DFI JOURNAL · Jet grouting has increasingly become a ground improvement technology used to address seepage concerns and provide strength improvement to soils underlying dams. The

Vol. 6, No. 1 July 2012

DFI JOURNALThe Journal of the Deep Foundations Institute

PAPERS:

Jet Grouting and Safety of Tuttle Creek Dam – Timothy D. Stark, Paul J. Axtell, Francke C. Walberg, John C. Dillon, Glen M. Bellew, and David L. Mathews [3]

Challenges and Uncertainties Relating to Open Caissons – Fathi Abdrabbo, Khaled Gaaver [21]

The Infl uence of RC Nonlinearity on p-y Curves for CIDH Bridge Piers – Leonardo Massone, Anne Lemnitzer [33]

Installation and Performance Characteristics of High Capacity Helical Piles in Cohesive Soils – Mohammed Sakr [41]

Results of Dynamic and Static Load Tests on Helical Piles in the varved clay of Massachusetts – Jorge Beim, Severino Carlos Luna [58]

Deep Foundations Institute is the Industry Association of Individuals and Organizations Dedicated to Quality and Economy in the Design and Construction of Deep Foundations.

DFI JOURNAL Vol. 6 No. 1 July 2012 [1]

From the Editors and Publisher 2012 DFI Board of TrusteesPresident:James A. MorrisonKiewit Engineering Co.Omaha, NE USA

Vice President:Patrick BerminghamBermingham Foundation SolutionsHamilton, ON Canada

Secretary:John R. WolosickHayward Baker Inc.Alpharetta, GA USA

Treasurer:Robert B. BittnerBittner-Shen Consulting Engineers, Inc.Portland, OR USA

Immediate Past President:Rudolph P. FrizziLangan Engineering & Environmental ServicesElmwood Park, NJ USA

Other Trustees:David BorgerSkyline Steel LLCParsippany, NJ USA

Maurice BottiauFranki Foundations Group BelgiumSaintes, Belgium

Dan BrownDan Brown and Associates, PLLCSequatchie, TN USA

Gianfranco Di CiccoGDConsulting LLCLake Worth, FL USA

Bernard H. HertleinAECOM Technical Services Inc.Vernon Hills, IL USA

Matthew JanesIsherwood AssociatesBurnaby, BC Canada

James O. JohnsonCondon-Johnson & Associates, Inc.Oakland, CA USA

Douglas KellerRichard Goettle, Inc.Cincinnati, OH USA

Samuel J. KosaMonotube Pile CorporationCanton, OH USA

Kirk A. McIntoshAMEC E&I, Inc.Jacksonville, FL USA

Raymond J. PolettoMueser Rutledge Consulting EngineersNew York, NY USA

Arturo L. Ressi di CerviaKiewit Infrastructure GroupWoodcliff Lake, NJ USA

Michael H. WysockeyThatcher Engineering Corp.Chicago, IL USA

Journal PublisherManuel A. Fine, B.A.Sc, P.Eng

Journal EditorsAli Porbaha, Ph.D., P.E. California State University Sacramento, CA, USADan A. Brown, Ph.D. Dan Brown and Associates, Sequatchie, TN, USAZia Zafir, Ph.D., P.E. Kleinfelder Sacramento, CA, USA

Associate EditorsLance A. Roberts, Ph.D., P.E.RESPEC Consulting & ServicesRapid City, SD USAThomas Weaver, Ph.D., P.E.Nuclear Regulatory CommissionRockville, MD USA

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Standard Serial Number (ISSN): 1937-5247

Mission/Scope The Journal of the Deep Foundations Institute publishes practice-oriented, high quality papers related to the broad area of “Deep Foundations Engineering”. Papers are welcome on topics of interest to the geo-professional community related to, all systems designed and constructed for the support of heavy structures and excavations, but not limited to, different piling systems, drilled shafts, ground modification geosystems, soil nailing and anchors. Authors are also encouraged to submit papers on new and emerging topics related to innovative construction technologies, marine foundations, innovative retaining systems, cutoff wall systems, and seismic retrofit. Case histories, state of the practice reviews, and innovative applications are particularly welcomed and encouraged.

DFI JOURNAL

The papers in the current edition cover diverse subject matters. Helical pile continues to be a prolific subject, with two articles in this edition, one of which deals with helical piles in cohesive soils, a follow-up to a paper on high capacity helical piles in cohesionless soils by the same author, which appeared in the June 2011 edition of the Journal. The other helical pile paper deals with dynamic and static load testing on helical piles. There is a paper resulting from the 2011 DFI Young Professor Paper competition on the subject of p-y curves for CIDH bridge piers. The other two papers are on subjects that are new to the DFI Journal; a paper on jet grouting relating to the safety of the Tuttle Creek Dam in Kansas, USA, and a paper from Egypt on challenges and uncertainties relating to open caissons.

Our current system of obtaining reviewers wherein DFI Staff email all members of the DFI Technical Committees that relate to the paper in question, with the abstract quoted in the email requesting volunteers, is working very well. We usually receive more responses than actually required within 24 hours or less. We wish to thank our dedicated reviewers for their attention to detail and service to the authors, the DFI Journal publication and the industry in providing input to enhance quality of the papers.

We continue to maintain the tradition of publishing high-quality practice-oriented technical papers with diverse subject matters. The papers all relate to practical issues in our industry, with a strong presence of case history papers, as well as papers that confirm existing design methodology or may influence changes in such methodology. The Vol. 6, No. 2 edition to be published in the late Fall of 2012 will again be open to any subject matter meeting these criteria. Sometime in 2013 we expect to again publish a themed edition on a subject still to be determined. Any Technical Committee desiring to have any specific topic as the focus of a future themed edition should contact the Publisher.

Other comments, suggestions, and submissions are welcome and may be submitted via the DFI website at www.dfi.org.


Jet Grouting and Safety of Tuttle Creek DamTimothy D. Stark, Professor of Civil and Environmental Engineering, University of Illinois at Urbana-

Champaign, USA, [email protected]

Paul J. Axtell, Dan Brown and Associates, Kansas City MO, USA,

[email protected]

Francke C. Walberg, URS Corporation, Overland Park, KS, USA, Francke [email protected]

John C. Dillon, Glen M. Bellew, and David L. Mathews, U.S. Army Corps of Engineers, Kansas

City District, Kansas City, MO, USA

ABSTRACTJet grouting has increasingly become a ground improvement technology used to address seepage concerns and provide strength improvement to soils underlying dams. The technique of jet grouting uses high pressure/volume jet fluids to erode existing soil, evacuate some or most of the soil, and mix the remaining cuttings with cement slurry to form soilcrete. While considered a useful technology, this paper discusses some of the problems that can develop while jet grouting in or below a dam with an operational reservoir and seepage condition. Jet grouting experience at Tuttle Creek Dam indicates concerns with respect to ground fracture; spoil return, column diameter consistency, and homogeneity of resulting soilcrete. Recommendations are presented to increase monitoring of downhole parameters during jet grouting to better understand the downhole pressures and soil response during jet grouting.

JET G ROUTING TECHNOLOGYUse of jet grout walls as a barrier to water flow under dams and other sites has been used in-creasingly in recent times. (Croce and Modoni, 2007; Burke, 2007, Martin, et al., 2004; Yilmaz, et al., 2007; and Fang, et al., 2006). Jet grout col-umns have the potential to provide a continu-ous wall of relatively low hydraulic conductivity material; however, the effectiveness of jet grout construction can be difficult to assess, particu-larly when the project involves an operational dam. Usually the reservoir is operating at non-critical conditions (normal pool or non-seismic state) during jet grouting and after construc-tion. However, the cutoff wall is designed for a critical loading condition, and is not fully tested until that condition is experienced. This reduces confidence in the constructed elements for the design event(s).

Jet grouting has been in commercial use since approximately 1975 (Kauschinger, 2008). Cur-rently, the three main jet grouting systems are single, double, and triple fluid systems. In the single fluid system, grout slurry is ejected un-der high velocity through a horizontal nozzle which works as a cutting or erosion fluid to mix and evacuate soil. The grout slurry mixes with

non-evacuated soil and then hardens to create a soilcrete column. Single fluid jet grouting is most effective in loose coarse-grained/cohesion-less soils. The double fluid system usually uses air and grout ejected from two different nozzles that are placed opposite each other on the drill rod. However, some double fluid systems use one nozzle and air surrounds the grout to increase cutting/erosion. Double fluid jet grouting is most effective in loose to medium dense coarse-grained soils and some soft fine-grained soils. In the triple fluid system, air and water are used as cutting fluids above the grout nozzle. Separation of the erosion and grout mixing processes is thought to yield more uni-form columns. However, Stark et al. (2009) show considerable soil inclusions may be present in large diameter triple fluid columns. Triple fluid jet grouting is believed to be best for eroding and evacuating dense coarse-grained and some fine-grained soils.

For all jet grouting methods, field trials are usu-ally required to establish site-specific jet grout parameters, energy correlations, achievable column diameters, and assessment of quality of treated soil. Because jet grouting does not use positive displacement or a known mechanical

[4] DFI JOURNAL Vol. 6 No. 1 July 2012

excavation and mixing tool, it is more techni-cally demanding and less forgiving than other ground improvement methodologies, such as excavated slurry walls and non-jet assisted soil mixing.

Recent Dam Jet Grouting Trends and Precedence

Initially jet grouting systems were used to cre-ate soilcrete columns with diameters of 1 to 3 m (3 to 10 ft). Recent modifications and pro-cedures are being used to construct soilcrete columns with diameters up 5 m (16 ft). To cre-ate these larger diameter columns, high velocity jets, extremely high fluid pressure and volumes, slow rotation of the drill string, and slow lift rates are being used to erode and excavate a larger volume of soil. These high pressures can cause ground fracturing of the insitu soils because they exceed the borehole resistance to fracture. With either double fluid or triple fluid methods the use of air is crucial because it is readily mixed with the spoil cuttings and reduc-es the weight of the borehole column of spoil. The bottom hole pressure is thus only a fraction of the induced fluid pressures, and is only that pressure necessary to lift spoil to the ground surface. However, ground fracturing can occur nearly instantaneously if the annulus between the drill rods and the borehole or casing be-comes blocked because air, water, and/or grout are continuously being injected at extremely high pressures during jet grouting.

TUTTLE CREEK DAMTuttle Creek Dam is a U.S. Army Corps of En-gineers, Kansas City District (USACE) project located on the Big Blue River near Manhattan, Kansas, 200 km (125 miles) west of Kansas City. Tuttle Creek Dam is 41.8 m (137 ft) high earth and rockfill embankment with a length of about 2,288 m (7,500 ft). Details of the fill zones and construction of the dam are available in Lane and Fehrman (1960) or Walberg et al. (2012).

Tuttle Creek Dam was originally designed and constructed in the 1950’s prior to the de-velopment of recent earthquake engineering technology which accounts for the behavior of materials subjected to seismic shaking. Shortly after the upstream slope failure caused by soil liquefaction at Lower San Fernando Dam in 1971, the USACE began a program to evaluate all of its dams based on evolving technology. Investigations to assess the seismic stability of

Tuttle Creek Dam were conducted in the 1980’s and 1990’s and concluded that seismic reha-bilitation (specifically an upstream cutoff wall and slope stabilization and downstream slope stabilization) was required to assure the proj-ect could withstand the design ground motion without an uncontrolled release of the reservoir towards downtown Manhattan, Kansas.

Tuttle Creek Dam Foundation Conditions

The dam is founded on native alluvial soils con-sisting of 2.4 to 8.2 m (8 to 27 ft) of silt (ML) and clay (CL, CH, and OH) underlain by sand, silty sand, and gravely sand to a depth of 12.2 to 24.4 m (40 to 80 ft). The silt and clay deposit immediately below the dam forms a natural low hydraulic conductivity fine-grained blanket that facilitates seepage control by dissipating some of the hydraulic head imposed by the reservoir. This material is referred to as the fine-grained blanket herein. Below the fine-grained blanket are layers of loose, fine to medium sand to silty sand, and sand with silt (SP, SW, SM, SM-SP, and SM-SW) and medium dense to dense, coarse to gravelly sand to bedrock. The fine to medium sand and coarse to gravelly sand deposits typi-cally vary in thickness from about 7.6 to 18.3 m (25 to 60 ft). Interspersed with the sands are occasional relatively thin layers or lenses of clay and silt typical of recent alluvial deposits. The ground water surface is typically located at a depth of 2.7 m (9 ft) or Elevation 310.2 m (1017 ft) for Stations 30+00 to 50+00 at the downstream toe, but is dependent on reservoir elevation. Walberg et al. (2012) presents a de-tailed description of the foundation conditions.

Fig. 1 presents a typical cross-section of Tuttle Creek Dam showing the seepage control sys-tem which consists of an extended upstream impervious embankment zone, the natural fine-grained soil blanket underlying the embank-ment, and a downstream relief well system. The natural fine-grained blanket dissipates 13.7 to 15.2 m (45 to 50 ft) of the reservoir head upstream of the impervious zone before it en-ters the permeable foundation sands. Once the seepage reaches the permeable foundation sands, seepage continues to the downstream toe with less hydraulic head loss because of the high hydraulic conductivity of of the sands and gravels. The majority of underseepage is eventually intercepted by the relief well system along the downstream toe of the dam. Relief well flow discharges into an adjacent collec-


tor that runs along the downstream toe with several lateral ditches that take the water fur-ther downstream. Flumes placed in the lateral ditches show a continuous flow rate of 2,200 gpm (295 ft3/min = 4.9 ft3/sec = 0.14 m3/sec) at multipurpose pool (MPP) elevation of 327.7 m (1075 ft). Relief well flow rate is also pool-level dependent. The relief wells at the downstream toe are critical to protect the dam against foun-dation erosion and piping. The large earth-quake induced deformations predicted for the downstream toe would likely disable the down-stream relief wells. With loss of, or damage to, the relief wells, an internal erosion/piping fail-ure of the foundation soils could occur even at the MPP.

TUTTLE CREEK DAM FOUNDATION MODIFICATION PROJECTThe selected seismic retrofit alternative for Tuttle Creek Dam was to stabilize foundation soils without drawing down the reservoir. The initial upstream stabilization included jet grout-ing to stabilize the liquefiable foundation silty clays and sands below the upstream slope and to install an upstream cutoff wall (depth of ap-proximately 36 m (120 ft) on average) to reduce seepage and piezometric levels to acceptable levels at the downstream toe in case the relief wells were damaged during an earthquake. Downstream slope stabilization was required to stabilize the liquefiable foundation silty clays and sands. Downstream stabilization was to include either soil mixing or jet grouting to re-duce downstream slope movement and relief well damage. In September 2005, the USACE entered into a contract with Treviicos South, a ground improvement contractor, to construct the upstream cutoff wall and slope stabilization and downstream slope stabilization/foundation improvement.

The cutoff wall objective was to dissipate suf-ficient hydraulic head (11.9 m or 39 ft) so the pressure relief system along the downstream toe of the dam was not necessary at MPP be-cause it could become inoperable during or after the design ground motion. The upstream cutoff wall was to be constructed to a minimum thickness of 3.0 m (10 ft) using multiple (at least two) rows of full or partial jet grout col-umns. The wall was to penetrate a minimum of 0.3 m (1 ft) into the bedrock except in the area of a deep buried channel where a 3 m (10 ft) bedrock socket was required because of the presence of slump blocks overlying the alluvial sands and gravels in the channel. The contract did not specify how the bedrock embedment of 0.3 m (1 ft) and 3 m (10 ft) was to be obtained and did not require predrilling. In addition, it was unclear how the depth of embedment would be measured. It was expected that the depth of embedment would be determined by observations and/or recordings obtained from the Lutz instrumentation system used with the jet grouting equipment. It also was not clear how effective against seepage the cutoff wall/bedrock contact would be because the air-water jets might not be effective in cutting the lime-stone and hard shales in the bedrock. This seal at the bottom of the cutoff wall is important be-cause an improper seal can lead to reduced ef-ficiency of the cutoff wall with time (Rice et al., 2009a and b). Finally, it should be realized that the depth is usually measured from the depth of the nozzles. Therefore, the drill bit may have to penetrate 1 to 1.2 m (3 to 4 feet) below the bedrock interface to achieve a wall penetration of 0.3 m (1 ft).

One of the major challenges of the upstream jet grouting for both the cutoff wall and slope stabilization was the reservoir had to be main-tained, meaning the active foundation seepage

[FIG. 1] Typical dam cross section around Station 50+00 and proposed upstream cutoff wall (thick dark line) and slope stabilization

Cutoff Wall

↑


previously described (2,200 gpm or 0.14 m3/sec) would be present during construction and is significantly greater than the seepage in the downstream test area. Treviicos South had re-cently constructed a cutoff wall under similar conditions at Paso de las Piedras Dam in Argen-tina which facilitated their understanding of the project. The jet grouting pressures used at Paso de las Piedras Dam (Treviicos, 2007) are presented herein for comparison purposes with the Tuttle Creek Dam pressures.

