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EFFECT OF COLUMN STIFFENER DETAILING AND WELD FRACTURE TOUGHNESS ON THE PERFORMANCE OF WELDED

MOMENT CONNECTIONS

R.J. Dexter, University of Minnesota, United States of America J.F. Hajjar, University of Minnesota, United States of America

D. Lee, Research Institute of Industrial Science & Technology, South Korea

ABSTRACT Full-scale cyclically loaded cruciform experiments with weak panel zones and welded unreinforced flange-welded web (WUF-W) prequalified moment connections performed well provided that the weld metal has minimum Charpy V-Notch (CVN) toughness. Welds with low CVN and brittle fracture was obtained in one specimen despite using weld metal certified to meet minimum CVN requirements. Unstiffened columns perform well and alternative stiffening details all performed well, indicating that variations in column stiffening details did not affect the potential for fracture or low-cycle fatigue. Low-cycle fatigue performance is compared to strain range-cycles curves extrapolated from high-cycle S-N curves.

INTRODUCTION Following the Northridge earthquake of January 17, 1994, extensive damage to steel moment connections was reported (1-3). This damage most often consisted of brittle fractures of the bottom girder flange-to-column flange Complete Joint Penetration (CJP) groove welds. The fractures were caused by the use of low toughness welds combined with a number of other connection detailing and construction practices that were typical prior to the earthquake (3-5). Additionally, column stiffening practices have been cited as a possible contributor to the fractures, largely as a result of observations that many of the connections fractured during the Northridge earthquake lacked continuity plates and that some had weak panel zones (6). Finite element analyses (7-10) also have shown an increase in stress and strain concentrations in the girder flange-to-column flange CJP welds associated with excessively weak panel zones or insufficient continuity plates. It is presumed that these stress and strain concentrations increase the potential for fracture. As a result of these observations, there has subsequently been a tendency to be more conservative than necessary in designing and detailing of the continuity plates and doubler plates in steel moment connections. Recommendations for the seismic design of new steel moment-frame buildings (3) provide equations for determining whether continuity plates are required, and indicate that any required continuity plates must be of equal thickness to the girder flange for interior connections (thinner continuity plates are permitted for exterior connections), unless connection qualification testing demonstrates that the continuity plates are not required. Furthermore, the connection of the continuity plates to the column flanges must be made with CJP welds, and reinforcing fillet welds should be placed under the backing bars. Design criteria for the limit states related to column stiffening are presented in the AISC

Connections in Steel Structures V - Amsterdam - June 3-4, 2004 177

LRFD Specification (11). The limit states of primary importance for stiffening of connections include Local Flange Bending (LFB), Local Web Yielding (LWY), and Panel Zone yielding (PZ). Additional provisions for seismic design of doubler plates and continuity plates were included in the AISC Seismic Provisions (12). However, AISC (13) removed all continuity plate design procedures for Intermediate and Special Moment Frames, requiring instead that they be proportioned based on connection qualification tests. The tendency towards being more conservative than necessary in column stiffener design has raised concerns about economy as well as the potential for cracking of the k-area in the column web near the web-flange junction during fabrication due to high residual stresses caused by highly restrained CJP welds on the continuity plates or doubler plates (14,15). Therefore, a combined experimental and computational research study was conducted at the University of Minnesota to reassess the recent column stiffener design and detailing provisions and recommendations, and to provide economical alternative stiffener details that minimize welding along the column k-line while retaining superior performance for non-seismic and seismic design (16). This paper examines the effect of variations in column stiffening, including no stiffening as well as more economnical alternative stiffening details, on the fracture and low-cycle fatigue performance of Welded Unreinforced Flange-Welded Web (WUF-W) moment connections (3). The design and results of these tests are reported elsewhere, including an assessment of LFB and LWY limit states (17) and the cyclic panel zone behavior and design (18). Related research included nine pull-plate experiments (19-22) that investigated the limit states of LFB and LWY, primarily for non-seismic design, and further tested the alternative doubler plate and continuity plate stiffener details. Finite element analyses of all experimental specimens were also conducted as part of this research as well as parametric studies to extend the results to member sizes and details not tested (23).

