large passes promotes grain growth in the HAZ andattendant low notch toughness. In evaluation of welds in buildings affected by theNorthridge earthquake, the parameters found to be most likely to result indamage-susceptible welds included root gap, access capability, electrodediameter, stick-out, pass thickness, pass width, travel speed, wire feed rate,current and voltage were found to be the significant problems in evaluation ofwelds in buildings affected by the Northridge earthquake.

Welding electrodes for common welding processes include:

AWS A5.20: Carbon Steel Electrodes for FCAWAWS A5.29: Low Alloy Steel Electrodes for FCAWAWS A5.1: Carbon Steel Electrodes for SMAWAWS A5.5: Low Alloy Steel Covered Arc Welding Electrodes (for SMAW)AWS A5.17: Carbon Steel Electrodes and Fluxes for SAWAWS A5.23: Low Alloy Steel Electrodes and Fluxes for SAWAWS A5.25: Carbon and Low Alloy Steel Electrodes and Fluxes for Electroslag

Welding

In flux cored arc welding, one would expect the use of electrodes that meeteither AWS A5.20 or AWS A5.29 provided they meet the toughness requirementsspecified below.

Except to the extent that one requires Charpy V-Notch toughness andminimum yield strength, the filler metal classification is typically selected by theFabricator. Compatibility between different filler metals must be confirmed bythe Fabricator, particularly when SMAW and FCAW-SS processes are mixed. Generally speaking, SMAW-type filler metals may not be applied to FCAW-SStype filler metals (e.g. when a weld has been partially removed) while FCAW-typefiller metals may be applied to SMAW-type filler metals. This recommendationconsiders the use of aluminum as a killing agent in FCAW-SS electrodes that can



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be incorporated into the SMAW filler metal with a reduction in impact toughnessproperties.

As an aid to the engineer, the following interpretation of filler metalclassifications is provided below:

E1X2X3T4X5 For electrodes specified under AWS A5.20 (e.g. E71T1)E1X2X3T4X5X6 For electrodes specified under AWS A5.29 (e.g. E70TGK2)E1XX7X8X9X10 For electrodes specified under AWS A5.1 or AWS A5.5. (e.g. E7018)

NOTES:

1. Indicates an electrode.

2. Indicates minimum tensile strength of deposited weld metal (in tens of ksi, e.g., 7 = 70ksi).

3. Indicates primary welding position for which the electrode is designed (0 = flat andhorizontal and 1 = all positions).

4. Indicates a flux cored electrode. Absence of a letter indicates a "stick" electrode forSMAW.

5. Describes usability and performance capabilities. For our purposes, it conveys whetheror not Charpy V-Notch toughness is required (1, 5, 6 and 8 have impact strengthrequirements while 2, 4, 7, 10 and 11 do not). A "G" signifies that the properties are notdefined by AWS and are to be agreed upon between the manufacturer and the specifier. Impact strength is specified in terms of the number of foot-pounds at a given temperature(e.g., 20 ft-lb. at 0 degrees F). Note that for electrodes specified under AWS A5.20, theformat for usage is "T-X".

6. Designates the chemical composition of deposited metal for electrodes specified underAWS A5.29. Note that there is no equivalent format for chemical composition forelectrodes specified under AWS A5.20.

7. The first two digits (or three digits in a five digit number) designate the minimum tensilestrength in ksi.

8. The third digit (or fourth digit in a five digit number) indicates the primary weldingposition for which the electrode is designed (1 = all positions, 2 = flat position and filletwelds in the horizontal position, 4 = vertical welding with downward progression and forother positions.)

9. The last two digits, taken together, indicate the type of current with which the electrodecan be used and the type of covering on the electrode.



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10. Indicates a suffix (e.g., A1, A2, B1, etc.) designating the chemical composition of thedeposited metal.