Paso de las Piedras Dam Jet Grout Cutoff Wall

Paso de las Piedras Dam has a central impervi-ous core, a vertical chimney drain, and hori-zontal drainage blanket downstream of the core (Bustinza et al., 1999; Rattue, 2005, and Treviicos, 2007). A cutoff wall through and below the core was designed because the dam had a history of seepage related issues that re-sulted in the dam operating at a reduced pool level for many years. The dam is about 30 m (100 feet) high with upstream and downstream slopes of 2.75H:1V and 2.5H:1V, respectively. The foundation soils consist of gravel and silt/sand. A jet grout cutoff wall was constructed between 1998 and 2001 a distance of 12.2 m (40 ft) upstream of the dam axis. The cutoff wall columns are an average of about 22.9 m (75 ft) long, but in deeper areas of the foun-dation the columns average 39.7 m (130 ft) in length. Approximately 42,000 m2 of triple fluid jet grout columns were completed (Rat-tue, 2005). Primary column diameter is about 1.5 m (5 ft) with columns spaced about 2.4 m (8 ft) apart. After the primary row of columns was installed, secondary (2.4 m/8 ft diameter) and tertiary (1.5 m/5 ft diameter) columns were installed at various locations to close windows/gaps in the wall. The top of the wall extends 3.0 to 3.7 m (10 to 12 ft) into the impervious core of the dam. A total of 1,299 jet grout holes were

drilled. Only 12 holes had to be abandoned due to verticality, drilling difficulties, obstructions, or other reason.

DOWNSTREAM TEST P ROGRAMIn 2006 the contractor and the USACE initiated a test program downstream of Tuttle Creek Dam to prove the viability of jet grout and jet assisted soil mixing technologies and develop appropriate site-specific parameters before beginning production of the upstream and downstream foundation modification elements. The test program site is located about 152 m (500 feet) downstream of the dam and is ap-proximately 56.4 m (185 ft) wide and 103.7 m (340 ft) long). The jet grout and soil mix col-umns were surrounded by a perimeter cement-bentonite cutoff wall constructed to bedrock which allowed the test section area to be dewa-tered so column excavation could occur below the groundwater surface after construction. Stark et al. (2009) and Walberg et al. (2012) present additional details on the Downstream Test Program.

The jet grouting test program consisted of twenty-seven jet grout columns in three groups of nine. Columns were installed using both double and triple fluid jet grouting systems. The double fluid jet grout system was used to create columns 19 through 27 with a target diameter of 2.4 m (8 ft) (see Fig. 2). The triple fluid jet grout system was used to create col-umns 1 to 18 with target diameters from 2.4 to 3.0 m (8 to 10 ft). In each group of double fluid and triple fluid columns, one set of three columns was overlapped to assess column over-lap strength and integrity. For example, triple fluid columns 1, 2, and 3 (8 foot diameter) and 13, 14, and 15 (10 foot diameter) are the two groups of triple fluid columns that were over-lapped. The double fluid columns 25, 26, and 27 (8 foot diameter) is the group of double fluid

[FIG. 2] Final layout of jet grout columns for Tuttle Creek Dam test program. (Treviicos, 2006)


columns that were overlapped. The water-to-cement ratio for the columns varied from 0.75, 0.9, and 1.0. The columns were about 11 m (36 ft) in length with the top of the columns at about elevation 309.9 m (1016 ft) (3 m/10 ft be-low ground surface) and the base was at about elevation 298.9 m (980 ft).

Soil Inclusions in Completed Columns (Performed following Jet Grouting on Dam)

After the jet grout columns were completed and subsequent coring of the columns had been conducted, the groundwater within the cement-bentonite slurry wall was lowered to 11.3 m (37 ft) below ground surface (b.g.s.), i.e., eleva-tion 299 m (989 ft) utilizing two (2) dewater-ing wells and monitored by six (6) observation wells inside and around the slurry wall. Once the water level was reduced to the target depth of 11.3 m (37 ft), excavation to expose the col-umns proceeded. Fig. 3 shows the test site near the completion of excavation. The triple fluid columns are closer together because the actual column diameters are much greater than the 2.4 and 3.0 m (8 and 10 ft) target diameters (see Fig. 3). For example, triple fluid jet grout columns 1, 2, and 3 had diameters that ranged from 3.0 to 3.7 m (10 to 12 ft) and columns 10, 11, and 12 had diameters that ranged from 3.7 to 4.3 m (12 to 14 ft).

Following these observa-tions, nine (9) columns were chosen to be sec-tioned which would expose the inside of the columns to determine column integrity and homogeneity. The up-per 4.9 m (16 ft) of the columns were to be removed to allow sec-tioning of a 3 m (10 ft) segment of the selected columns between about elevation 305 m (1000 ft) and elevation 302 m (990 ft). Column groups 1-2-3 and 10-11-12 of the jet grout columns were chosen for sec-tioning. Fig. 3 shows that columns 1-2-3 cor-respond to 2.4 m (8 ft)

diameter columns that were overlapped while columns 10-11-12 correspond to 3 m (10 ft) di-ameter columns that were not overlapped. Both groups were constructed using triple fluid tech-nology. The columns were cut along the line of the group, effectively cutting each column in half.

The sectioning revealed that the cross-section of the jet grout test columns contained more than 40 to 50% native soil that was not broken up and evacuated during the jet grout process (Stark et al. 2009). In other words, 40 to 50% of the sectioned column consisted of chunks of native soil and about 60 to 70% of the sectioned triple-fluid column consisted of native soil. The inclusions were encountered throughout the entire diameter of the column. Most of the in-clusions were greater than 75 to 100 mm (3 to 4 inches) which could lead to blockage of spoil return. This large amount of soil inclusion could also have impacted the hydraulic conductivity of the resulting column.

The observed inclusions in the completed jet grout columns included significant amounts and large pieces of both fine-grained (silts and clays) and coarse-grained (fine sands and sands) soils. The following are two explanations for the inclusions: (1) fine grained soils can be dif-ficult to erode and break up into small enough

[FIG. 3] Downstream parametric jet grout and soil mix column groups during excavation and columns to be sectioned

Three rows

of soil mix

1-2-3 jet grout 10-11-12 jet grout


particles that can be evacuated to the surface through the drill rod annulus, especially at low natural water content, and (2) when large diam-eter columns are being excavated, the excavated roof of the cylindrical cavity may be unstable and may collapse introducing large pieces of soil into the unhardened slurry (Stark et al. 2009). This roof instability probably allowed large slabs of the natural fine-grained blanket material to break off, fall into the slurry, and then not be reduced by the cutting and mixing action of the rotating jets.

In summary, the significant amount and large size of the soil inclusions found in the complet-ed columns suggests that blockage of the an-nulus between the boring casing and jet grouting drill rod should be expected. This may lead to accumulation of high pressures in the subsurface and likely induce ground fracturing.

Downstream Jet Grouting Parameters and Ground Fracturing

Table 1 presents the triple fluid jet grouting parameters used to construct the two clusters of columns (1, 2, and 3 and 10, 11, and 12) that were sec-tioned. Table 1 shows that high water, grout, and air pressures were used to construct these columns. This resulted in some air bubbles being observed at the ground surface during downstream jet grouting. In addition, there was anecdotal evidence of ground fractur-ing during the downstream jet grout-ing including grout being found in the bottom of the two dewatering wells inside the cement-bentonite cutoff wall. It is recognized that other possible occurrences could have contributed to this condition, such as migration of lean grout through the coarse-grained soil directly above the bedrock, among others.

UPSTREAM CUTOFF WALL CHALLENGESAt the outset, it was recognized that construction of the upstream cutoff wall posed many challenges including column diameter consistency, verti-cality, variable stratigraphy, drilling through the upstream embankment

materials (rock fill), and securing the top of the cutoff wall to the extended impervious embank-ment material. The following paragraphs brief-ly discuss these challenges.

Column Diameter Consistency

Based on the Downstream Test Program, it was not certain that the required column diameter and adequate column consistency could be achieved, especially in the dense sands and stiff cohesive layers, to ensure the proposed cut-off wall would have a thickness of 3 m (10 ft) thick, would not have gaps, and could achieve the required 11.9 m (39 ft) head drop to create a safe condition if the relief wells were ren-

[TABLE 1] Triple fl uid jet grouting parameters for sectioned downstream parametric columns

Jet Group # 1, 2, 3 Group #10, 11, 12

Column Group

Column Diameter (m/ft)

2.4/8 3.1/10

Nozzles(# & diameter[cm/inches])

1 & 0.64/0.25

(w)

1 & 0.76/0.3

(c)

1 & 0.64/0.25

(w)

1 & 0.76/0.3

(c)

Water Pressure (MPa/psi)

45.0/6,525 44,988/6,525

Grout Pressure (MPa/psi)

25.0/3,626 24,994/3,626

Air Pressure (MPa/psi)

1.0/145 1,000/145

Grout Flow Rate (gal/min)

112.3 112.3

Station time (sec/four cm)

17.5 32

Rotation Speed (rpm)

4-8 2-4

Water Quantity ([gal/m] / [gal/ft])

800/244 1,417/432

Grout Quantity ([gal/m] / [gal/ft])

814/248 1,444/440

Cement Quantity ([kg/m] / [lbs/ft])

1,848/1,239.5 3,297/2,211

Specific Energy (MJ/m)

130-140 240-250

Grout Mix (C/W & B/W)

0.75 & 0.01 0.75 & 0.01


dered inoperable or damaged. The option of smaller diameter columns and additional rows (allowed in the contract) were viable options, but not options the contractor adopted due to cost concerns.

Variable Stratigraphy

Another jet grouting challenge at Tuttle Creek was the variable foundation stratigraphy and varying grain size distributions. Because of its cohesiveness, clayey material is difficult to evacuate from the boring during jet grouting. The downstream test program demonstrated that the columns contained numerous inclu-sions, but also that column diameter can be reduced significantly when a cohesive layer is encountered. Brill et al. (2003) show that plastic clays erode as chunks and pieces rather than small particles and that the chunks often lead to clogging of the drill rod annulus and subse-quent spoil return blockage. Even within the sand layers there is considerable variation in density/Standard Penetration Test (SPT) blow count values which can cause column variabil-ity. However, SPT or density variation in the sands will have less of an influence than the presence of clays because of the difficulty in eroding and breaking-down cohesive material which can result in annulus blockage. Success-ful jet grouting performance is a challenge with highly variable stratigraphy because jet grout parameters may have to be adjusted to provide different energy levels required to erode and grout different materials.

Securing the Top of Cutoff Wall to Embankment Material

Observations and coring of completed jet grout columns in the Downstream Test Program in-dicate that a void was usually created at the top of a jet grout column due to “bleed” of the unhardened grout and settlement of the materi-als involved. It was estimated that 75 to 90% of the completed jet grout columns had a void at the top. All of these voids were filled with grout following construction by topping-off the column. If a void develops at the top of a com-pleted column and is not filled, this can result in a preferential flow path over the top of the column. This flow path would exhibit a larger gradient than is currently being experienced in the foundation sands because a smaller amount of hydraulic head would be dissipated in the im-pervious zone and/or the natural fine-grained

blanket. This could result in unacceptable gra-dients and flow at the downstream relief well system. Additionally there was concern that the presence of voids could result in progressive erosion back to the reservoir creating an open flow path from the reservoir, across the top of the cutoff wall, into the foundation sands, and to the downstream toe. Such a condition could lead to an unstable seepage condition that also could not be controlled by the pressure relief well system. A similar concern occurred if ground fractures remained open and allowed di-rect communication between the pervious foun-dation and the reservoir. These scenarios could occur at high flood pools when implementation of a remedial measure would be difficult result-ing in a serious threat to dam safety.

Field observations also suggested that the pres-ence and/or quality of the extended impervious zone was in question. Historic as-built draw-ings suggest the presence of this zone and refer to the material as “impervious fill”. Several in-vestigative borings indicate this zone may have actually been shale “rock fill”. Original embank-ment construction specifications required dif-ferent placement techniques and compactive efforts for the two different types of fill. This discrepancy was recognized and considered a possible contributor to some of the observa-tions presented herein and made sealing of the cutoff wall to the embankment more suspect.

UPSTREAM JET GROUT CUTOFF WALL PARAMETRIC COLUMNSThe contractor continued the jet grout field trial program on the upstream face of Tuttle Creek Dam to confirm that the jet grouting param-eters developed in the downstream parametric columns were applicable to the upstream soils, higher hydraulic head imposed by the adjacent reservoir, and significantly increased confin-ing stresses because of the overlying embank-ment. This upstream work initiated concurrent with excavation and complete evaluation of the downstream test columns so some of the find-ings previously discussed in this paper were not yet fully understood as the upstream jet grout-ing commenced.

Pre-Drilling through Embankment Material for Jet Grouting

Fig. 1 shows the cutoff wall was to be located near the upstream extent of the extended im-


pervious zone. To install the cutoff wall the contractor pre-drilled borings through the upstream rock fill shell (see vertical jagged line above cutoff wall in Fig, 1). The borings extended to a depth of about 20.1 m (66 ft) to the proposed top of the cutoff wall. The initial diameter of the boring was 0.4 m (1.3 ft) and the hole was maintained using temporary steel casing and grout. Before the grout hardened in the boring, a 254 mm (10 inch) diameter PVC casing with no end cap (to prevent the casing from floating back up) was inserted in the bor-ing. Spacers or “spiders” were attached to the PVC casing at various depths in an effort to maintain casing verticality and its location in the center of the boring. These spacers had to span the 75 mm (3-inch) annulus in the 406 mm (16-inch) borehole. It became apparent that the flimsy and variable spacers aligning the casing in the large hole would not be able to maintain casing verticality.

After setting the PVC casing in the unhardened grout, a 203 mm (8 inch) drill bit was used to clean out the PVC casing and advance the bor-ing and jet grout tools to bedrock. If perfectly centered, this results in a 25 mm (1-inch) wide annulus of hardened grout in the casing prior to drilling the hole to rock for the jet grouting of the column. The hardened grout tended to shrink or detach from the casing wall which re-sulted in relatively large pieces of the grout an-nulus falling into the hole created for jet grout-ing. Evidence of the grout annulus falling into the hole is large pieces of curved grout were returned to the surface in the jet grout spoil.

Subsequently, using either sonic or augering drilling methods the contractor also installed ungrouted 254 mm (10 inch) diameter steel casing to the proposed top of the cutoff wall, elevation 314.2 m (1030 ft). A 203 mm (8 inch) drill bit was used to advance the jet grout bor-ing from the bottom of the 254 mm (10 inch) diameter steel casing to bedrock. The steel cas-ing provided a larger return annulus in the cas-ing because hardened grout was not present in the casing, avoided the issue of hardened grout falling into the column, and could be withdrawn upon completion of the column and grout back-filling to the work platform.

The jet grout drill rod diameter ranged from 60 to 114 mm (2.4 to 4.5 inches) depending on the jet grout system being used, e.g., double v. triple fluid jet grout system. The triple fluid

system utilized a drill rod diameter of 114 mm (4.5 inches). This created an annulus of about 44.5 mm (1.75 inches) in both the cased and uncased portion of the boring if the 25 mm (1-inch) wide annulus of hardened grout in the casing was still present. If the grout broke from the casing, the triple fluid annulus within the casing increased from 44.5 to 70 mm (1.75 to 2.75 inches). Thus, a soil clump with any di-mension of 44.5 to 70 mm (1.75 to 2.75 inches) could cause a partial blockage of the drill rod annulus. The sectioned columns previously de-scribed show sand and clay chunks much larger than these dimensions in the completed col-umns. The annuluses of 44.5 to 70 mm (1.75 to 2.75 inches) correspond to an annulus area of 0.022 to 0.041 m2 (34.3 to 62.6 in2), respectively. Thus, the spoil had to be evacuated through an annulus area of less than 0.041 m2 (62.6 in2). Of course larger column diameters require a larger volume of material to be eroded and evacuated through the annulus.

It is also recommended that the PVC or steel casing should extend to the bottom of the bore-hole, i.e., top of jet grout column, to provide the best connection possible to the top of the cutoff wall. Extending the casing to the bot-tom of the hole would provide the best condi-tions for a stable hole at the critical connection between the embankment materials and the top of the wall. This connection is important because if the connection is lost, leakage during jet grouting can occur at this location due to the high pressure air, water, and/or grout. This is discussed below as a possible cause for the ob-served air and possibly grout observed in Tuttle Creek Reservoir during upstream jet grouting. Due to elevation variations, equipment must be available to adjust the length of the PVC or steel casing. For example, the PVC casing was furnished in 6.1 m (20 ft) sections so additional 1.8 m (6 ft) of casing had to be added to reach a depth of 20.1 m (66 ft). If not, a gap of 1.8 m (6 ft) was created which could serve as a release point for high pressure air, water, and/or grout during grouting.

After installation of the casing to a depth of 20.1 m (66 ft) and drilling an 203 mm (8 inch) boring inside the 254 mm (10 inch) diameter PVC or steel casing to accommodate the jet grout drill rod, the grouting drill rod would then drill past the bottom of the cased hole to the desired depth of the column bottom, i.e., to


bedrock, a depth of approximately 36 m (120 ft) on average for the bottom of the cutoff wall. The jet grouting would then start from the bot-tom of the column and move upward to create the soilcrete column.

Observations during Upstream Jet Grouting of Triple Fluid Column S2-A3

Jet grout construction of triple-fluid column S2-A3 was attempted on Tuesday, 11 July 2006 using the jet grouting parameters in Table 2. The jet grouting started at 14:30 and ended about 145 minutes later at 16:55. Jet grouting was to occur from elevation 37.6 m (123.4 ft) to 27.9 m (91.5 ft). The contractor started stroking the hole to regain spoil return but only partial return was restored. At 15:20 the contractor reduced the water and grout pressure to 100 bars (1450 psi) and continued stroking the hole. At 15:35, the contractor reduced the water and grout pressure to 70 and 30 bars (1015 psi and 435 psi), respectively, and continued stroking the hole with the nozzles just below the PVC borehole casing. At 15:50, the contractor re-gained spoil return. However, spoil return was lost, but regained return shortly after, at the fol-lowing times and depths: 16:15 (31.2 m), 16:30 (29.9 m), 16:45 (28.5 m) and 16:50 (28.3 m).