FULL-SCALE CONNECTION TESTS Six full-scale, girder-to-column cruciform specimens were tested (Table 1). The SAC (24) loading history was applied including six cycles at each interstory drift level of 0.375%, 0.5%, and 0.75%, four cycles at 1.0% interstory drift level, and two cycles at each interstory drift level of 1.5%, 2.0%, 3.0%, and 4.0%. ASTM A992 wide-flange sections and A572 Grade 50 plate was used. The specimens used the pre-qualified (3) WUF-W connection detail (Figure 1). The column stiffening was varied in these specimens including three alternative doubler plate details (i.e., back-beveled fillet-welded doubler plate, square-cut fillet-welded doubler plate, and groove-welded box doubler plate; see Figure 2) and a fillet-welded ½ in. thick fillet-welded continuity plate detail. Table 2 presents the design strength-to-demand ratios using minimum specified material properties for the LFB and LWY limit states. Note that the design strength for LFB and LWY is the design strength of the column shape alone and does not include the column reinforcement, if any. The demand is calculated with various methods:

• Yield strength of the girder flange, • A value of 1.8 times the yield strength of the flange as in the 1992 AISC Seismic

Provisions (12) • Equation included in AISC Design Guide No. 13 (25) which amplifies the plastic

moment due to shear and reduces this moment to a force couple of the flanges. This equation, or slightly modified forms of it, has been widely used for the design of

178 Connections in Steel Structures V - Amsterdam - June 3-4, 2004

column stiffeners in steel moment connections (3). Table 2 shows that 1992 AISC Seismic Provisions (12) and AISC Design Guide No. 13 (25) provide similar demand values. For these cases, the demand exceeds the capacity for LFB for all specimens except CR1, whereas only specimen CR3 has continuity plates (although the box detail functions as a continuity plate as well as a doubler). The remaining specimens CR2 and CR5 are underdesigned for LFB for seismic demand. Table 1. Test matrix of cruciform specimens.

CR1 CR2 CR3 CR4 and CR4R CR5

Girder W24x94 W24x94 W24x94 W24x94 W24x94

Column W14x283 W14x193 W14x176 W14x176 W14x145 Doubler Plate

(DP) None Detail II Detail II Detail III Box (Offset) Detail I

DP Thickness NA 0.625 in. 2 @ 0.5 in. 2 @ 0.75 in. 2 @ 0.625 in.Continuity Plate (CP) None None Fillet-

welded None None

CP Thickness NA NA 0.5 in. NA NA

Figure 1. Typical welding details used for cruciform specimens (Specimen CR1).

However, the girder flange demand predicted by the latter two methods is very conservative and can be put in perspective by comparing to the maximum possible uniaxial tensile strength of A992 steel. The stress in the flange is 1.8 times 50 ksi or 90 ksi, well above the likely tensile strength of A992 steel. For example, a survey of more than 20,000 mill reports from (26,27) showed that A992 steel has a mean tensile strength of 73 ksi. The 97.5 percentile tensile strength was 80 ksi, and the maximum value reported was 88 ksi. Also shown in Table 2 are the panel zone capacity-to-demand ratios (including the strength of the doubler plate in the capacity) and column-girder moment ratios calculated from the


2002 AISC Seismic Provisions (28), assuming no axial compression in the column. The panel zone demand exceeds the capacity for all specimens. Therefore, if column stiffening were necessary to prevent premature brittle fracture or low-cycle fatigue, these specimens are a worst-case test since they are underdesigned.

Figure 2. Doubler details: (a) back-beveled fillet-welded doubler (Detail I), (b) square-cut

fillet-welded doubler (Detail II), (c) box (offset) doubler (Detail III). Table 2. Nominal capacity/demand ratios of PZ yielding, LFB, and LWY limit states.