Electrode Diameter: (See AWS D1.1 Section 4.14.1.2) The issue of maximumelectrode diameter has not been studied sufficiently to determine whether or notelectrode diameter is a critical variable. Recent tests have produced modifiedframe joints with acceptable test results using the previous standard-of-practice 0.120 in. diameter wire. The use of smaller diameter electrodes will slow the rateof deposition (as measured by volume) but will not, in and of itself, produce anacceptable weld. The following lists the maximum allowable electrode diametersfor prequalified FCAW WPS’s according to D1.1:

• Horizontal, complete or partial penetration welds: 1/8 inch (0.125")*• Vertical, complete or partial penetration welds: 5/64 inch (0.078")• Horizontal, fillet welds: 1/8 inch (0.125")• Vertical, fillet welds: 5/64 inch (0.078")• Overhead, reinforcing fillet welds: 5/64 inch (0.078")

* This value is not part of D1.1-94, but will be part of D1.1-96.

For a given electrode diameter, there is an optimum range of weld bead sizesthat may be deposited. Weld bead sizes that are outside the acceptable size range(either too large or too small) may result in unacceptable weld quality. The D1.1code controls both maximum electrode diameters and maximum bead sizes (widthand thickness). Prequalified WPS’s are required to meet these coderequirements. Further restrictions on suitable electrode diameters are notrecommended.

Low-hydrogen electrodes. Low hydrogen electrodes should be used to minimizethe risk of hydrogen assisted cracking (HAC) when conditions of high restraintand the potential for high hardness microstructures exist. Hydrogen assistedcracking can occur in the heat affected zone or weld metal whenever sufficientconcentrations of diffusible hydrogen and sufficient stresses are present alongwith a hard microstructure at a temperature between 100 C and –100 C. Hydrogen is soluble in steel at high temperatures and is introduced into the weldpool from a variety of sources including but not limited to: moisture from coatingor core ingredients, drawing lubricants, hydrogenous compounds on the basematerial, and moisture from the atmosphere.

At the present time, the term “low hydrogen” is not well defined by AWS. Thedegree of hydrogen control required to reduce the risk of hydrogen assistedcracking will depend on the material being welded, level of restraint,preheat/interpass temperature, and heat input level. When a controlled level ofdiffusible hydrogen is required, electrodes can be purchased with a supplementaldesignator that indicates a diffusible hydrogen concentration below 16, 8, or 4 ml



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H2/100g in the weld metal can be maintained (H16, H8, and H4 respectively)under most welding conditions .

The diffusible hydrogen potential (measured in ml/100g deposited weld metal)will depend on the type of consumable, welding process, plate/joint cleanliness,and atmospheric conditions in the area of welding. Some consumables mayabsorb moisture after exposure to the atmosphere. Depending on the type ofconsumable, this may result in a significant increase in the weld metal diffusiblehydrogen concentration. In situations where control of diffusible hydrogenconcentrations is important, the manufacturer should be consulted for advice onproper storage and handling conditions required to limit moisture absorption.

Hydrogen assisted cracking may be avoided through the selection andmaintenance of an adequate preheat /interpass temperature and/or minimum heatinput. Depending on the type of steel and restraint level, a trade-off between aneconomic preheat/interpass temperature and the diffusible hydrogen potential ofa given process exists. There have been several empirical approaches developedto determine safe preheat levels for a given application that include considerationof carbon equivalent, restraint level, electrode type, and preheat. When followed,the guidelines for preheat that have been established in AWS D1.1 and D1.5 aregenerally sufficient to reduce the risk of hydrogen assisted cracking in most mildsteel weldments.

Hydrogen assisted cracking will typically occur up to 72 hours after completionof welding. For the strength of materials currently used in moment frameconstruction, inspection of completed welds should be conducted no sooner than24 hours following weld completion.

8.2.4 Preheat and Interpass Temperatures


8.2.5 Postheat


8.2.6 Controlled Cooling


8.2.7 Metallurgical Stress Risers




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8.2.8 Welding Preparation & Fit-up


Documents

FEMA 267