At 16:55, contractor and USACE personnel noticed vigorous air bubbling in the reservoir about 15.3 m (50 ft) upstream of the work area. After this observation, the contractor stopped jet grouting, pulled the drill rods, and grouted up the drill hole around 17:00. The arrows in Fig. 4 indicate large areas of intense air bub-bling in Tuttle Creek Reservoir in the vicinity of jet grout column S2-A3. The bubbling air in the reservoir appeared from approximately Sta-tion 65+00 to 67+00 and suggested that ground fracture may have occurred. The air bubbling continued for almost 48 hours after jet grout-ing indicating a large volume of air was stored, probably in the pervious foundation materials, prior to release into the reservoir.

The loss of spoil return and the presence of high air, water, and grout pressures can frac-ture insitu materials as pressure builds up. For triple-fluid column S2-A3, water and grout were being injected at the rate of about 341 and 448 liters/minute, respectively (see Table 2), while the air pressure was being maintained at 9 bars (130.5 psi). While bubbles of air escaping in the reservoir may have indicated damage to the embankment and/or the natural fine-grained

blanket, the first manifestation of damage, e.g., increased seepage or gradient, can occur long after fracturing has initiated, particularly when the pool level increases. As a result, the USACE carefully monitored the downstream relief wells and toe area for signs of erosion, sand boils, and increased seepage for three weeks after S2-A3. Piezometers were also monitored to identify any changes in the hydraulic heads in the foundation soils. No abnormal reading or observations were observed so the heightened surveillance was reduced after three weeks. No higher than expected piezometric levels have been observed since completion of all construc-tion at the dam up to a pool level approximately 9.2 m (30 feet) above MPP.

Table 2 also presents the specific energy in-duced by the jet grouting (EJG) which can be re-lated to column diameter (Schlosser, 1997). For double-fluid jet grouting the specific energy is calculated by adding the energy imparted by the grout and air (see Equation (1)) (Pagliacci et al., 1994). For triple-fluid jet grouting the specific energy is calculated by adding the energy im-parted by the water and air (see Equation (2)).

[1]

[2]

where PG, P

W, and P

A are the grout, water, and

air pressures in MPa, respectively, QG, Q

W, and

QA are the grout, water, and air flow rates in

m3/hour, respectively, and VS is the withdrawal

speed in m/hour. Use of these units will yield values of E

JG in MegaJoules/meter.

Restart of Upstream Jet Grouting and Double Fluid Column S2-A7

Because of the critical nature of the project, the contract specifications required that the work had to be performed without inducing ground fracturing. Suspension of jet grouting was di-rected by the USACE for assessment of hydrau-lic fracturing and to develop techniques for con-trolling downhole pressures. Discussion with the contractor revealed no means to accurately measure downhole pressure in the columns nor could they demonstrate that triple fluid jet grouting with increased erosive capability would not suffer blockage and cause hydraulic


fracturing. A revised plan of action for spoil blockage allowed a loss of spoil return for 30 seconds before action to remove the blockage, e.g., stroking the hole with the jet grout drill rod, had to be undertaken. Upstream jet grout-ing resumed Friday, 14 July 2006 with double-fluid column S2-A7. During jet grouting of column S2-A7, air bubbles again were observed in the reservoir just upstream of the S2-A7 location. The air bubbles occurred shortly after spoil return became intermittent. Af-ter observation of the air bubbles, jet grouting was terminated and the hole grouted. Spoil return was regularly lost for as much as 10 seconds but never for 30 seconds. Thus the plan of action was not implemented before air bubbles were observed in the res-ervoir because the maximum blockage time was less than 30 seconds. This necessitated development of another response plan that would not allow hydraulic fracture before remedial ac-tions were implemented.

The experience of S2-A7 where lack of spoil return was not sufficient to trigger the action plan was troubling because it provided no means for de-tecting the onset of possible fracture until the presence of air bubbling was present in the reservoir and potential damage had already occurred.

Second Restart of Upstream Jet Grouting and Triple Fluid Column S2-A3B

Vibrating wire piezometers and open tube devices were installed to monitor

the effects of jet grouting during a moratorium period of about two months to determine lon-ger term effects of the observed air release. No adverse conditions were noted, and jet grouting for additional upstream parametric columns was resumed on 19 September 2006 with the vibrating wire piezometers and open tube devic-es providing real-time monitoring of jet grout-ing pressures. Some of the adjustments made before jet grouting resumed are use of a larger annulus, i.e., use a 254 mm (10 inch) casing and 228.6 mm (9 inch) drill bit so the grout annulus was reduced from 25 mm to 12.5 mm (1 inch to 0.5 inch), starting air and fluid circulation in

[TABLE 2] Triple fl uid jet grouting parameters for upstream parametric column S2-A3 and Paso de las Piedras Dam

Jet Grout

Parameters

Tuttle Creek Dam S2-A3 in Foundation

Sands

Paso de las Piedras Dam

Primary Columns in Silt and Sands

Air Pressure (bars) 9 12 - 20

Air Pressure (MPa) 0.9 1,200 – 2,000

Air Pressure (psi) 130.5 174.0 – 290.1

Air Flow Rate (liter/min) 7,700 12,000 –18,000

Air Flow Rate (m3/hour) 462.0 12,000 –18,000

Water Pressure (bars) 440 450

Water Pressure (MPa) 44.0 45,000

Water Pressure (psi) 6,381.7 6,526.7

Water Flow Rate (liter/min) 341 170

Water Flow Rate (m3/hour) 20.5 170

Grout Pressure (bars) 300 200 – 220

Grout Pressure (MPa) 30.0 20,000 – 22,000

Grout Pressure (psi) 4,361.1 2,900.8 – 3,815.8

Grout Flow Rate (liter/min) 448 180

Grout Flow Rate (m3/hour) 26.9 180

Withdrawal Rate (m/hour) 8.4

Specific Energy (MJ/m) 156 32 - 71

Nominal Column Diameter (ft) Minimum 8.5 5.2 – 7.9

Nominal Column Diameter (m) Minimum 2.6 1.6 – 2.4

[FIG. 4] Photograph of air bubbling in Tuttle Creek Reservoir during jet grouting of upstream column S2-A3 with reservoir turbidity curtain in background


the casing instead of at the bottom of the bore-hole, and starting jet grouting at the column top and moving to the bottom instead of starting at the bottom and moving to the top. This al-lowed better evacuation of spoil by not operat-ing under a full column of spoil, especially near the top of the column after starting. The triple fluid system utilized a drill rod diameter of 114 mm (4.5 inches) which created an annulus of 57.2 mm (2.25 inches), instead of 44.5 mm (1.75 inches) if the 12.7 mm (0.5 inch) wide annulus of hardened grout in the casing was still pres-ent. The first three jet grout columns, S2-A3B, S2-A6B, and S2-A4B, were completed on 19, 20, and 21 September 2006, respectively, with-out a visible air release into Tuttle Creek Reservoir.

Table 3 provides a comparison of the jet grouting parameters used at Paso de las Piedras Dam and up-stream triple fluid column S2-A3B, which did not experience reservoir air bubbling. The triple-fluid system used at Tuttle Creek involved grout pressure and water and grout flow rates that exceeded those used at Paso de las Piedras Dam. However, the air pressure and flow rate used at Paso de las Piedras did exceed those at Tuttle Creek which may explain the presence of air bubbling in the reservoir at Paso de las Piedras and not during S2-A3B. The biggest dif-ference between these two projects is the contractor proposed a minimum column diameter of 2.6 m (8.5 ft) for Tuttle Creek even though it had only constructed 1.6 to 2.4 m (5.2 to 7.9 ft) diameter primary columns for the cutoff wall at Paso de las Pie-dras Dam. This increase in column diameter requires a larger volume of material to be evacuated through the borehole annulus which may be problematic for achieving good spoil return. In addition, higher energies would be required for the second-ary columns because higher energies were required for the secondary col-umns at Paso de las Piedras Dam due to densification that occurred during primary column construction (38 to 100 MJ/m instead of 32 to 71 MJ/m) (8543 to 22481 ft-kips per ft instead

of 7194 to 15962 ft-kips per ft). It was encour-aging that no air bubbling occurred during the first three jet grout columns even though some of the grouting parameters exceeded those at Paso de las Piedras Dam, which lead to some optimism that subsequent jet grouting could continue safely at Tuttle Creek.

Jet Grouting of Upstream Double Fluid Column S2-A7B

Unfortunately during double fluid column S2-A7B on 22 September 2006, air bubbles again occurred in Tuttle Creek Reservoir. This air release occurred in close proximity to the air release that occurred during double fluid S2-A7 and occurred within about 10 seconds

[TABLE 3] Triple fl uid jet grouting parameters for upstream parametric column S2-A3B and Paso de las Piedras Dam

Jet Grout

Parameters

Tuttle Creek Dam S2-A3B

Paso de las Piedras Dam

Primary Columns in Silt and Sands

Air Pressure (bars) 12 12 – 20

Air Pressure (MPa) 1.2 1.2 – 2.0

Air Pressure (psi) 174.0 174.0 – 290.1

Air Flow Rate (liter/min) 8,000 12,000 –18,000

Air Flow Rate (m3/hour) 480.0 12,000 –18,000

Water Pressure (bars) 440 450

Water Pressure (MPa) 44.0 45,000

Water Pressure (psi) 6,381.7 6,526.7

Water Flow Rate (liter/min) 412 170

Water Flow Rate (m3/hour) 24.7 170

Grout Pressure (bars) 290 200 – 220

Grout Pressure (MPa) 29.0 20,000 – 22,000

Grout Pressure (psi) 4,206.1 2,900.8 – 3,815.8

Grout Flow Rate (liter/min) 415 180

Grout Flow Rate (m3/hour) 24.9 180

Withdrawal Rate (m/hour) 6.8

Specific Energy (MJ/m) 245 32 – 71

Nominal Column Diameter (ft) Minimum 8.5 5.2 – 7.9

Nominal Column Diameter (m) Minimum 2.6 1.6 – 2.4


of spoil return loss. Table 4 presents the jet grouting parameters for upstream parametric column S2-A7B.

Significant pressure, probably air and grout, developed in an adjacent vibrating wire piezom-eter (P-65-3) during drilling for column S2-A7B. Piezometer P-65-3 also exhibited high pressures during jet grouting of column S2-A7B and cul-minated in the appearance of air bubbling in the reservoir shortly thereafter. Piezometer P-65-3 was tipped in the shale and limestone embank-ment material at elevation 312.6 m (1025 ft). Data from piezometer P-65-9 in boring 65-3 which is tipped in the foundation sand just be-low the natural fine-grained blanket at elevation 303.5 m (995 ft)) also showed a rapid increase and decrease in piezometric head from the equilibrium value of about Elevation 313.7 m (1028.5 ft) with initiation and termination of jet grouting.

Upstream Ground Fracture Mechanisms

A number of mechanisms were considered to explain the sustained air bubbling in Tuttle Creek Reservoir. These mechanisms are sum-marized below and include:

Air permeation into permeable foundation • sands, storage of large quantities of air, fol-lowed by air exiting through pre-existing/natural defects in the natural fine-grained soil blanket, and air entering the reservoir at or near the upstream toe of the dam.

Air permeation into permeable foundation • sands, storage of large quantities of air, fol-lowed by fracture of the natural fine-grained soil blanket by elevated air, water, and/or grout pressures, and air entering the reser-voir at or near the upstream toe of the dam.

Ground fracturing of the natural fine-• grained soil blanket and air entering the res-ervoir from the jet grouting.

The pre-drilling for the jet grout drill rod • for column S2-A3 puncturing the PVC cas-ing due to lack of verticality and causing an air release.

Lack of seal at the bottom of the ungrouted • steel casing in the shale and limestone em-bankment fill in column S2-A7 and S2-A7B, air traveling through pervious zones of the shale and limestone fill, and into the reser-voir along the upstream face of the dam.

The continuation of bubbling for at least 48 hours after jet grouting suggests that a large amount of air was stored and it is unlikely that the natural fine-grained blanket or shale and limestone rock fill (predominantly com-pacted clayey shale) under the reservoir, see Fig. 1, could store that large volume of air. In addition, the close proximity of the shale and limestone suggests that the air would have ap-peared quicker and more gradually if the leak occurred at the bottom of the steel casing used for column S2-A7 or through a punctured PVC casing for S2-A3. Finally, sonic drilling samples obtained from near the base of the S2-A7B cas-ing suggested the presence of granular material at or near the base of the steel casing in the shale and limestone. This also would facilitate movement of air to the reservoir which should have resulted in air appearing quicker and not continuing for over 48 hours after termination of jet grouting.

Air exiting from the permeable foundation sands through natural defects in the natu-ral fine-grained blanket also is not plausible because the reservoir had been filled to the MPP level since 29 April 1963. Thus, an ac-tive seepage condition had been occurring for

[TABLE 4] Double fl uid jet grouting parameters for upstream parametric column S2-A7B

Jet Grout

Parameters

Tuttle Creek Dam S2-A7B

Air Pressure (bars) 6

Air Pressure (MPa) 0.6

Air Pressure (psi) 87.0

Air Flow Rate (liter/min) 4,500

Air Flow Rate (m3/hour) 270.0

Grout Pressure (bars) 438

Grout Pressure (MPa) 43.8

Grout Pressure (psi) 6,352.7

Grout Flow Rate (liter/min) 503

Grout Flow Rate (m3/hour) 30.2

Specific Energy (MJ/m) 179

Nominal Column Diameter (ft) Minimum 8.5

Nominal Column Diameter (m) Minimum 2.6


about 43 years which would have filled any pre-existing/natural defects in the natural fine-grained blanket with sediment. More impor-tantly, pre-existing defects in the fine-grained blanket are not consistent with a 13.7 m (45 ft) hydraulic head loss that is currently occurring across the blanket based on piezometers mea-surements routinely made over the life of the dam. This mechanism also suggests that loss of spoil return has no consequences, when it is well known in the industry that loss of spoil return will result in a near instantaneous in-crease in bottom hole pressure and can easily cause hydrofracture.

Therefore, a plausible mechanism for air bub-bling in the reservoir during jet grouting for columns S2-A3 and S2-A7 is the accumulation of sufficient pressure with reduced spoil return, storage of large quantities of air in the perme-able and high void ratio foundation sands, and subsequent fracture of the natural fine-grained soil blanket that released the stored air for at least 48 hours after cessation of jet grouting. However, the most plausible mechanism for the air bubbling in the reservoir for double fluid column S2-A7B appears to be different and is discussed below. After blanket fracturing oc-curred, the stored air exited the foundation sands and entered the reservoir. The bubbling was able to continue for over 48 hours because of the large void space of the foundations sands allowed a substantial amount of air to be stored. It is likely that the reason air bubbling was not observed sooner in the reservoir is: (1) the distance air had to travel, from elevation 305.7 (1002.3 ft) or lower to the natural fine-grained blanket, (2) the time required for suf-ficient pressure to accumulate to fracture the natural fine-grained blanket, and (3) the time required to fracture and/or permeate the reser-voir sediment and/or the overlying shale/lime-stone fill. This assumes that the air traveled upstream of the impervious fill, which is rea-sonable given the hydraulic conductivity of the impervious fill and the additional vertical stress imposed by the thicker embankment. This scenario also explains air bubbles appearing in the reservoir shortly/immediately after spoil returned slowed in column S2-A7 because the air pressure of 1,200 kPa (174.0 psi) fractured the lower sands and nearby natural fine-grained blanket and shale/limestone fill had been previ-ously fractured by column S2-A3.

Additional evidence of ground fracturing during the upstream parametric columns was obtained via the subsequent subsurface investigation. The evidence consisted of core samples con-taining a grout fracture from a boring near up-stream jet grout column S2-A1. The arrows in Fig. 5 point to grout lenses found in the natural fine-grained blanket at a depth of 23.8 m (78 ft). This core sample was obtained at a ground sur-face distance of only 1.2 m (4 ft) from the cen-ter of jet grout column S2-A3. This ground frac-ture is significant because it is located down-stream of the column. This proved that ground fracturing can occur downstream (towards higher confining stress) as well as upstream of the jet grouting. Appearance of bubbles was not possible downstream of the jet grouting be-cause the surface of the work platform was dry. Hence, air release may have occurred but would have gone undetected by site personnel. Frac-tures created downstream of the cutoff wall can provide seepage paths through the fine-grained blanket after the wall is complete. These seep-age paths could lead to undesirable erosion and piping downstream of the cutoff wall. This is of critical importance because a completed cutoff wall could address the upstream fractures but cannot address downstream fractures. This scenario becomes increasingly dangerous at el-evated pool levels that are certain to occur in a flood control reservoir.

Additional evidence of upstream ground frac-ture included hardened grout being observed in the spoil return for nearby upstream jet grout column S2-A9B and sonic coring for other upstream parametric columns revealing grout and spoil outside of the completed column and downstream of the cutoff wall alignment. Some piezometers upstream of the jet grouting also may have grout in them as was observed in de-watering well No. 2 in the downstream test area described above.

The most plausible mechanism for the air bub-bling in the reservoir during jet grouting for column S2-A7B is different than the mechanism presented above for columns S2-A3 and S2-A7. This air release does not appear to be caused by accumulation of sufficient pressure, storage of large quantities of air in the foundation sands, and subsequent fracture of the natural fine-grained soil blanket that released the stored air for at least 48 hours after cessation of jet grouting. This mechanism appears to involve


an air release at the bottom of the ungrouted steel casing in the shale and limestone embank-ment fill, through the shale and limestone fill, and into the reservoir. Alternatively, air could have connected to the reservoir through a previ-ous fracture caused by jet grouting for columns S2-A3 and/or S2-A7. The mechanism is likely different than for columns S2-A3 and/or S2-A7 because the air release occurred shortly after jet grouting started and did not last for days after the jet grouting ceased as discussed below.