LFB φRn/Ru LWY φRn/Ru

*

*pc

pb

MM

∑∑

(AISC)

PZ φvRv/Ru (AISC) Yield a

1.8Yield a 1992

Seismic DG13 a Yield a

1.8Yield a 1992

Seismic DG13 a

CR1 1.50 0.72 3.04 (3.38)b

1.69 (1.88)

1.64 (1.82) 2.38 1.32 1.29

CR2 0.99 0.66 1.47 (1.63)

0.82 (0.91)

0.80 (0.89) 2.20 1.22 1.19

CR3 0.89 0.74 1.22 (1.36)

0.68 (0.76)

0.66 (0.73) 2.51 1.39 1.36

CR4 CR4R 0.89 0.93 1.22

(1.36) 0.68

(0.76) 0.66

(0.73) 3.19 1.77 1.73

CR5 0.73 0.74 0.84 (0.93)

0.47 (0.52)

0.46 (0.51) 2.34 1.30 1.27

a Equation used to calculate demand, Ru b Values in parentheses reflect use of φ = 1.0 Weld details E70T-1 (Lincoln Outershield 70) wire with 100% CO2 shielding gas was used for all shop welding. The girder flange-to-column flange CJP groove welds were made in the flat position with E70T-6 (Lincoln Innershield NR-305) wire. Welds made with E70T-6 wire are required by AWS A5.20 (29) and AISC 2002 Seismic Specifications (28) to have notch toughness of 20 ft-lbs at -20°F. FEMA 350 (3) has recommended minimum notch toughness requirements at two temperatures, 20 ft-lbs at 0°F and 40 ft-lbs at 70°F. According to the Lincoln Electric Company product family literature, the typical values for NR-305 are 21 to 35 ft-lbs at -20°F and 21 to 54 ft-lbs at 0°F. As shown in Table 3, Specimens CR1 and CR4 were fabricated with a 5/64 in. diameter NR-305 wire and the remaining were fabricated with 3/32 in. diameter NR-305 wire. All CJP welds were ultrasonically tested by a certified inspector in conformance with Table 6.3 of AWS D1.1-2000 (30) for cyclically loaded joints. The out-of-position field welds, including the CJP welds connecting the girder web to the column flange and all reinforcing fillet welds were made with 0.068 in. diameter E71T-8 (Lincoln Innershield NR-203MP) wire for Specimens CR1 and CR4, and 5/64 in. diameter


E71T-8 (Lincoln Innershield NR-232) wire for the other specimens. The shear tab was designed to extend approximately 0.25 in. into the top and bottom access holes and acted as the backing bar for the CJP welds of the girder web to the column flange. This extension acted as a short runoff tab, allowing the weld to extend the full depth of the girder web. Ricles et al. (31) recommended that these runoff tabs of the vertical web weld be ground smooth, which is labor intensive. Since it was felt that this might not be necessary, these runoff tabs were not ground smooth in the specimens tested in this work. Table 3. Tested weld material properties (E70T-6 only).