Dam Safety Implications and Cancellation of Jet Grouting at Tuttle Creek Dam

If the mechanism or cause of the reservoir air bubbling is fracturing of the fine-grained blan-ket, there could be detrimental consequences to Tuttle Creek Dam because new flow pathways could be created in an uncontrolled manner through this important seepage control feature. This can lead to increased seepage and hydrau-lic gradients in the foundation soils which can cause erosion and possibly piping of the foun-dation soils that would undermine the embank-ment. More importantly, the discovery of grout fractures downstream of the cutoff alignment indicates that fractures can occur downstream of the jet grouting. This could render the im-pervious zone embankment material and the natural fine-grained blanket less effective in re-sisting or reducing seepage over the top of the cutoff wall.

Reducing the effectiveness of the impervious zone would result in seepage directly through the blanket downstream of the cut-off wall and higher gradients acting across the blanket than with the up-stream impervious zone being pres-ent. This is significant because of the variability of the natural fine-grained blanket across the valley. This vari-ability was mitigated during design by the extended impervious zone so damage to the impervious zone could result in greater seepage and hydrau-lic gradients in the foundation sands. In addition, the natural fine-grained blanket could be damaged directly by jet grouting because grout was found downstream of the upstream para-metric columns in the fine-grained blanket (see Fig. 5). The potential for ground fracture is about the same up-stream and downstream of the cutoff

wall because the work platform was essentially flat so the difference in vertical effective stress between upstream and downstream is small. The biggest difference between the upstream and downstream sides of the cutoff wall is the downstream is slightly more overconsolidated because of the previous greater height of the embankment in the work platform area, thereby increasing the horizontal effective stress.

Potential fractures or seepage paths could develop along the downstream side of the wall and through fractures in the natural fine-grained blanket and/or the shale and limestone fill downstream of the wall. These scenarios would result in higher gradients occurring in the foundation sands which also could result in unacceptable hydraulic gradients and flow at the downstream relief well system. The po-tential for seepage along the downstream side of the wall appeared extremely likely because previous jet grout experience suggests that the completed columns will serve as pressure relief points by allowing pressure to escape alongthe sides of the completed column. If com-pleted columns were to serve as pressure relief points, a flow path could be created along the downstream side of a completed column which would allow the reservoir head to enter the foundation sands along the downstream side of the completed columns. Piezometer measure-ments made prior to jet grouting show a head loss in the shale and limestone fill so fracturing of this material would allow a higher gradient

[FIG. 5] Grout fracture in the core for column S2-A1 at a depth of 23.8 m (78 feet) (arrows indicate grout found in the embankment materials)


to act at the top of the fine-grained blanket than was present prior to jet grouting.

Because of these dam safety risks and the in-ability to control these risks, the USACE decided to abandon all upstream jet grouting which meant terminating the upstream cutoff wall and slope stabilization. This reduced the remedial work to only downstream slope stabilization us-ing a non-jet assisted technique. The option of smaller diameter columns and additional rows (allowed in the contract) were viable options, but not options the contractor adopted due to cost concerns. The smaller diameter columns would have reduced the fracture potential by reducing the pressures required but these pres-sures still were probably high enough to cause ground fracturing. Downstream stabilization involved construction of transverse shear walls using a self-hardening cement-bentonite slurry. The transverse shear walls were excavated using a clamshell device mounted on a crane (see Wal-berg et al., 2012).

DAM SAFETY RISKS IN JET GROUTING PROCESSThis section discusses some of the risks of jet grouting in or below an operational dam, such as lack of spoil return and the buildup of down-hole pressures. At present, there appears to be no viable means in the industry for measur-ing downhole pressures, detecting the onset of ground fracture, or detecting grout outside of the completed columns to determine the extent of grout migration. This section briefly discusses each of these risks and deficiencies in current industry practice.

This case history shows that even if spoil • return is maintained, high and detrimental ground pressure can still develop and cause ground fracture.

At present, the authors are not aware of a • usable device, in-rod instrumentation, or technique for accurately measuring down-hole pressure which allows selection and/or adjustment of appropriate jet grout parameters so ground fracture does not oc-cur, or at least would be recognized when it was occurring.

The ability to locate grout migration is dif-• ficult using standard drilling techniques because of the small size of the sample and the small amount of grout being sought. In

addition, the more problematic fractures are ones that are not filled with grout and would go undetected if the investigative means consist only of borehole drilling.

RECOMMENDATIONS FOR FUTURE JET GROUTING AT OPERATIONAL DAMSThis section discusses some recommendations for constructing a jet grout cutoff wall in dams with an operational reservoir.

Control Spoil Blockage

First, it is important to control/manage spoil blockages and poor spoil return to reduce the potential for ground fracture. Brill et al. (2003) present some methods for controlling spoil re-turn but the keys to continuous spoil return ap-pear to be an adequate annulus to evacuate the required material, effectively disaggregating the native soils, and low viscosity spoils. It may be possible to reduce spoil viscosity by adjusting the jet grouting parameters/ methodology or by a precutting technique. A precutting process was used for the Posey Tube project near Oak-land, California (Lee et al. 2005) because of dif-ficulties in obtaining good, uniform columns in the stiff native clay layers. Through many test columns the contractor found that precutting was required to detach and disaggregate the clayey materials to facilitate spoil return and allow thorough mixing with grout. Lee et al. (2005) state that to enhance the consistency and structural integrity of the production columns, the contractor “remixed” the new columns by re-inserting the drill rod to the bottom of the columns and re-grouting the columns upward. The first phase would involve precutting with water or a dilute grout followed by a second phase of grout injection from the bottom of the column upward.

Other jet grouting recommendations for im-proving spoil return and reducing the potential for spoil blockage and ground fracture include:

Increase hole diameter/annulus as large • as possible.

Install fragmenting PVC drill casing through • lower clay zones as suggested by Sembenelli and Sembenelli (1999) to increase the poten-tial for the hole to remain open and remove constrictions.

Staring erosion at the top of the column and • continuing to the bottom so top-down ero-


sion as well as bottom-up erosion and jet grouting occurs.

Adjust jet grouting parameters and methods • to decrease spoil density and viscosity

Increase station time to increase erosion • and disaggregation of in situ materials

Initiate air flow and pressure with the moni-• tor still in the casing so elevated air pres-sure is not required to overcome the column head in the borehole and the tendency for cuttings to settle into the boring while the drill rods are idle, e.g., during verticality measurements.

Install pressure relief wells, i.e., a passive • response system, near jet grouting to relieve some pressure to reduce the potential for wide spread ground fracture assuming a high insitu hydraulic conductivity. However, relief wells most likely will not be able to prevent localized ground fracturing.

Develop a response plan for blockages and • instrumentation prior to jet grouting to prevent or limit the amount of ground frac-ture. The response plan should recognize that poor spoil return and temporary block-ages can lead to elevated bottom hole pres-sures occurring before the response plan is implemented. As a result, the response plan must have early triggers to change grouting methodology so large pressures and ground fracturing do not occur. This response plan also should include criteria, e.g., an unac-ceptable increase in pressure in adjacent piezometers, for requiring the contractor to “stroke the hole” with the jet grouting drill rod to re-establish suitable spoil return and reduce bottom hole pressures. If high pres-sures or poor spoil return continue, the plan should require a change in jet grout meth-odology to address these conditions.

Measure and control downhole pressures

It is also recommended that a device or tech-nology be developed for measuring downhole pressures so the onset of ground fracturing can be monitored which is important for a dam with an operational reservoir. This device is desired because contract language for the Tuttle Creek project that stated “ground fracture shall not occur” was not sufficient. This is because ground fracture likely occurred even though precautionary measures were undertaken. It is anticipated that some in-rod instrumentation

could be developed to measure the pressure outside of the monitor during jet grouting. The instrumentation could send the pressure mea-surements to the surface via cable or possibly by wireless technology. If measuring downhole pressure is not possible, it is recommended that a correlation between downhole pressure and acceptable spoil return/properties be de-veloped so downhole pressures can be indi-rectly estimated from observable spoil return. Kauschinger (2006) describes such a device that monitors borehole pressure during jet grouting at one or more locations in the borehole. The device would notify the jet grouting operator when there is a risk of soil fracture because the borehole pressure exceeds a predetermined lim-it or predetermined rate of increase, e.g., 69 kPa (10 psi) in 10 seconds (Kauschinger, 2006). It is important that the pressure required for ground fracture should be agreed upon before the onset of jet grouting so this point is not a source of confusion/contention if indications of ground fracture develop.

Jet grouting impact on seepage control elements

Given the difficulties measuring downhole pres-sures, maintaining adequate spoil return, and identifying the onset, location, and extent of ground fracture, some technology should be de-veloped for determining the integrity of existing seepage control elements, e.g., the natural fine-grained soil blanket, upstream impervious ex-tended zone material, and the clay core, before, during, and after jet grouting. This technology should monitor fracturing/damage, if any, to the seepage control elements during grouting and estimate the continued effectiveness of the elements as jet grouting production proceeds. One possible technology for accomplishing this objective is mapping and monitoring of subsurface water flow which may provide an insight into flow patterns and changes in flow patterns after the initiation of jet grouting. This is accomplished by energizing a water-bearing zone, e.g., the natural fine-grained blanket and/or impervious zone material, with an AC cur-rent. As the AC current flows through the water a magnetic field is generated and is measured at multiple points on the ground surface using patented technology. These data are used to prepare contour maps of subsurface flow that can be used to determine if changes in flow, i.e., ground fractures, have occurred as a result on the jet grouting.


Finally, a double fluid system may be more desirable for jet grouting with an operational reservoir than a triple fluid system because if ground fracture does occur, it may be sealed with either a cement based spoil or a grout fluid. The triple system uses a pair of opposing water jets shrouded with air for ground erosion and another jet for injecting the necessary ce-ment grout or other binder that provides the long term durability and low hydraulic conduc-tivity in the soil. In a double system the erosion is caused by two opposing jets of grout also shrouded with air. In principle the triple sys-tem may offer less risk of ground fracture but fractures may partially or fully close before the grout jet can inject cement or binder to seal the fracture. Conversely, the double fluid system causes erosion by the grout jet so if ground fracture does occur a cement/binder would have a better opportunity to fill the fracture(s) and seal it than with the triple fluid system. Control and sealing of fractures is an important consideration given an operational reservoir, the cutoff wall usually being located on the upstream side of the dam, and the impervious core being in close proximity to the embank-ment material.

CONCLUSIONS AND RECOMMENDATIONSJet grouting has seen increased usage in ground improvement efforts to address strength and compressibility of soils. The practice has be-come popular because of its relative ease of installation (compared to excavations) and its versatility. Recently its use has extended to seepage control and/or strengthening of dams with an operational reservoir. Results of recent jet grout projects involving dams suggest that the process may create some dam safety risks including ground fracture, which can lead to un-desired seepage pathways and hydraulic gradi-ents in erodible materials, inadequate securing of the cutoff wall or seepage control measure to the impervious embankment material, and variability in column diameter and consistency. Of course the largest risk is the potential for ground fracture due to the high air, water, and grout pressures required to create large diam-eter soilcrete columns.

This paper presents recommendations for re-ducing dam safety risks including control and maintenance of spoil return, measuring/moni-toring of downhole pressures, and estimating

the effect(s) of jet grouting on the existing seep-age control measures in the dam. Given the large consequence of failure, the impact of jet grouting on the embankment and foundation materials should be understood when an opera-tional reservoir is present.

REFERENCESBrill, G.T., Burke, G.K., and Ringen, A.R. 1. (2003). “A Ten Year Perspective of Jet Grouting: Advancements in Applications and Technology, Grouting and Ground Treatment”, Proceedings of Third International Conference on Grouting and Ground Treatment, ASCE Geotechnical Special Publication No. 120, New Orleans, pp. 218-235.

Burke, G. (2007). “Vertical and Horizontal 2. Groundwater Barriers using Jet Grout Panels and Columns”, Geotechnical Special Publication No. 168, Geo-Denver, Denver, Colorado, ASCE.

Bustinza, J.A., Pujol, A.O. & Schefer, J.C. 3. (1999) "Paso de las Piedras Dam - Repair of foundation problems." International Symposium on Dam Foundation: Problems and Solutions. Antalya, Turkey.

Croce, P., and Modoni, G. (2007). Design 4. of Jet Grout Cut-off Walls. Ground Improvement, 11(1): 11-19.

Fang, Y.S., Kao, C.C., Chou, J., Chain, K.F., 5. Wang, D.R., and Lin, C.T. (2006). Jet Grouting with the Superjet-midi Method. Ground Improvement, 10(2): 69-76.

Kauschinger, J., (2008). "Methods and 6. systems for monitoring pressure during jet grouting," U.S. Patent No. 7,455,479, Application No. 11/456,130, www.uspto.gov.

Lane, K.S. and Fehrman, R.G., (1960). "Tuttle 7. Creek Dam of Rolled Shale and Dredged Sand," Journal of Soil Mechanics Division, ASCE, Vol. 86, No. SM6, December, pp. 11-34.

Lee, T.S., Murray, R., and Kiesse, M. (2005). 8. “Jet Grouting at Posey Tube, Oakland, California”, Geotechnical Special Publication No. 136, Geo-Frontiers Conference, Austin, TX, ASCE, pp. 1-15.

Martin, J.R., Olgun, C.G., Mitchell, J.K., and 9. Durgunoglu, H.T. (2004). High Modulus Columns for Liquefaction Mitigation. Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 130(6): 561-571.


Pagliacci, F., Trevisani, S., and Chong, L. 10. (1994). “Recent development in jet grouting techniques – Singapore Pulau Seraya power station, a case history.” Proceedings 3rd International Conference on Deep Foundation Practice, Singapore, CI-Premier, 219-225.

Rattue, A. (2005). “Innovation to Tradition, A 11. Matter of Time,” Keynote address, Canadian Dam Association Conference, October 3, 2005, Calgary, Alberta http://www.cda.ca/proceedings%20datafiles/2005/2005-00-01.pdf

Rice, J.D. and Duncan, J.M., (2009a). 12. “Findings of Case Histories on the Long-Term Performance of Seepage Barriers in Dams,” Journal of Geotechnical Engineering, ASCE, 136(1), 2-16.

Rice, J.D. and Duncan, J.M., (2009b). 13. “Deformation and Cracking of Seepage Barriers in Dams due to Changes in the Pore Pressure Regime,” Journal of Geotechnical Engineering., ASCE, 136(1), pp. 16-25.

Schlosser F (1997). “Soil improvement and 14. reinforcement.” Proc. 14th International Conference on Soil Mechanics and Foundation Engineering, Hamburg, ISSMGE, 2445-2466.

Sembenelli, P.G. and Sembenelli, S. (1999). 15. “Deep Jet-Grouted Cut-offs in Riverine Alluvia for Earthen Cofferdams,” Journal of Geotechnical and Geoenvironmental Engineering, ASCE, 125(2), February, pp. 142-153.

Stark, T.D., Axtell, P., Lewis, J.R., Dillon, 16. J.C., Empson, W.B., Topi, J.E., and Walberg, F.C. (2009). "Soil Inclusions in Jet Grout Columns," DFI Journal, Deep Foundations Institute, Vol. 3, No. 1, May, pp. 44-55.

Treviicos South. (2006). 17. Tuttle Creek Dam Foundation Modification Project, Manhattan, KS, Test Program Construction (Option), Final Report.

Treviicos South. (2007). Personal 18. communication and PowerPoint presentation.

Walberg, F.C., Stark, T.D., Nicholson, P.J., 19. Castro, G., Byrne, P.M., Axtell, P., Dillon, J.C., Empson, W.B., Topi, J.E., Mathews, D.L., and Bellew, G.M. (2012). “Case Study: Seismic Retrofit of Tuttle Creek Dam,” accepted for publication in ASCE Journal of Geotechnical and Geoenvironmental Engineering (in press).

Yilmaz, D., Babuccu, F., Batmaz, S., and 20. Kavruk, F. (2007). “Liquefaction Analysis and Soil Improvement in Beydag Dam”, Geotechnical and Geological Engineering, 26(2): 211-224.


Challenges and Uncertainties Relating to Open CaissonsFathi Abdrabbo, Professor, Structural Engineering Department, Faculty of Engineering, Alexandria

University, Alexandria, Egypt; [email protected]

Khaled Gaaver, Associate Professor, Structural Engineering Department, Faculty of Engineering,

Alexandria University, Alexandria, Egypt; [email protected]

ABSTRACTOpen caissons are used for many geotechnical engineering applications. Open caissons may be used as deep foundation elements bypassing weak soils to tip in firm deeper strata, and in rivers and maritime construction to reduce the risk of scour. Open caissons are also used for collecting sewage water through gravity sewer pipe networks or from sewer force mains. In such applications, the design and construction of open caissons require a detailed soil investigation program. In this way, the design and construction plan of an open caisson can be developed with full knowledge of the prevailing subsoil conditions. The engineering and construction techniques are key factors to achieve functional caissons. Based on close observations during construction stages, the current study presents some challenges that were encountered during the construction of two open caissons of internal diameters 20 m and 10 m (65.6 ft and 32.8 ft). This paper describes the procedure followed to alleviate the construction difficulties encountered. Site exploration program and control measures required to satisfy design and construction requirements are crucial aspects. Sinking of open caissons in dense or very dense sands is risky. Incorrect sinking of open caissons may cause extra cost, delay in construction, and harm to nearby structures. Air/water jetting near the cutting edge of an open caisson, outside slurry trench, and/or inside open trench may be used to drive an open caisson downward. Unsymmetrical work around an open caisson may lead to tilting of the caisson. If this occurs, the tilt should be immediately corrected before resuming the sinking process. Improper cleaning of fine materials on the caisson’s excavation bed, and/or inappropriate pouring of underwater concrete may result in a defective concrete seal. The paper contains a series of practical guidelines to assist those intending to use open caissons, and shares good caisson sinking practice with practitioners. Finally, the study aims to understand the difficulties encountered and to anticipate future problems.

INTRODUCTIONSinking of open caissons is appropriate where the prevailing soil consists of soft to medium clays, silty sands, or loose sands. These soils can be readily excavated using grab buckets within the open caisson and do not offer high skin friction along caisson-soil interface. Open caissons can feasibly extend to great depth at relatively low cost; however, they have some disadvantages, Tomlinson (1986). For example, construction may be halted if obstructions, such as large boulders or tree trunks, are encountered. The available literature is scant regarding the sinking of open caissons because diaphragm trenches and large-diameter secant piles are implemented in the construction of open caissons (Puller, 1996). Moreover, the use of suction caissons is considered as an alternative construction method. Contrary to open caissons, the penetration of suction caissons into soil is due to self-weight of shaft in addition to suction pressure created inside the caisson. Therefore,

adequate seal provided by caisson’s self-weight penetration is essential to apply suction pressure inside the caisson. Suction caissons attracted the attention of many authors, including Chen & Randolph (2007), Byrne & Houlsby (2004), Clukey et al. (2004), Iskander et al. (2002), Randolph & House (2002), and El-Gharbawy & Olson (1999).

The construction procedure of open caissons may differ between countries (Allenby et al. 2009). Depending on experience gained in Egypt, open caissons are constructed using consecutive lifts of reinforced concrete cast-in-situ walls. The caisson walls sink in place successively while the soil inside the caisson is excavated using grab buckets. Upon reaching the design depth level, a concrete seal is cast underwater using tremie pipes. After the concrete seal has matured, water inside the caisson is pumped out. The caisson can be used either as is for collecting sewage water or filled with concrete as a deep foundation (Nonveiller 1987).


Open caissons have two essential stages to be carefully studied, construction stage and in-service stage. During the in-service stage, most of the applied loads on the caissons are transferred to the soil via end bearing at their bases. The contribution of skin friction developed along the caisson-soil interface may be ignored to support superstructure loads in certain circumstances. In contrast, skin friction along the caisson-soil interface should be precisely estimated during the construction stage to check for feasibility of sinking the caisson through the soil. The uncertainties arising during the calculation of skin friction along the caisson-soil interface may be attributed to the disturbance of adjacent soil due to the construction process. Ranges of skin friction values have been provided by Terzaghi and Peck (1967) for each type of soil. Similarly, values of estimated skin friction during sinking of both pneumatic and open caissons in different soil conditions have been reported by Tomlinson (1986). Puller (1996) mentioned that a comparison of the values recommended by Terzaghi & Peck (1967) with those given by Tomlinson (1986) shows a considerable scatter of skin friction in similarly described soils.

The methodology used in this study is based on the observational procedure, which is one of the design approaches listed in Eurocode 7 (Glass & Powderham 1994). Peck (1969) pioneered the applications of the observational method to geotechnical engineering. The philosophy behind the observational technique is to initially base the design on whatever information can be obtained and then to examine all conceivable differences between assumptions and reality. The observational method saves cost and time, and limits construction risks. Nowadays, the observational method is well known to the geotechnical profession (Wu 2011). The method was used to explore some challenges during the construction stages of two open caissons. In this paper, two case studies were considered.

CASE STUDY NO. 1This case study presents some difficulties encountered during sinking of an open caisson of

20.00 m (65.6 ft) internal diameter. (Caisson PS4). The caisson was designed to collect sewage water through a sewer network at El-Agamy district, west of Alexandria, Egypt. Caisson PS4 is located approximately 400 m (1312 ft) from the shore of the Mediterranean Sea. Service and residential buildings are located near the caisson site. Five pumps were to be installed in the caisson to pump sewage water at a rate of 5000 liter/sec (1320 gal/sec) to a treatment plant located 20 km (12.4 miles) away. Caisson PS4 is the main pump station of the sewer network of the district, see schematic diagram, Fig. 1. The total height of the caisson was 33.31 m (109.3 ft). It is common practice to reduce the wall thickness above the cutting shoe by 30 to 100 mm (1.2 to 3.9 in) from the outside to reduce skin friction along the caisson walls. Therefore, the wall thickness is 1.60 m (5.25 ft) for the top 27.31 m (89.6 ft) height and 1.70 m (5.6 ft) for the bottom 6.00 m (19.7 ft), as shown in 2. The elevation of the ground surface at the construction site is +2.30 m (+7.6 ft) above sea level. The designed upper floor level of the caisson is at +3.15 m (+10.3 ft), while the tip level is designed to be at -30.16 m (-99.0 ft). The wall is reinforced vertically using 10 bars of 22 mm/m (#7 bars at 4 in centers) each side and hoop reinforcement using 8 bars of 22 mm/m (#7 bars at 5 in centers). At the designed sinking level, the dry weight of the caisson’s walls is 90.48 MN (10,170 ton), while the buoyant weight is 56.32 MN (6,330 ton).

[FIG. 1] Schematic diagram of caisson PS4 and emergency caisson


Prior to the start of construction, geotechnical investigations were conducted at the site by drilling three boreholes up to 50.00 m (164.0 ft) depth. The retrieved soil samples revealed six successive soil strata of similar thicknesses, as shown in Table 1. The subsoil exploration showed groundwater table at elevation -0.30 m (-1 ft). The data collected from the geotechnical investigation was used to design the caisson. The open caisson technique was recommended by the engineer as the most appropriate con-

struction method. The engineer’s decision was based on engineering judgment and construc-tability of the structure. The caisson was sunk successfully under its own weight according to the construction plan, while grabbing the soil inside to an elevation of -24.60 m (-80.7 ft). At this level, it was very difficult to further ad-vance the caisson. The caisson was obstructed; nevertheless the periphery walls were construct-ed to the full design height. Therefore, part of the caisson walls, 5.56 m (18.2 ft), remained above the ground surface. At this stage, the buoyant weight of the caisson walls was 63.58 MN (7,147 ton). Thus, the average skin fric-tion developed along caisson-soil interface was 32.29 kPa (4.68 psi). The estimated value is in good agreement with the lower limit value rec-ommended by Terzaghi & Peck (1967) and 1.42 times the value reported by Tomlinson (1986). Terzaghi & Peck (1967) mentioned values that vary from 33.50 to 67.00 kPa (4.86 to 9.72 psi) for dense sand, while Tomlinson (1986) report-ed a value of 22.80 kPa (3.31 psi) for sand.

After the caisson had been obstructed, four boreholes 50.00 m (164 ft) deep were drilled to further explore the subsoil difficulties en-countered. The recovered soil samples from the boreholes were classified according to ASTM D 2487 for soil and ASTM D 6032 for rock. Fig. 3 illustrates a typical borehole log along with the corresponding standard penetration tests (SPT), N-values. Retrieved soil samples from the boreholes revealed that the soil at the site consists of a top layer derived from oo-litic very poor, weak limestone extending from ground surface to a level that varied from -0.20 to -4.70 m (-0.66 to -15.4 ft) . This limestone was underlain by a layer of poorly graded very dense sand intermixed with pieces of sandstone of various sizes. At a level that varied from -15.40 to -16.70 m (-50.5 to -54.8 ft), a layer

comprising different soils of sandy silt with clay and silty clay with sand was encountered. This layer ex-tended to a level that varied from -33.30 to -34.70 m (-109.3 to -113.8 ft) and it contained two relatively thin layers of very poor weak sandstone. The thick-ness of the top sandstone layer varied from 0.35 to

[TABLE 1] Soil strata revealed from boreholes drilled prior to construction

Layer Top level (m) Bottom level (m) Soil classification

1 + 2.30 -12.70 Sandy silt

2 -12.70 -15.70 Sandstone

3 -15.70 -22.70 Sandy silt

4 -22.70 -33.70 Silty clay

5 -33.70 -36.70 Sandstone

6 -36.70 -47.70 Silty sand

[FIG. 2] Design details of open caisson PS4


3.00 m (-1.1 to -9.8 ft), while the bottom sand-stone was a discontinuous layer of 2.00 m (6.6 ft) maximum thickness. Groundwater table was measured at a level of -0.20 m (-0.66 ft).

The properties of sand in the second layer were interpreted based on the standard penetration test results and visual classification of the retrieved soil samples. Accordingly, the relative density of the sand was about 75%, and the unit weight was 18 kN/m3 (3093 lb/yd3). The corresponding angle of shearing resistance of sand is 40 degrees. In this situation, it is important to note that the interpreted properties of the second layer should be used with caution due to the effect of sandstone pieces on the SPT results. Core samples recovered from limestone in the first layer revealed that the recovery values varied from 25 to 40% while the rock quality designation was zero. Laboratory tests on rock samples including compressive strength were conducted. Undisturbed samples of cohesive soil in the third layer were tested in direct shear using shear box apparatus. The achieved results of the laboratory tests are presented in Table 2 and Fig. 3. The average value of undrained shear strength (Cu) of the cohesive soil is 22 kPa (3.2 psi) and the corresponding average undrained angle of shearing resistance is 13.60º. The average values were implemented in the stability analysis.

During the excavation process inside the cais-son, the second layer was visually observed and found to be weak sandstone, contrary to the laboratory classification of soil samples retrieved from the boreholes. The contradic-tion may be attributed to the disturbance of the retrieved samples by the sampling process. Block samples were recovered from the second layer during excavation inside the caisson. The unconfined compressive strength of the tested

[TABLE 2] Summary of soil properties obtained from post-construction explorations

Layer First Second Third

Top level (m) + 2.30 -0.20 / -4.70 -15.40 / -16.70

Bottom level (m) -0.20 / -4.70 -15.40 / -16.70 -33.30 / -34.70

Soil classification Limestone Sand/Sandstone Sandy silt/Silty clay

Natural unit weight (kN/m3) 18.20 18.00 15.80 – 16.90

Unconfined compressive strength (MPa) 1.94 0.45 – 1.37 -

Undrained shear strength (kN/m2) - - 18.00 – 26.00

Undrained angle of shearing resistance (Degrees) - 40.00 11.00 – 16.00

[FIG.3] A typical borehole in case study # 1


samples varied between 0.45 and 1.37 MPa (65.3 and 198.7 psi). Therefore, it is important to carefully consider soil types that reveal anoma-lously high N-values in SPT. Double or triple cores may be used to retrieve relatively undis-tributed rock samples.

Appreciable difference of the subsoil condition was observed when comparing the post-con-struction and the pre-construction explorations, especially above the level of -16.70 m (-54.8 ft). Soil investigations prior to construction showed a layer of sandy silt to elevation -12.70 m (-41.7 ft) overlaid sandstone of 3.00 m (9.8 ft) thick. In contrast, post-construction geotechni-cal explorations indicated weak limestone over-laid sandstone extending up to an elevation of -15.40 to -16.70 m (-50.5 to -54.8 ft). The differ-ence in soil types may be attributed to the pro-cedure of soil sampling. Washed soil samples using tricone do not represent the in-place state of soil formations, especially in rock layers. Also soil sampling using a split barrel in weak rocks leads to unrealistic soil descriptions and proper-ties compared to the real state conditions.

The post-construction boreholes clearly showed that the adopted method of sinking the open caisson was not the proper construction meth-od due to the following reasons:

There are high values of standard penetra-1. tion test results recorded with the geo-material in the second layer. Also, there are relatively high values of unconfined compressive strength of both limestone in the first layer and sandstone in the second layer as real state geomaterial conditions. The higher values of SPT are sufficient to provide an early indication that sinking of the caisson is not the proper construction technique.

There are irregularities in the thickness of 2. the sandstone layers, which adds difficul-ties to the sinking process and extends the excavation time. The existence of sandstone increases the frictional resisting force along the caisson-soil interface during the sinking process. Also, sandstone underneath the cutting edge of the caisson will not slump towards excavation inside the caisson and offers high bearing resistance.

Pieces of sandstone interbedded in sand 3. cause difficulties in excavation using a clam-

shell. Excavation of this type of geomaterial requires special trenching equipment.

The high level of groundwater table presents 4. additional difficulties to the underwater excavation process since excavation must be controlled by divers.

Due to these difficulties, the caisson stuck at the level of -24.60 m (-80.7 ft) for approximately 31 months. During this period, traditional meth-ods of sinking the caisson were used including pumping out water from the caisson using pow-erful surface pumps, and excavating geomaterial inside the caisson using vibratory excavators. Also excavation of soil outside the caisson up to 3.00 m (9.8 ft) depth was carried out to reduce skin friction on caisson-soil interface. The ex-cavation depth of soil outside the caisson was restricted to ensure stability of nearby struc-tures. Secant piles were designed and installed to maintain the safety of neighboring structures. Moreover, excavation beneath the cutting edge of the caisson walls was carried out by divers. All of these methods failed to move the caisson downward. It was a challenge to complete sink-ing the caisson by developing a reliable proce-dure to complete the sewage project.

In such circumstances, many factors control the adopted restoration procedure such as:

Legal liability and responsibility of the 1. owner/engineer or the contractor.

Expected cost of the proposed rehabilitation 2. procedure.

Expected time of the proposed restoration 3. technique.

The free space between the caisson wall and 4. the existing nearby structures is limited, thus the capability of this space to accom-modate the machinery involved in the resto-ration procedure must be considered.

Soil stratification and groundwater table.5.

The risk measurement of buildings nearby 6. the caisson.

In conclusion, due to the critical function of the caisson PS4 and the legal situation, it was impossible to backfill the obstructed caisson and to construct a new one nearby implement-ing a proper construction procedure. In addi-tion, relocation of the caisson would require rerouting of the approach inlet and outlet sewer


pipes, which had been installed already using a pipe jacking technique. Thus, complete sinking of the obstructed caisson by using a remedial technique was essential.

A restoration proposal was developed to con-struct a circular slurry trench outside the cais-son and as close to the wall as possible such that the drilling machine could accomplish the drilling process. The outside slurry trench was formed by constructing slurry piles, each 600 mm (23.6 in) in diameter, at a spacing of 1.00 m (3.28 ft) and extending to level -33.50 m (-109.9 ft). see Fig. 4. Each slurry pile incorpo-rated a steel tube of 50 mm (2 in) in diameter provided with an end nozzle. The tubes in-serted into the slurry piles are used to jet air at a pressure of 20 bars (290 psi), if needed. The slurry trench acts as a separator between the walls of the caisson and the geomaterial extend-ing further away from the caisson. The slurry trench reduces the skin friction along caisson-soil interface.

The construction of the slurry trench faced difficulties due to drilling through subsoil hard formation and the presence of saltwater from the nearby sea. The saltwater affected the performance of the slurry mud pumped inside the drilling holes for base and side stability. To overcome the effect of saltwater, the slurry mud mixture comprised 1.00 kg (2.2 lb) of sodium

[FIG. 4] Layout of the caisson and slurry piles

carbonate, 40.00 kg (88 lb) of bentonite, and 1,000 liters (264 gal) of water. Sodium carbon-ate was added to the slurry mud to overcome the effect of the sodium chloride and sulfate present in groundwater. An interior open trench inside the caisson and close to the caisson walls was excavated. The width of the inside trench is 3.00 m (9.8 ft) and extending to a level of -33.50 m (-109.9 ft). Thus it is believed that the outside slurry trench and the inside open trench bounded a geomaterial wall extending down from the cutting edge of the obstructed caisson up to a level of -33.50 m (-109.9 ft), see Fig. 5.

To study the stability of the geomaterial wall, which extended from level -24.60 to level -33.50 m (-80.7 to -109.9 ft), first we consid-ered section (a-a) at level -24.60 m (-80.7 ft), see Fig. 5. A trial plane of failure was considered to be inclined at an angle β (where β = 45 – φ/2 = 25°). The buoyant weight of 1.00 m (3.28 ft) length along the periphery of the caisson wall is 868 kN/m (59,478 lb/ft). Therefore, the driving force on the trial plane of failure is 366 kN/m (25,079 lb/ft) and the resisting force is 660 kN/m (45,225 lb/ft). This means that the geomaterial under the tip of the caisson was stable up to a level of -27.70 m (-90.9 ft), the top surface of the sandy silt/silty clay layer. The geomaterial wall sustained the imposed loads resulting from the caisson walls with-

All dimensions are in meters


out any slumping toward the excavated open trench inside the caisson. Then, consideration was given to the stability of a trial section (b-b) through the sandy silt/silty clay layer at a level of -31.00 m (-101.7 ft), see Fig. 5. The weight of the caisson wall and the enclosed geomate-rial up to a level of -31.00 m (-101.7 ft) was 933 kN/m (32 ton/ft). Thus the driving force on the trial plane of failure was 574 kN/m (19.7 ton/ft), whereas the resisting force was 225 kN/m (7.7 ton/ft). This analysis indicated that the geomaterial wall was unstable through the sandy silt/silty clay layer. Consequently, it was expected that shear failures occur due to excessive shear stresses imposed at a level of -31.00 m (-101.7 ft).