E70T-6a 5/64 in. wire

E70T-6 3/32 in. wire

CR1a CR4a CR2 CR3 CR4R CR5 CVN @ 0°F

(ft-lbs) 2.6 2.0 34.3 44.3 33.0 33.0

CVN @ 70°F (ft-lbs) 19.3 2.3 54.3 73.3 58.7 53.7

Fy (ksi) NA NA 59.5 50.0 56.0 53.5 Fu (ksi) NA NA 79.5 72.5 78.2 75.5

% Elongation NA NA 25.0 23.0 27.5 26.0 a For Specimens CR1 and CR4, the CVN tests were performed on specimens machined after the experiment from the welds that did not fracture in the cruciform joints BRITTLE FRACTURE Specimen CR4 exhibited premature brittle failure in three of four girder flange-to-column flange CJP welds in the early stage of the SAC (1997) loading history and was stopped after one-half cycle at 2.0% interstory drift. It was found that this specimen was unintentionally prepared with low-toughness weld metal, as shown in Table 3. Note that the AWS Certificate of Conformance for this wire indicated that the weld metal meets the minimum toughness requirement of AWS A5.20 (28) of 20 ft-lbs at -20°F (32). Specimen CR4R was essentially a replicate test of Specimen CR4, except that the weld metal used for Specimen CR4R met the minimum requirements of FEMA 350 (3). Specimen CR4R not only performed acceptably according to the SAC (24) requirements, it performed as well as any of the other specimens successfully tested in this experimental study. The fact that the box (offset) doubler plate detail performed well in Specimen CR4R indicates that the detail itself was probably not a factor in the fracture that occurred in Specimen CR4. Following the premature brittle failure in Specimen CR4, it was found that the previously tested Specimen CR1 also had relatively low weld-metal notch toughness, with an average of 2.6 ft-lbs at 0°F and 19.3 ft-lbs at 70°F as presented in Table 3. Specimen CR1, which was welded using the same wire and the similar welding procedure as Specimen CR4, but had marginally better notch toughness, performed very well, experiencing 14 cycles of 4.0% interstory drift before the significant strength degradation. It is important to note that this Specimen CR1 had no doubler plates or continuity plates, even though doubler plates would be required as shown in Table 2. Thus this test shows that column stiffeners are not absolutely required to avoid brittle fractire or low-cycle fatigue, even with this very poor notch toughness. These two tests have closely bracketed the minimum notch toughness required for adequate performance of CJP welds. It is believed that the FEMA 350 requirements (3) for minimum notch toughness are


adequate, provided they can be consistently met. FEMA 353 (4) requires toughness testing on each production lot of the specified filler metal. However, lot testing is typically not done since FEMA 353 also allows this requirement to be waived (upon approval of the Engineer), instead relying upon the consumable manufacturer’s certification. However, the certification is not necessarily reliable; for example, this 5/64-in. diameter E70T-6 wire that produced these brittle welds had been certified by the manufacturer as meeting the minimum 20 ft-lbs at -20°F required by the AWS certification test (32). Therefore, further evaluation of the CVN testing requirements for weld metal is warranted. LOW-CYCLE FATIGUE The remaining connections exhibited no brittle fractures, but low-cycle fatigue failures occurred after significant cyclic loading. Figures 3 and 4 show low-cycle fatigue cracks forming at the beam flange weld. These beam flange weld cracks were the only type of low-cycle fatigue cracks that actually propagated to cause failure, which is defined here as significant strength reduction. Low-cycle fatigue cracks did originate at the weld of the beam web to the column flange and at the weld access hole, as shown in Figure 5. However after propagating for a short distance they arrested and did not propagate further or lead to failure, so therefore they are not structurally significant.

(a) (b)

Figure 3. Low cycle fatigue crack developing at the toe of the beam flange weld in a moment-frame connection after (a) 11 cycles and (b) 17 cycles of 4% drift.

Table 4 shows the cycles at 3% or 4% drift when the first crack was first visible in the CJP welds, the final cycles at 4% drift when the strength was reduced, and the measured strain ranges. The pairs of measured strain ranges for each specimen are from the west top flange and the east bottom flange, respectively (except for CR1 where only the west top flange data were avaialble). The average of five gages across the width was used to eliminate some of the scatter and the effect of strain gradients. It is believed that the variation in the measured strain ranges is random, and that the strain range at 4% drift was relatively consistent among the specimens, averaging approximately 4.1%.