The process of sinking the caisson was resumed by blowing compressed air through the steel tubes installed in the slurry trench via a header pipe over a three day period. Air bubbles were observed to rise through the water inside the

caisson. This observation was a positive sign for the initiation of caisson sinking. Movement of the caisson was observed during injection of the compressed air. It was observed that the caisson moved vertically by 20, 50, and 100 mm (0.8, 1.0 and 2.0 in) in the three successive days respectively. Overnight, the caisson sank by 7.90 m (26.0 ft) and stayed at a level of -32.50 m (-106.6 ft).

According to the stability analysis of the geo-material wall, it was evident that failure of the geomaterial wall was expected at a level of -31.00 m (-101.7 ft). To reduce the risk level in the remedial procedure, the compressed air sys-tem was designed and installed in the outside slurry trench. The objective of the compressed air system was to overcome any shortcoming resulting from probable deviation of predicted outcome from actual performance. The predic-tion procedure may be deficient if one or more of the following is missed or deficiently predict-ed; soil stratigraphy, soil properties, subsoil het-erogeneity, prediction method and capability, stress history, and stress path (Focht 1994). The result of the completion of sinking the caisson while implementing the above procedure was that the caisson moved down to 2.34 m (7.7 ft) below the designed level. Therefore the top level of the caisson walls was at a level of +0.81 m (+2.7 ft), instead of +3.15 m (+10.3 ft). The cais-son walls were extended to a level of +3.15 m (+10.3 ft) by pouring concrete inside shuttering and scaffolding.

It is important to note that sudden sinking of the caisson had no side effect on the buildings located at a distance of 4.00 to 6.00 m (13 to 20 ft) from the caisson. This may be due to the gentle movement of the caisson. The caisson moved downward through a cohesive geomate-rial, which exhibited neither strain softening nor strain hardening. In other words, the geo-material behaved as an elastoplastic material such that the caisson sank smoothly. The move-ment of the caisson was monitored after com-plete sinking for 30 days and no movement was observed. At this stage, the dominant resisting force was the bearing stress developed at the caisson’s tip. The tip is at a distance between 0.80 to 2.20 m (2.62 to 7.2 ft) above sandstone layer. Thus sandstone contributed to the bear-ing capacity at the caisson tip. If friction along caisson-soil interface is ignored, the imposed bearing stress at the tip of the caisson is about

[FIG. 5] Stability analysis in case study # 1


518 kPa (75 psi). If half of the friction at the caisson-soil interface is considered, the im-posed bearing stress at the tip of the caisson is 162 kPa (23.5 psi). Simple analysis, using bear-ing capacity theories, shows that the caisson is stable at the level of -32.50 m (-106.6 ft).

The contractor resumed the work to complete the excavation inside the caisson and to ex-ecute the concrete seal. Tremie pipes of 200 mm (8 in) diameter were used to deliver the concrete from concrete trucks via a concrete pump to construct the concrete seal. After the concrete seal had been matured, it was discov-ered that the concrete seal did not function properly. Groundwater seeped through the seal when water inside the caisson was pumped out. The concrete seal was constructed without cleaning debris and soft deposits on the excava-tion bed. A layer of soft deposits was trapped under the concrete seal, and pockets of soft deposits intervened into the seal. As a result, the cast concrete seal was of poor quality and was thinner than the designed thickness. Ac-cordingly, the concrete seal became unable to resist stresses induced by uplift water pressure. Furthermore, it was observed by divers that cracks in the concrete seal developed more and more as pumping of water inside the caisson continued. Inadequate planning and improper execution of tremie concreting led to a defective concrete seal. Unfortunately, another remedy to an unexpected problem was needed.

The goal of the remedial work was to lower the groundwater table to beneath the toe level of the caisson to demolish the defective concrete seal and construct a new seal in dry condi-tions. The design of the remedial work required another borehole to explore soil stratification up to 100 m (328 ft) depth. The soil samples ob-tained revealed very poor weak sandstone bed extending from depth 50 m up to 100 m (164 ft to 328 ft) below the ground surface. The sand-stone is moderately weathered. As the caisson was 400 m (1300 ft) from the sea, additional im-pacts and uncertainties arose in the calculation of water seepage into the caisson. Moreover, another uncertainty arose from the permeabil-ity coefficient of the rock mass. The sandstone contained cracks and joints that were filled with fine material. Water channels were developed through cracks and joints in the sandstone with the progress of pumping water. Dewater-ing calculations were performed utilizing deep

wells outside the caisson to construct a new seal in dry conditions, but a large number of deep wells with large pump capacities were required. Furthermore, dewatering outside the caisson might have affected the stability of the adjacent buildings. Due to difficulties arising from pump-ing large amount of water to lower water inside the caisson to the level of -32.50 m (-106.6 ft) and side effect of dewatering on the adjacent buildings, a water cut-off wall was designed to be constructed around the caisson walls.

The choice of the geometry of the cut-off wall including material, thickness, and length is significant. Cut-off walls should have suffi-cient thickness to prevent hydraulic fracture. The depth of the cut-off wall should be deter-mined to prevent piping and heave in soil at the excavation bed and to reduce the seepage. One of the following materials was proposed to be used in the construction of the cut-off wall: soil-bentonite, soil-cement-bentonite, cement-bentonite, and plain concrete. The selected construction material was controlled by the design of the appropriate thickness to prevent hydraulic fracture in the cut-off wall. U.S. Army Corps of Engineers (USACE 1986) recommended that the minimum width of a soil-bentonite cut-off should be 0.03 m per 0.30 m (0.1 ft per ft) of differential hydraulic head. According to this criterion, the required width of the soil-benton-ite trench is excessively thick. Therefore, plain concrete was used to construct the cut-off wall with a thickness of 1.20 m (4 ft). It is known that the flow of water beneath impervious cut-off walls may produce heave or piping in soil. Heave occurs if the uplift force at the sheeting toe exceeds the submerged weight of the overly-ing soil. When the velocity of water at the exit exceeds the critical velocity of water, piping occurs. To prevent both heave and piping, the cut-off wall should be of sufficient depth below the excavation bed. Design charts provided by NAFAC DM-7 (1982) for two-dimensional flow were used to assess the penetration depth of the proposed cut-off wall. These charts were produced for homogeneous dense sandy soil including safety factor of 2. Interconnecting panels of diaphragm wall, which extended to the level of -87.70 m (-288 ft), were designed to form the cut-off wall. The clear space between the outer surface of the caisson walls and the inner surface of the cut-off wall was about 2.00 m (6.6 ft).


At this stage, the contractor faced another chal-lenge; it was difficult to obtain a diaphragm wall machine capable of operating at depths of up to 90 m (295 ft) below the ground surface. At that time, no suitable machine was locally available; therefore the equipment was imported from Europe. Another difficulty was that the free space between the caisson walls and the adja-cent buildings was insufficient to accommodate the imported diaphragm wall machine. Conse-quently, it was essential to backfill inside the caisson and the area around the caisson using structural fill to prepare a working platform for the diaphragm wall machine. After constructing the cut-off wall, the backfill inside the caisson was removed to complete the work.

To lower the groundwater table below the level of -32.50 m (-106.6 ft), seven deep wells of diameter 0.60 m (24 in) and extending to level -55.00 m (-180 ft) were designed and installed in the annular space between the caisson and the cut-off wall. An electric submersible pump of capacity 200 m3/hour (785 yd3/hr) at 60 m (197 ft) head was mounted in each well. During removal of the contaminated concrete seal, the concrete was observed to be inhomogeneous and containing soft spots. Additionally, free-from-cement aggregates were observed, indicat-ing that the concrete of the seal was washed out during concrete pouring. A large amount of fine material had been deposited at the lower surface of the defective seal. The defective seal was demolished and a new seal and reinforced concrete base of the caisson were constructed under dry conditions. The construction of the cut-off wall, dewatering process, and difficulties from the inaccurate interpretation of soil condi-tions, prior to construction, doubled the con-struction cost and increased the construction time to about five times the anticipated time.

Based on the presented case study, it can be concluded that improper interpretation of sub-surface ground conditions leads to inappropri-ate design of the caisson. The difficulties arising from inaccurate interpretation of soil conditions cause challenges that may increase the cost and time of construction. Improper cleaning of fine material deposited on the excavation bed, and/or incorrect procedures in pouring of underwa-ter concrete may produce an inadequate con-crete seal. Legal liability may divert the deci-sions of the engineer from adopting the proper procedure. Therefore, geotechnical engineers

are advised to avoid the use of open caissons in such circumstances. The alternative method is to use interconnecting panels of reinforced con-crete diaphragm walls to form the caisson walls and to grout the soil below the excavation bed before excavation inside the caisson to form an impervious blanket.

CASE STUDY NO. 2Case study No. 2 refers to the construction of a wastewater pump station at a village in El-Behera province, west of the Nile river delta, Egypt. An open caisson of 10.00 m (32.8 ft) internal diameter, 1.00 m (3.28 ft) wall thickness, and 12.00 m (39.4 ft) depth below the ground surface was considered for the station. The caisson is bounded by a store on the eastern side, 4.00 m (13 ft) from the caisson wall. On the southern side, there is a bank building 6.00 m (20 ft) from the caisson. Farmer buildings on the other sides are located approximately 15.00 m (49 ft) from the caisson. Two boreholes were drilled at the site up to 15.00 m (49 ft) below the ground surface. Fig. 6 presents a typical borehole log. The recovered soil samples from the boreholes were classified

[FIG. 6] A typical borehole in case study # 2


in accordance with ASTM D 2487. The subsoil is poorly graded sand with silt extending to 15.0 m (49 ft) depth below the ground surface. Groundwater table was encountered at a depth of 1.30 m (4.3 ft) below ground surface. Standard penetration test results indicated that the sand is medium dense throughout the top 5.00 m (16.4 ft) and became very dense below this level. Prior to the excavation process, the caisson walls were constructed to 12.00 m (39.4 ft) above the cutting edge.

After the concrete had matured, excavation inside the caisson walls began using a cable-suspended grab. Due to the excavation process and the body weight of caisson, the caisson sank up to 8.00 m (26 ft) below the ground surface. At this depth, the open caisson was stuck. The dry weight of the caisson walls was 10.37 MN (1165 ton). When the caisson sank 8.00 m (26 ft) below the ground surface, its weight reduced due to buoyancy force to 8.05 MN (905 ton). Thus, the average skin friction developed along caisson-soil interface was 26.70 kPa (3.87 psi). The average calculated value is less than the lower limit value recommended by Terzaghi & Peck (1967) for dense sand by 20%, which varied from 33.50 to 67.00 kPa (4.86 to 9.72 psi). Also, the average calculated skin friction is 17% greater than the value reported by Tomlinson (1986) of 22.80 kPa (3.31 psi). The uncertainties in skin friction at caisson-soil interface produce a high level of risk during caisson sinking.

The caisson was observed while its tip was at 8.00 m (26.2 ft) below the ground surface. It was found that the caisson had tilted southward by 0.4%. Ground loss of soil around the caisson was observed in the southern direction. Ground loss that occurred at one side of the caisson may have been due to heterogeneity of the subsoil condition, improper excavation process, and due to inclination of the caisson. The flow of soil inside the caisson can cause more tilting of the caisson. Ground loss may also damage nearby structures; therefore, the excavation process inside the caisson was stopped. At this stage, the soil inside the caisson was 1.00 m (3.3 ft) above the cutting shoe. It was evident that the friction resistance along caisson walls and the bearing resistance at caisson tip were greater than the body weight of the caisson.

To sink the caisson to the required depth, fric-tion resistance along the caisson-soil interface

needed to be deliberately decreased and the vertical orientation of the caisson needed to be corrected. Correction of the caisson vertical-ity became difficult as the embedded depth of the caisson increased. To decrease the friction at the caisson-soil interface, a slurry trench around the caisson walls was installed. Seventy holes were drilled around the caisson, and the drilled holes were filled with slurry mud. The slurry piles of diameter 400 mm (16 in) were extended to 15.00 m (39.2 ft) depth below the ground surface. The slurry mud consisted of 1:2 (bentonite: cement). Steel pipes of 50 mm (2 in) diameter were inserted in the holes for water or air jetting, if required. The pipes were pro-vided with slotted holes along the depth from 8.50 m to 15.00 m (28 ft to 49 ft). The caisson sank 0.40 m (1.3 ft) on completion of the slurry trench without inside excavation. Unfortunately, the caisson tilted by 4% due to unsymmetrical air jetting. Therefore, it was unreasonable to continue using air or water jetting to advance the downward movement of the caisson be-fore adjusting the tilt of the caisson. This was achieved by dewatering and excavating more intensively below the cutting edge at the higher side of the caisson than the lower side.

Based on the above discussion, it is not rea-sonable to sink open caissons in dense or very dense sands (N-Value for 300 mm (1 ft) ≥ 20). The values given by Terzaghi & Peck (1967) for skin friction on the caissons may be sensibly used to predict the friction resistance during sinking open caissons in sand. Incorrect sinking of open caissons may cause extra cost, delay in construction, and harm nearby structures. Moreover, uncertainties involved in the subsoil conditions and skin friction along caisson-soil interface impact risk measurements on the construction of open caissons. Unsymmetrical work around the open caisson leads to tilting of the caisson. The tilt should be corrected before resuming sinking process of the caisson. Fur-thermore, this case study demonstrates that improper construction of an engineering design, due to lack of knowledge and experience, may lead to further engineering problems that need to be rectified.

CONCLUSIONSThis paper presents some challenges that were encountered during the construction of two open caissons under two different subsoil


conditions. Procedures used to overcome the encountered construction challenges were described. The following conclusions may be drawn:

Proper interpretation of subsurface soil 1. conditions is a crucial aspect during design and selection of the proper technique for the construction of open caissons. Difficulties arising from the erroneous interpretation of subsoil conditions cause extra cost and delay in construction.

Sinking of open caissons in dense or very 2. dense sands (N-Value for 300 mm (1 ft) ≥ 20) is risky due to high friction resistance on caisson-soil interface.

Incorrect sinking of open caissons may 3. cause extra cost, delay in construction, and harm to nearby structures.

Air/water jetting near the cutting edge of 4. an open caisson, outside slurry trench, and/or inside open trench may be used to drive an open caisson downward.

A unique procedure to calculate skin 5. friction along soil-caisson interface does not exist, and the values recommended by Terzaghi and Peck (1967) may be sensibly used.

Improper cleaning of fine material 6. deposited on the excavation bed, improper pouring of underwater concrete, and improper interpretation of subsurface soil conditions caused some challenges to the open caisson in case study No. 1. These challenges doubled the construction cost and increased the construction time to approximately five times the anticipated time.

Unsymmetrical work around the open 7. caisson in case study No. 2 led to tilting of the caisson by 4% from the vertical. The tilt should be immediately corrected before resuming sinking the caisson.

Improper construction of an engineering 8. design may lead to further engineering problems that need to be remedied.

REFERENCESAllenby, D., Waley, G., and Kilburn, D. 1. (2009), Examples of open caisson sinking in Scotland, Proceedings of the ICE - Geotechnical Engineering, 162(1), 59-70.

Byrne, W. B., and Houlsby, G. T. (2004), 2. Experimental investigations of the response of suction caissons to transient combined loading, Journal of the Geotechnical and Geoenvironmental Engineering, ASCE, 130(3), pp. 240-253.

Chen, W., and Randolph M. F. (2007), Uplift 3. capacity of suction caissons under sustained and cyclic loading in soft clay, Journal of the Geotechnical and Geoenvironmental Engineering, ASCE, 133(11), pp. 1352-1363.

Clukey, E. C., Templeton, J. S., Randolph, M. 4. F., and Phillips, R. (2004). Suction caisson response under sustained loop current loads, Proceedings. 36th Annual Offshore Technology Conference, Houston, Paper No. OTC 16843.

US Army Corps of Engineers (USACE 1986), 5. Engineering manual EM 11110-2-1901. USA

El-Gharbawy, S. L., and Olson, R. E. (1999), 6. The cyclic pullout capacity of suction caisson foundations, Proceedings 9th International Offshore and Polar Engineering Conference, ISOPE’99, vol.2, France, pp 660-667.

Focht, J. A. (1994), Lessons learned 7. from missed predictions, Journal of the Geotechnical and Geoenvironmental Engineering, ASCE, 120(10), pp. 1653-1683.

Glass, P. R. and Powderham, A. J. (1994), 8. Application of the observational method at Limehouse Link, Geotechnique (44), 665- 679.

Iskander, M., El-Gharbawy, S. L., and Olson, 9. R. E. (2002), Performance of suction caissons in sand and clay, Canadian Geotechnical Journal, 39(3), pp 576-584.

NAFAC DM-7 (1982), Design manual 7.1, 10. Department of the Navy, Naval Facilities Engineering Command.

Nonveiller, E. (1987), Open Caissons 11. for Deep Foundations, Journal of the Geotechnical and Geoenvironmental Engineering, ASCE, 113(5), pp. 424-439.

Peck, R. B. (1969), Advantages and limitation 12. of the observational method in applied soil mechanics, Geotechnique, 19 (2), pp. 171-187.

Puller, M. (1996), Deep excavations - 13. a practical manual, Thomas Telford Publishing, London.


Terzaghi, K. and Peck, R. B. (1967), Soil 14. mechanics in engineering practice, 2nd edition, John Wiley, New York.

Tomlinson, M. J. (1986), Foundation design 15. and construction, Longman, Harlow.

Randolph, M. F., and House, A. R. (2002), 16. Analysis of suction caisson capacity in clay, Proceedings 34th Annual Offshore Technology Conference, Houston, Paper No. OTC 14236.