The performance requirement is that the connections must complete 2 cycles at 4% drift without a significant reduction in strength in order to be prequalified connections (3). One could conclude that all these connections (except the original CR4, which experienced brittle fracture) met this performance requirement and that therefore the performances of the specimens are equally good. However, there may be some significance to the final number of cycles before


strength reduction. In particular, it is believed that the differences in number of cycles to strength reduction between 12 and 16 are due to random variability, and that therefore specimens CR1, CR2, CR3, and CR4R performed equally well. This means that the variation in column stiffener detailing and panel zone strength among these specimens had no significant influence on the low-cycle fatigue performance. For specimen CR5, the number of cycles before strength reduction is 6. As shown in Table 2, this specimen had the lowest ratio of capacity to demand for LFB – less than one even for non-seismic (nominal yield strength of flange) and less than 0.5 for either seismic demand. It is likely that the relatively smaller number of cycles in this under-designed specimen is due to lack of continuity plates.

Figure 4. Cross section of beam flange weld showing low cycle fatigue crack

developing at the weld toe.

Figure 5. Low cycle fatigue cracks forming at the end of the beam web to column flange weld and at the weld access hole.

Table 4. Low-cycle fatigue data.

Specimen CR1 CR2 CR3 CR4R CR5 Column W14x283 W14x193 W14x176 W14x176 W14x145

Doubler Plate None Detail II Detail II Box Detail I Continuity Plate None None Fillet welded None (box) None

Cycles when Crack Visible 11@4% 2@3% 2@3% 2@4% 1@3%

Cycles at 4% when Strength Reduced 14 16 14 12 6

Strain Range in Girder Flange at

4% Drift 4.2% 4.5%, 3.7% 3.8%, 4.5% 3.8%, 3.8% 3.7%, 4.8%

Column Web

Column Flange

LCF Crack

Slag Inclusion, LOF


As noted above, specimen CR1 had much lower notch toughness than the others. Therefore, it appears that, as long as the notch toughness is sufficient to preclude brittle fracture, notch toughness above this minimum level also has little influence on low-cycle fatigue performance. Finally, for Specimens CR1 through CR4, there is little correlation betweeen the cycles when the crack was first detected in the CJP welds, which was highly variable, and the final number of cycles when strength was reduced. Therefore, it is believed that the number of cycles at which the crack is first detected is not significant. Most past research on low-cycle fatigue has involved pressure vessels and some other types of mechanical engineering structures. Since low-cycle fatigue is an inelastic phenomenon, the strain range is the key parameter rather than the stress range as in high-cycle fatigue. The Coffin-Manson rule (33) has been used to relate the strain range in smooth tensile specimens to the fatigue life. Manson suggested a conservative lower-bound simplification, called Manson’s universal slopes equation (34):

6.06.012.05.3 −− +=∆ NNE f

u εσ

ε (1)

where: ∆ε is the total strain range, σu is the tensile strength, and εf is the elongation at fracture. Note that the first term in Equation 1 is the elastic part of the total strain range (which is relatively insignificant when there are fewer than 100 cycles) and the second term is the plastic part of the total strain range. Figure 6 shows a plot of Manson’s universal slopes equation where σu is 450 MPa and εf is 25%, typical minimum properties for Grade 50 structural steel. Many studies have shown that Manson’s universal slopes equation is conservative compared to experimental data from smooth specimens (34,35). However, because of buckling at greater strain ranges, most of the experimental data are for strain ranges less than 1%, i.e., for cycles greater than 1000. Limited data exist at higher strain ranges – some are shown in Figure 6 for A36 steel smooth specimens machined from the flanges of wide-flange sections (35). At this time, very little is understood about low-cycle fatigue in welded or bolted structural details. For example, it is a very difficult task to predict accurately the local strain range at a location of cyclic local flange buckling. However, Krawinkler and Zohrei (36) and Ballio and Castiglioni (37,38) showed that the number of cycles to failure by low-cycle fatigue of welded connections could be predicted by the local strain range in a power law that is analogous to an S-N curve. Ballio and Castiglioni (37,38) showed that the power law would have and exponent of 3, just like the elastic S-N curves. Krawinkler and Zohrei (36) also showed that Miner’s rule (39) could be used to predict the number of variable-amplitude cycles to failure based on constant amplitude test data. Therefore, it may be possible to predict and design against low-cycle fatigue using strain-range vs. number-of-cycles curves that are extrapolated from the high-cycle fatigue design S-N curves. Figure 6 shows the AISC S-N curves (11) for Categories A and C, converted from stress range to strain range by dividing the stress ranges by the elastic modulus, and extrapolated up to one cycle. There are only limited data to support this approach. Figure 6 shows the strain range-number of cycles data from these WUF-W beam-to-column connection tests. The number of cycles plotted in Figure 6 is the number of cycles at 4% drift. If the effect of all the previous cycles is included using Miner’s rule (39), they add up to an equivalent of about one additional cycle at 4%. Since flange strains right at the weld toe were used rather than nominal values, this is analogous to a hot-spot approach for high-cycle fatigue. For high-cycle fatigue, the Category C S-N curve is a suitable baseline S-N curve for the hot-spot approach (40,41). It appears that the Category C S-N curve is also a good lower bound to these low-cycle fatigue data. The scatter in the data is substantial, as is also true in high-cycle fatigue.


Also shown in Figure 6 are previously unpublished data for smaller coupon-type specimens with transverse butt welds, which would be expected to be Category C details. These are some of the only available data with fewer than 5,000 cycles. These coupon data are also in reasonable agreement with the extrapolated Category C curve as a lower bound.

0.1

1

10

100

1 10 100 1000 10000Cycles

welded couponsmooth specimenWUF-W connection tests

Category A

Category C

Manson's Equation

Figure 6. Comparison of standard S-N curves presented in terms of strain range and Manson’s universal slopes equation for Grade 50 (350 MPa yield strength) steel to low-cycle fatigue test

data and the connection test data. CONCLUSIONS AND RECOMMENDATIONS

1. When properly detailed and welded with notch-tough filler metal, the WUF-W steel moment connections can perform adequately even though relatively weak panel zones and low local flange bending strengths were chosen. 2. The failure mode of the specimens other than the original CR4 was low-cycle fatigue (LCF) crack growth and eventual rupture of one or more girder flange-to-column flange complete joint penetration (CJP) groove welds. Low cycle fatigue may be conservatively predicted using strain-range vs. cycles curves derived from the stress based S-N curves for high-cycle fatigue. 3. Specimens CR1 and CR4 were unintentionally prepared with weld metal that had CVN values that were much lower than the minimum requirements. The premature brittle failure of specimen CR4 reconfirmed that achieving the required minimum CVN toughness in the girder flange-to-column flange CJP welds is critical. These low toughness welds occurred despite the certification of the filler metal; the certification is only required annually, unlike the way that each heat of steel is tested. A study should be conducted to fully characterize the typical variability in the CVN and other properties of the weld. An evaluation of the need for lot testing should be performed. Consideration should also be given to use of filler metals with a distribution of CVN such that there is a sufficiently small probability of not meeting the minimum required values, and therefore lot testing may not be required. 4. Application of the alternative column stiffener details (i.e., back-beveled fillet-welded doubler plate detail; square-cut fillet-welded doubler plate detail; groove-welded box doubler plate detail; fillet-welded 1/2 in. thick continuity plates) in the WUF-W steel moment connections was successfully verified. No cracks or distortions were observed in the welds connecting these stiffeners before the rupturing of the girder flange-to-column flange CJP

Stra

in R

ange

[%]