Wu, T. H. (2011), 2008 Peck lecture: The 17. observational method: case history and models, Journal of the Geotechnical and Geoenvironmental Engineering, ASCE, 137(10), pp. 862-873.


The Infl uence of RC Nonlinearity on p-y Curves for CIDH Bridge PiersLeonardo Massone, Assistant Professor, Department of Civil Eng., University of Chile, Santiago, Chile

Anne Lemnitzer, Assistant Professor, Department of Civil & Environmental Engineering, University

of California, Irvine, USA ; [email protected]

ABSTRACTThe p-y method is one of the most popular methods in pile design and has been calibrated for various boundary conditions using numerical and experimental studies during recent years. Most studies on reinforced concrete (RC) piles have included the impact of flexural nonlinearity, (e.g. nonlinear moment–curvature relations) but not considered associated pile shear deformations when deriving p-y curves from field data. Common p-y curves may be better applicable for piles with flexure dominated failures (e.g. piles with free- head boundary conditions). For piles with fixed head boundaries (i.e. rotation restrained piles) shear deformations could be of significant influence. To study this problem, a coupled shear flexure interaction model for axial-bending-shear behavior coded in OpenSees was applied to a 0.61 m (2 ft) diameter flagpole and a 0.61m (2 ft) diameter fixed head pile specimen to investigate the possible influence of shear deformations to the overall pile responses. The surrounding soil was represented by p-y curves derived from prior large scale testing on piles with similar boundary conditions. Analysis results show that for flagpole piles, shear forces and shear deformations were insignificant. Considerable contributions of pile shear displacements and forces were observed for the fixed head pile, with shear displacements contributing up to 40% of the total pile displacement. Results suggest that nonlinear shear deformations for reinforced concrete piles should be considered for fixed-head or similar conditions, and that currently used p-y curves may underestimate the actual lateral pile displacement and possibly lead to unsafe design for the particular boundary condition.

INTRODUCTIONCommon bridge foundation systems consist of large diameter drilled shafts or groups of relatively small-diameter bored or driven piles in various configurations. Typical design in the U.S. includes the use of reinforced concrete Cast-in Drilled Hole (CIDH) Shafts, which are also known as bored piles. Among the various models currently used by design engineers, the p-y approach has become the most popular method to estimate soil-pile interaction, load displacement relationships and ultimate pile capacities due to lateral loading such as earthquake shaking, tsunami impact or ship collisions. In the p-y analysis, the horizontal response of the soil is expressed through a series of non-linear springs, also called p-y springs, which represent the lateral resistance p of the soil coupled with the lateral displacement y of a pile into the soil at a given pile depth. The properties of these non-linear springs are defined through hyperbolic p-y curves, which have been originally proposed by McClelland and Focht (1958). In large scale experiments, p-y curves are traditionally obtained from pile

curvature measurements using sensors such as strain gauges (SGs), Linear Variable Displacement Transducers (LVDTs) or Fiber Optic Sensors (FOS). Average strain readings can first be converted into average curvature readings over the respective sensor gauge lengths and then translated into moment profiles using section specific moment curvature relationships. Double differentiation of pile moment profiles will yield the soil force profile p along the pile depth; double integration of pile curvatures will result into pile displacement profiles y. While free head boundary piles (also called flagpole piles) simply fail under single curvature yielding, as shown in Fig. 1, fixed head piles show a complex nonlinear failure mechanism, which is characterized by two possible yielding locations and higher moment gradients under lateral loading. This may lead to considerably higher shear forces, which are not yet addressed in common p-y studies. Most p-y relations derived using the experimental approach described above, integrate linear moment curvature (M-φ) relations or nonlinear beam-column models


proposed future experimental study explained in this paper. For piles with boundary conditions such as flagpole piles (free rotation at pile top), this interaction might be negligible and the failure under lateral loading will be flexure dominated.

The objective of this study is to use analytical tools to assess the potential impact of nonlinear shear deformations on p-y curves by assessing the relative contribution of shear and flexure deformations to pile displacement and force profiles over the pile length. To accomplish this, two different modeling options for reinforced concrete piles were used along with corresponding p-y curves to study the lateral load behavior of CIDH piles embedded in clayey soil. Both pile models account for nonlinear flexural behavior using a fiber section model; however, in one model shear deformations are neglected whereas in the other model they are considered using biaxial material relationships that allow for shear-flexure interaction at the section level. The shear flexure interaction model was originally developed by Massone and Wallace (2006) during experimental research on laterally loaded RC shear walls and was modified to investigate the response of two 0.61 m (2ft) diameter reinforced concrete shafts representative of a fixed head and a flagpole pile configuration. By considering these two extreme boundary conditions, the impact of shear deformations on experimentally calibrated p-y curves is studied by evaluating the response profiles over the pile height. The studies were performed using the finite element program OpenSees (http://peer.berkeley.edu/OpenSeesNavigator/) provided by the Pacific Earthquake Engineering Research center (PEER).

MODELING STUDIESInput parameters for the analytical model were selected based on pile tests conducted by Stewart et al. (2007) on bridge foundations

that do not include the interaction of shear and flexural forces or deformations. Examples of these are the most popular p-y curves proposed by the American Petroleum Institute (API, 1993). However it has been widely observed in experimental studies that often flexural yielding in structural members is accompanied by significant nonlinear shear behavior, even when the nominal shear strength is substantially larger (>2 times) than the shear demand (Oesterle et al. 1974, 1979; Hiraishi 1984; Massone and Wallace, 2004). For slender, flexure dominated walls in example, inelastic shear behavior was shown to contribute up to 30% of the total lateral displacements at the first level where most nonlinear flexural deformations are developed (Massone and Wallace, 2004). If similar behavior could be observed for RC piles, significant shear deformations contribute to the lateral pile displacement, which are not captured by any analytical models yet. This implies that current lateral pile design for fixed head reinforced concrete piles may be unconservative, not recognizing that much larger lateral deformations exist in the field due to internal pile shear deformations. Deformation due to such pile shear forces have been observed previously during a large scale experiment of a 9 pile RC fixed head pile group conducted by Lemnitzer et al. (2010). Fig. 2 shows a photograph after test completion and specimen excavation that clearly indicates large diagonal shear cracks in one of the RC piles of the group specimen. This observation sparked the interest to pursue the presented analytical and

[FIG. 2] Photograph of Diagonal Shear Crack

[FIG. 1] Bending Behavior of Free Head and Fixed Head Piles under Lateral Loading


near the LAX airport in Hawthorne, California. This was done to enable comparisons with experimental results, and also to verify the accuracy of the analytical model developed for this study. The experimental tests were designed based on CALTRANS’ Seismic design criteria (SDC 2006) who represented the sponsor of the field test program. The pile test specimens were subjected to quasi static reverse cyclic loading up to ultimate pile capacity (i.e. structural pile failure). Testing was performed under displacement control. Maximum displacements of 51 cm (20 in) and 13 cm (5 in) were reached for the flagpole and fixed head specimens, respectively. Identical loading procedures were implemented in the OpenSees model.

Pile & Soil Characteristics

The cross-sectional properties of the flagpole and the fixed head shafts are identical and consisted of 0.61m (2ft) diameter reinforced concrete shafts. Laboratory testing revealed an in-situ concrete compressive strength between 31 and 43 MPa (4,500 and 6,235 psi) for the flagpole specimen cylinders and 30 and 36 MPa (4,350 and 5,220 psi) for the fixed head specimen concrete cylinders. The longitudinal reinforcement for both shafts consisted of 8 #9 bars (d

b = 29mm) A706, Grade 60 steel with

a yield stress of 469 MPa and 483 MPa (68 ksi and 70 ksi) for the flagpole and fixed head shaft, respectively. Transverse reinforcement was installed in form of 48 cm (19 in) diameter spirals made of #5 bars (db = 16 mm) spaced at 11 cm (4.3 in) pitch over the length of the pile. The clear concrete cover over the spiral was 6 cm (2.36 in). Both piles extended about 7.6 m (25 ft) below ground. The flagpole pile extended 4.06 m (13.3 ft) above ground and has a small cap to connect the hydraulic actuator to the test specimen. The fixed head pile had a concrete cap with dimensions of 2.2 m x 1.5 m x 1.8 m (7.2 ft x 4.9 ft x 5.9 ft) and longitudinal reinforcement of the pile extended approximately 1.8 m (5.9 ft) into the pile cap. The cap was restrained from rotation during the duration of the experiment. Photographs of both test specimens are shown in Figs. 3a&b. The specimens were placed in a soil profile consisting of a 5.9 m (19.4 ft) layer of partially saturated overconsolidated silty sandy clay followed by a silty, medium to fine grained sand layer that extends to a depth of approximately

9 m (29.5 ft).Detailed soil profiles and in-situ strengths can be found in Lemnitzer et al. (2010)

Analytical Model

The vertical model discretizations of the OpenSees models for both pile specimens are shown in Fig. 4a&b. The flagpole pile was modeled using a total of 35 elements, 10 elements above ground-line with an element length of 41 cm (16 in) and 25 elements below ground-line with an element size of 30 cm (12 in). Lateral loads or displacements were applied at the top of the pile, which was free to rotate. The fixed head pile had an identical number and length of the elements below ground-line as the flagpole shaft, and a single, rigid element of 91 cm above ground-line which allows the lateral load or displacement to be applied at a location consistent with the test condition. The top node was restrained against rotation, but allowed to translate laterally.

Vertical displacement at the top of the pile was unrestrained for both the flagpole and the fixed-head configurations. The cross-sections of the pile shafts are represented using a fiber model consisting of 8 fibers. Each fiber was defined through a coordinate location, a fiber area and the respective material properties accounting for confined concrete, unconfined concrete, longitudinal and transverse rein-forcement steel. The material properties were selected based on the presented in-situ speci-men properties. The confined concrete charac-teristics were determined using the Saatcioglu and Razvi (1992) relationship. The constitutive relationship employed in the analytical model

[FIG. 3a&b] Photograph of Flagpole (left) and Fixed Head (right) Field Specimens


for concrete considers also the effects of biaxial compression softening and tension stiffen-ing. The reinforcement is modeled using the Menegotto-Pinto (1973) model and accounts for the softening of the average stress-strain relationships of reinforcing bars embedded in concrete. The model that does not account for shear flexure interaction is hereafter referred to as flexure only model (or uncoupled model) and the model that incorporated the interac-tion is called shear flexure interaction model (or coupled model).

The soil in the analytical model is represented using p-y curves obtained from the experimen-tal test results. These curves were derived using an alternative approach as described in detail by Stewart et al. (2007) and do not follow the traditional procedure of double integration and differentiation of field curvature measurements. It is important to note that the implemented experimental p-y curves were derived with non-linear M-φ relationships, but did not account for possible shear flexure interaction. Hence a good fit should be obtained for the analytical flexure-only model with the measured test results. For simplification, a tri-linear fit to the p-y curves was implemented in OpenSees. Curves for six representative depths for the flagpole and fixed head specimen are shown in Fig. 5.

A detailed description of the coupled shear flexure interaction model used for this analy-sis is presented in Massone et. al (2006). The basic principle of the shear flexure interaction model relies on an interaction at the fiber level using assigned shear and normal (uniaxial)

[FIG. 4a&b&c] Vertical Flagpole (a) and Fixed Head (b) Model Discretization along with Cross-sectional Fiber Model (c) Implemented in OpenSees

c)

[FIG. 5] Tri-linear Fit to Experimentally Determined p-y Curves by Stewart et al (2007)


springs that are defined through biaxial con-stitutive material relationships. Shear, flexure and axial forces interact upon assembly at the element level, where many biaxial fibers are considered. Strains and deformations in each fiber are determined using the six degrees of freedom (u

x,u

y and θ) at both ends of the model

element by assuming that plane sections re-main plane after deformation (Bernoulli). The unknown quantity ε

x (transverse normal strain)

in each fiber was solved iteratively until stress equilibrium is achieved and all strains and deformations in the cross-section can be de-termined. Even though the model is capable of including axial pile loading, p-∆ effects are not considered in the current study and the cap self-weight is considered negligible. This ap-proach enabled the validation of the analytical through comparison with experimental data. The model results presented only monotonic loading, as cyclic material relations for shear-flexure interaction are not yet available.

MODEL RESULTSThe analysis was conducted under displacement control, that is, lateral displacement at the top of the shaft is incrementally increased until a lateral deformation of 51 cm (20 in) and 13 cm (5 in) is reached for the flagpole and fixed head cases, respectively. The overall model load displacement response was compared with the experimental test results. The model response included the results of the (i) uncoupled (flexure-only) model as implied by various beam models (e.g. Bernoulli) and (ii) the coupled (shear-flexure interaction) model. Fig. 6 shows the load-deflection relationship of the two models comparing them to the experimental data.

The model load displacement response for the flagpole pile is identical for the flexure-only and shear flexure interaction model, therefore only one curve is shown in Fig. 6. Since the pile is slender and shear deformations are small, this result is expected. The pile model matches the maximum field lateral force of about 125 kN (28,100 lb) and has an initial backbone stiffness of approximately 14.5 kN/cm (8,280 lb/in). The model results indicate that the tri-linear fits used for the p-y curves reported by Stewart et al. (2007) are adequate to capture overall load-deformation responses. For the fixed head pile, a clear contribution of shear deformations can be recognized in the diverging load displace-ment response of the uncoupled (flexure only) and coupled (shear flexure interaction) model. The ‘flexure-only’ model matches the experi-mental results with respect to initial stiffness and ultimate capacity very well and reaches the experimentally obtained lateral force of about 1214 kN (136.5 ton) at a displacement of 5 cm (1.97 in). This is expected given the approach that was used to derive the p-y relations, in which the beam model did not have the abil-ity to consider the interaction between shear and flexural deformation. The maximum lateral load for the coupled (shear flexure interaction) model is only 1110 kN (111.4 ton), or approxi-mately 10% less than the “uncoupled” model. This difference might seem small with respect to the overall response, but may influence the shape and capacity of the corresponding p-y curves significantly. No strength degradation is observed for either model curves until lateral displacements of about 12 cm (4.7 in); lateral strength degradation is observed in the test results at approximately 7.5 cm (2.95 in).

[FIG. 6] Comparison of Analytical OpenSees Model Results for the Flagpole (left) and Fixed Head Pile (right) with Test Data


Shear force profiles are shown in Fig. 7. A maxi-mum shear force of about 213 kN (47,900 lb) is reached in the flagpole pile at a depth of 2.75 m (9 ft) below ground. The fixed head pile shear forces range between 1130 kN (127 ton) at the top (maximum applied force) and -240 kN (-27 ton) at a pile depth of 3.0 m (9.8 ft). A nominal shear capacity (V

n) of approximately

950 kN (107 ton) for the flagpole and fixed head shaft was calculated for comparison using the ATC-32 (1996) recommendations. Shear forces for the flagpole configuration reached only 25% of the nominal shear strength. This observa-tion is in alignment with the matching flexure dominated load displacement response of the flagpole pile shown in Fig. 6 (i.e. shear forces/deformation are negligible). Higher shear forces and displacements observed in the fixed head pile are in alignment with the diverging curves in Fig. 6b.

Displacement Profile Response

Figs. 8 and 9 show the flexural, shear and total lateral pile deformations obtained for both the flagpole and fixed-head conditions at various lateral top displacement levels using both the flexure only and shear-flexure interaction models. In the interaction model, shear strain is assumed constant along every element, such that the shear lateral deformation profile is determined by simple integration (cumulative)

over the height. In other words, the shear deformation at each pile depth was calculated as the accumulated lateral shear displacement contributed from the lower pile depths (or elements). At each level the shear displacements were calculated by multiplying the shear strain with the respective element height. The flexural component can be estimated similarly, but in this case the integration over the height is based on rotation from integration of curvature multiplied by the relative height to the point where the lateral displacement is estimated. For simplicity, the flexural component can be determined as the total lateral deformation profile minus the shear component. Fig. 8

[FIG. 7] Shear Force Profi les for Flagpole (left) & Fixed Head Pile (right)

[FIG. 8] Pile Deformation Profi le of the Flagpole Pile

[FIG. 9] Pile Deformation Profi le of the Fixed Head Pile


shows that the overall response of the flagpole pile is flexure dominated and that shear displacements are negligible (< 0.05 cm or 0.02 in). Even at the maximum top displacement of 51 cm (20 in), contribution of shear deformations is less than 1%. Flexural bending deformations are mainly taking place in the shaft sections extending above ground line. The largest shear deformations occur just below ground line and approach zero at 150 cm (59 in) above and below ground line. For the fixed head pile, (Fig. 9) shear deformations are much larger (> 50 times) at similar top displacement levels than for the flagpole pile, and start contributing to the overall pile displacement much before the nominal shear capacity is approached. Shear deformations contribute between 30 to 40% at its maximum to the total pile lateral displacements. Most deformations occur right below ground line (up to about 60 cm (24 in) depth) and are almost zero at other locations. Flexural deformations account for 60 to 70% to the total displacements between ground line and a depth of 250 cm (98 in).