welds. 5. Specimens CR1, CR2, CR4R, and CR5, none of which had continuity plates (although Specimen CR4R included the offset doubler plate detail), met the requirements for two cycles at 4.0% interstory drift, athough only Specimen CR1 met the seismic requirements of AISC and FEMA with respect to continuity plates. Continuity plates may thus not be necessary in many interior columns in steel moment connections, and design provisions permitting the design, or lack of inclusion, of continuity plates are recommended for reintroduction into the AISC Seismic Provisions. 6. For a wide range of column sections and doubler plate detailing, strain gradients and strain magnitudes well above the yield strain in the girder flanges did not prohibit the specimens from achieving the connection prequalification requirement of completing two cycles at 4.0% interstory drift without significant strength degradation. This was even the case for specimen CR1, which had notch toughness in the weld metal that was significantly below the requirements. These results indicate that the column reinforcement detailing may not have a significant effect on the potential for brittle fracture at the girder flange-to-column flange weld. Note that this is contrary to previous finite-element analyses reported in the literature using theoretical fracture criteria that have predicted a significant effect of using or omitting continuity plates. 7. If continuity plates are required, fillet-welded continuity plates that were approximately half of the girder flange thickness performed well. The results showed that only minor local yielding occurred in these continuity plates in part of the most stressed cross sections at peak drift level and that these strains were not sufficient to cause cracking or distortion in the continuity plate or to change the strain gradients in the girder flange substantially. Since continuity plates do not significantly yield, it may not be necessary to size the welds large enough to develop the continuity plate. Rather the weld and the plate may only need to be designed for the difference between the demand and the capacity of the column shape without the continuity plate. Continuity plates with undersized fillet welds should be tested to confirm that the weld need not develop the full continuity plate strength. 8. In all the tests except the original CR4, the seismic performance of the relatively weak panel zones was stable and ductile, and the panel zones exhibited good energy dissipation. Lee et al. (17-23) provide recommended changes to the AISC panel zone strength equations, as well as detailed evaluations of current AISC provisions for local flange bending, local web yielding, and panel zone shear.

ACKNOWLEDGEMENT This research was sponsored by the American Institute of Steel Construction, Inc. and by the University of Minnesota. In-kind funding and materials were provided by LeJeune Steel Company, Minneapolis, Minnesota; Danny’s Construction Company, Minneapolis, Minnesota; Braun Intertec, Minneapolis, Minnesota; Nucor-Yamato Steel Company, Blytheville, Arkansas; Lincoln Electric Company, Cleveland, Ohio; and Edison Welding Institute, Cleveland, Ohio. Supercomputing resources were provided by the Minnesota Supercomputing Institute. The authors wish to thank T. V. Galambos and P. M. Bergson, University of Minnesota, L. A. Kloiber, LeJeune Steel Company, and the members of the technical advisory group on this project for their valuable assistance. REFERENCES 1. Youssef, N. F. G, Bonowitz, D., and Gross, J. H. (1995). A Survey of Steel Moment-

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20. Prochnow, S. D., Ye., Y., Dexter, R. J., Hajjar, J. F., and Cotton, S. C. (2002). “Local Flange Bending and Local Web Yielding Limit States in Steel Moment-Resisting Connections,” Connections in Steel Structures IV, Roanoke, Virginia, October 22-24, 2000, American Institute of Steel Construction, Chicago, Illinois, pp. 318-328.

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22. Hajjar, J. F., Dexter, R. J., Ojard, S. D., Ye, Y., and Cotton, S. C. (2003). “Continuity Plate Detailing for Steel Moment-Resisting Connections,” Engineering Journal, AISC, Vol. 40, No, 4, Fourth Quarter, pp. 189-211.

23. Ye, Y., Hajjar, J. F., Dexter, R. D., Prochnow, S. C., and Cotton, S. C. (2000). “Nonlinear Analysis of Continuity Plate and Doubler Plate Details in Steel Moment Frame Connections,” Report No. ST-00-3, Department of Civil Engineering, University of Minnesota, Minneapolis, Minnesota.

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27. Bartlett, F.M., R.J. Dexter, M.D. Graeser, J.J. Jelinek, B.J. Schmidt, and T.V. Galambos (2003). “Updating Standard Shape Material Properties Database for Design and Reliability”, Engineering Journal, AISC, Vol. 40, No. 1, 1st Quarter, pp 2-14.

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EFFECT OF COLUMN STIFFENER DETAILING AND …citg.home.tudelft.nl/.../-_Table_of_Contents/doc/s4-1-Dexter.pdfeffect of column stiffener detailing and weld fracture toughness on the