SUMMARY AND RECOMMENDATIONSA 0.61 m (2 ft) diameter reinforced concrete flagpole pile and a fixed head pile were modeled using OpenSees to investigate the effects of shear-flexure interaction on the pile response and to provide a basis for future research work that could include the modification of previously derived p-y curves. The pile configurations and material parameters were calibrated to match those used in the experimental study of large scale pile foundation systems conducted by Stewart et al., (2007). Results showed that for the flagpole pile, shear forces were less than 25% of the nominal shear strength and shear deformations were insignificant, contributing less than 1% of to the peak lateral displacement. For the fixed head specimen, results indicated that shear deformations significantly influence the overall top load displacement response for displacements exceeding 0.6 cm, even though the shear force in the pile does not reach the nominal pile shear capacity until a lateral displacement of 5.0 cm (2 in). This implies that even when nominal shear capacities per design code are met, shear deformations strongly influence the pile behavior and may result into less safe design. Shear deformations along the pile depth for the coupled model

were found to account for up to 40% of the total lateral displacement and reached their maximum contribution (analytically) where significant lateral strength degradation was observed experimentally, hence providing a good correlation between in-situ pile behavior and model findings. The coupled model underestimated the lateral top load at lateral displacements exceeding 2.5 cm (1 in) by approximately 10% because the p-y curves used in this study were calibrated to provide a good fit for a ‘flexure only’ (uncoupled) model. The results suggest that the actual p-y curves are different than those derived by Stewart et al. (2007). A recalibration study is currently in progress. Fig. 10 proposes an innovative sensor layout for future experimental studies that would enable the in-situ measurement of combined flexural and shear deformations using longitudinally and diagonally arranged sensor layouts over the contributory pile depth. Proposals for this experimental work are currently underway. Successful results will provide an experimental confirmation of the analytical studies and enable the revision of p-y curves for reinforced concrete fixed head piles. Results will significantly contribute to the improvement of lateral pile design by eliminating unconservative horizontal displacement assumptions that are made by neglecting shear deformations such as done in current p-y models.

REFERENCESAmerican Petroleum Institute (API), (1993). 1. “Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms.” API recommended practice 2A (RP 2A)

Hiraishi, Hisahiro (1984); “Evaluation of 2. shear and flexural deformations of flexural type shear walls”; Bulletin of the New Zealand National Society for Earthquake Engineering, v. 17, No 2, p. 135–144.

[FIG. 10] Proposed Combination of Diagonal and Longitudinal Sensors


Lemnitzer, A., Khalili-Tehrani, P., Ahlberg, 3. E.R., Rha, C., Taciroglu, E., Wallace, J.W., Stewart, J.P. (2010). “Nonlinear efficiency factors for bored pile group under lateral loading” Journal. Geotechnical and Geoenvi-ronmental Engineering, ASCE, 136 (12).

Lemnitzer, A., Massone, L. and Wallace, J., 4. 2010, “Influence of reinforced concrete pile shear-flexure interaction on p-y curves”. 14th European Conference on Earthquake Engineering, Skopje, Macedonia, Aug 30 – Sept 3rd, 2010

Massone, L.M. and Wallace, J.W., 2004, 5. “Load – Deformation Responses of Slender Reinforced Concrete Walls”, ACI Structural Journal, V. 101, No. 1, pp. 103-113.

Massone, L.M., Orakcal, K., Wallace, J.W. 6. (2006). “Modeling flexural/shear interaction in RC walls”, ACI-SP-236, Deformation capacity and shear strength of reinforced concrete members under cyclic loadings, American Concrete Institute, Farmington Hills, MI Paper 7, pp. 127-150, May 2006

McClelland, B. and Focht, J. (1958). “Soil 7. Modulus for Laterally Loaded Piles”, Transactions, ASCE, Vol. 123, Paper 2954, 1049-1086.

Menegotto, M. and Pinto, E., 1973, “Method 8. of Analysis for Cyclically Loaded Reinforced Concrete Plane Frames Including Changes in Geometry and Non-Elastic Behavior of Elements under Combined Normal Force and Bending”, Proceedings, IABSE Symposium, Lisbon, Portugal, pp. 15-21.

Oesterle, R. G. 9. et al. (1976); “Earthquake resistant structural walls - test of isolated walls”; PCA report to National Science Foundation no GI-43880/RA-760815.

Oesterle, R. G. 10. et al (1979); “Earthquake resistant structural walls - test of isolated walls - phase II”; PCA report to National Science Foundation no ENV77-15333.

OpenSees Development Team (Open Source 11. Project). OpenSees: Open System for Earthquake Engineering Simulation. http://opensees.berkeley.edu, 1998-2009

Saatcioglu, M. and Razvi, S.R. (1992) 12. “Strength and Ductility of Confined Concrete”. Journal. Structural. Eng., ASCE, 118 (6), 1590-1607.

Stewart, J.P., Taciroglu, E., Wallace, J.W., 13. Ahlberg, E.R., Lemnitzer, A., Rha, C., Khalili-Teherani, P., Keowen, S., Nigbor, R., Salamanca, A. (2007). “Full scale cyclic large deflection testing of foundation support systems for highway bridges. Part I: Drilled shaft foundations,” Report No. UCLA SGEL-01, University of California, Los Angeles.


Installation and Performance Characteristics of High Capacity Helical Piles in Cohesive SoilsMohammed Sakr, PhD., P.Eng, Core Manager, Geotechnical Services, WorleyParsons Canada,

Infrastructure & Environment, Edmonton, AB Canada ; Ph: 780-916-4137,

[email protected]

ABSTRACTThis paper presents the results of a full-scale axial compression and tension (uplift) testing program executed on large-capacity helical piles installed in very stiff to very hard clay till soils (i.e. cohesive soils). A total of ten tests including six axial compression tests and four tension (uplift) tests were carried out. The results of the axial compressive and tensile pile load tests and field monitoring of helical piles with either a single helix or double helix are presented in this paper. Helical piles are traditionally used to support light to medium loads (i.e. loads up to 500 kN or 56 tons). However the results of the present study confirmed that helical piles can support significantly higher axial compressive loads up to 2450 kN or 275 tons, which is about 5 times traditional design loads. The study also demonstrated the high tensile capacities of helical piles which was as high as 85% of their compressive capacity. Hence helical piles can provide a viable design option for foundations supporting relatively heavy loads.

INTRODUCTIONThe use of helical piles to support large loads for multi-million dollar projects has increased considerably in the last few years. The increased use of helical piles was empowered by advancements in hydraulic heads and the industry’s need to have a robust deep-foundation option to meet the challenging global economy. As described by several researchers (Bobbitt and Clemence, 1987; Perko 2009; Sakr 2009), there are numerous advantages for helical piles in terms of speed and ease of installation, and high axial capacities.

The axial capacity of helical piles may be estimated analytically using either the individual bearing or cylindrical shear methods. The individual bearing method, recently adapted in the Canadian Foundation Engineering Manual (Meyerhof and Adams 1968; Vesic 1971; and CFEM 2006) assumes that bearing failure occurs at each individual helix. The cylindrical shear method (Vesic 1971; Mitsch and Clemence 1985; Das 1990; Zhang 1999) assumes that a cylindrical shear failure surface, connecting the uppermost and lowermost helices, is formed and that its axial capacity is the sum of shear resistance along the cylindrical surface, adhesion along the top portion of the steel shaft above the helix level and bearing resistance above the top helix (for

uplift loading) and bearing resistance below the bottom helix (for compression loading).

The main objectives of the present study are to validate the installability of relatively large diameter helical piles and to evaluate the axial capacities of high performance helical piles installed in very stiff to very hard clay till based on the results of full-scale loading tests. The specific objectives of the test program were: (1) to define an appropriate failure criterion for helical piles, (2) to evaluate their axial compressive and tensile capacities, (3) to compare their axial compressive and tensile capacities and (4) to validate design methodology for high-performance helical piles. In order to achieve these objectives, ten full-scale load tests were carried out. The testing program included six axial compressive tests and four axial tensile (uplift) tests. Details of the testing program are described in the following sections.

SUBSURFACE CONDITIONSThe testing site is located in northern Alberta, Canada. The testing site is a large site with variable soil conditions. Seven test locations were selected to represent different soil conditions, including sandy soils (i.e. cohesionless material), and glacial till and or clay shale (i.e. both cohesive materials). A total of twenty one axial compression and uplift load


tests were performed, including eleven pile load tests in cohesionless soils performed at four different locations (Sites 1 to 4). The remaining ten load tests were carried out at three different locations, at Sites 5 to 7, where subsurface soils are mainly cohesive soils. The results of helical pile load tests on cohesionless soils are reported elsewhere (Sakr 2011). Subsurface soil stratigraphy and groundwater conditions for cohesive soils at test sites 5 to 7 are described in detail in the following sections.

Table 1 summarizes subsurface stratigraphy and soil properties at test sites 5 to 7.

[TABLE 1] Summary of Soil Properties

Depthm

Soil description

SPT blow count per 300 mm,

Total unit

weight, kN/m3

Undrained Shear

Strength, kPa

Frictional resistance angle, Ø (o)

Site 5

0 - 2Clay Till, very

stiff17 18 100 0

2 - 4 Clay Till, hard 32 20 200 0

4 – 6.5Clay Till, very

hard64 20 400 0

> 6.5Sand, dense to

very dense42-60 20 0 40

Site 6

0 – 1.7 Sand, compact NA 18.5 - 31

1.7 – 9.9Glacial Till,

stiffNA 18 85 0

9.9 – 13.9Glacial Till, very stiff

NA 18 115 0

13.9 - 15.0Sand, dense to

very denseNA 19.5 0 37

Site 7

0 – 1.9 Sand, compact NA 18.5 - 33

1.9 – 6.9Glacial Till,

stiffNA 18 80 0

6.9 – 13.7Glacial Till, very stiff

NA 18 115 0

13.7 – 16.4Glacial Till, very stiff to

hardNA 19 180 0

Site 5Soil stratigraphy at test Site 5 consists of glacial till that extended to a depth of about 6.4 m (21 ft), underlain by very dense sand that extended to the end of the test hole at a depth of about 30.6 m (100 ft). Clay till was sandy with some silts and contained a trace of fine to coarse subrounded gravel up to 50 mm (2 in) in size. A fine to coarse subrounded gravel lens was encountered at a depth of about 2.2 m (7.2 ft) and extended to about 2.5 m (8.2 ft). A seam of black woody debris was also encountered at a depth of about 2.5 m (8.2 ft). Standard Penetration Testing (SPT) blow counts varied between 17 and 64 blows per 300 mm (per ft) of penetration, indicating very stiff to very hard consistency. It should be noted that a very hard soil layer was encountered during drilling at the interface between clay till and sand layers at depths of about 6.0 m to 6.4 m (19.7 ft to 21.0 ft). The undrained shear strength for each soil sublayer, C

u, measured in kPa, was estimated

assuming Cu = 6.25 N, where N is the SPT blow

count per 300 mm (per ft) of penetration. Groundwater level at the test hole locations was relatively shallow and was measured upon completion of test holes at about 5.2 m (17.0 ft) below existing ground surface.

Site 6Subsurface conditions at test Site 6, based on Cone Penetration Testing (CPT) data, consists of surficial sandy silt, to a depth of about 1.7 m (5.6 ft), over glacial till layers, to a depth of about 13.9 m (45 ft), over dense, gravelly sand. CPT hit refusal at a depth of about 15 m (49 ft) below existing ground. The average cone tip resistance and sleeve friction for different layers are presented in Table 1 below. The estimated undrained shear strength, C

u,

values for stiff and very stiff glacial till were 85 and 115 kPa (12.3 and 16.7 psi), respectively. Groundwater level, based on pore pressure measurements, was estimated to be at a level of about 14.5 m (48 ft) below existing ground.

Site 7Subsurface conditions at test Site 7 consisted of surfacial sandy silt, to depth of about 1.9 m (6.2 ft) below existing ground surface, over a stiff to very stiff silty clay layer, to depth of about 13.7 m (45.0 ft), underlain by a hard silty clay layer that extended to the end of log at 16.4 m (53.8 ft). The estimated undrained shear strength, C

u, for the stiff and very stiff silt clay


were 80 kPa and 115 kPa (11.6 psi and 16.7 psi), respectively. The estimated undrained shear strength of the lower hard silty clay layer was 180 kPa (26.1 psi). Groundwater level at test Site 7 was estimated at a depth of about 10.8 m (35.4 ft) below existing ground surface.

TEST PILE CONFIGURATIONSThe configurations for different piles considered for the helical pile load test program are summarized in Table 2. Fig. 1 provides typical helical pile configurations. Helical pile types identified by even numbers were for piles with double helices (i.e. type 4 and type 8) while piles identified with odd numbers were for piles with a single helix. All piles were round shaft type with shaft diameters varying between 324 mm and 508 mm (12.8 in and 20.0 in) and helix diameters varying between 762 mm and 1016 mm (30 in and 40 in). Helices for double helix piles were spaced either at two or three times their helix diameter. Pile shafts were ASTM A252 Grade 3 seamless steel pipes with yield strength of 310 MPa (45 ksi). Helices were formed from CSA G40.21 44W hot-rolled structural steel with yield strength of 300 MPa (43.5 ksi). Helices were welded to a pile shaft using 9.5-mm (3/8-in) fillet weld. The leading edge of each helix was sharpened to reduce installation torque requirements.

[TABLE 2] Summary of Pile Confi gurations

PileType

Shaft Helices

Diametermm

Thicknessmm

Diametermm

Thicknessmm

No of Helices

Spacing,S (m)

3 324 9.5 762 25.4 1 NA

4 324 9.5 762 25.4 2 2.29

5 406 9.5 914 25.4 1 NA

7 508 9.5 1016 25.4 1 NA

8 406 12.7 813 25.4 2 1.63

PILE INSTALLATIONHelical piles are typically installed through the use of mechanical torque applied at the pile head. Table 3 provides a summary of the pile installations at test Sites 5 to 7, including the torque recorded at the end of the pile installation, predrill depth, thickness of the soil plug, and depth of embedment. Measured torque values were recorded at different depths for different piles tested in this study, and the results are presented in Figs. 2a through 2c for piles installed at Sites 5 to 7. Predrilling for test piles was performed only at test Site 5 to reduce the torque requirements and to facilitate pile installation. The typical predrilling process included drilling a pilot hole at the location of the test pile with a diameter approximately 51 mm (2 in) less than the diameter of the pile shaft. For example, for a pile shaft diameter of 406 mm (16 in), an auger with a diameter of about 355 mm (14 in) was used. Test piles ST5, ST6, ST7, ST13, ST14, and ST15 were predrilled to depths varying between 3.9 m and 5.2 m (13 ft and 17 ft) below existing ground surface.

[FIG. 1] Typical Test Pile Confi gurations: (a) Single Helix Pile; and (b) Double Helix Pile


Upon completing the predrilling process, depths of open holes were measured, and all holes were open to their full drilled depth. Fig. 3 shows a typical drilled hole for the helical pile installation at test Site 5. It worth mentioning that the measured torque values at Site 5 (Fig. 2a), especially within predrilled depths, should be used with caution due to the presence of predrilled holes. However, several general observations can be noted from Fig. 2a: 1) As expected, the measured torque values within the predrill depth (i.e. depths of 3.9 to 5.2 m or 13 ft to 17 ft) were scattered. 2) A clear trend of a sharp increase in torque values at a depth of about 4 m (13 ft) was observed for most of the installations at Site 5. The sharp increase in torque value may indicate the presence of harder soil material at that depth. 3) Measured torque at the end of the installation for piles ST5, ST6, and ST7 (Types 3 and 4) with a shaft diameter of 324 mm (12.8 in) was 211 kN.m (155.6 kip-ft), and the corresponding embedment depths varied between about 5.7 m and 5.9 m (18.7 ft and 19.4 ft). Piles ST13 and ST14 (Type 5), with a shaft diameter of 406 mm (16 in), were successfully installed using a 338 kN.m (250 kip-ft) drive head to a maximum embedment depths of 5.8 m and 5.6 m (19.0 and 18.4 ft), respectively. The measured torque value at the end of installation of pile ST15 (Type 7), with a shaft diameter of 508

[TABLE 3] Summary of Pile Installation

Test ID Pile TypeShaft

Diameter

mm

Installation Torque at end of

installation

kN.m

Embedment Depth

m

Soil Plug Thickness

m

Predrill Depth

m

Site 5

ST5 4 324 211.5 5.9 1.7 5.2

ST6 4 324 211.5 5.7 1.8 4.8

ST7 3 324 211.5 5.7 2.0 5.0

ST13 5 406 338.3 5.8 - 4.1

ST14 5 406 338.3 5.6 - 3.9

ST15 7 508 338.3 5.4 2.8 4.1

Site 6ST61 8 406 297.7 14.1 2.9 NA

ST62 8 406 297.7 14.3 2.9 NA

Site 7ST71 8 406 338.3 18.5 - NA

ST72 8 406 338.3 18.5 - NA

(a)

(b)


mm (20 in) , was about 338 kN.m (250 kip-ft) and the corresponding embedment depth was 5.4 m (17.7 ft).

Piles at Sites 6 and 7 were advanced into subsurface layers by applying torque at the pile head using the rotary driving head without drilling a pilot hole. Torque values were measured versus depth as well (Figs. 2b and 2c). Therefore, torque measurements may be used to qualitatively assess soil conditions at depth. It can be seen from Fig. 2b that the measured torque during installation for piles ST61 and ST62 at test Site 6 increased with depth until the end of the pile installation. Torque values increased at a higher rate up to depth of about 4 m (13 ft) at which point they increased at a lesser rate up to depths of about 8 m (26 ft). A considerable increase of torque values was observed at depth of about 9 m (30 ft), and another sharp increase in torque values

was observed at a depth of about 12 m (39 ft), followed by a slight reduction. The measured torque at the end of installation for both piles was 298 kN.m (219.8 kip ft), and the corresponding embedment depths were 14.1 m and 14.3 m (46.3 ft and 46.9 ft), respectively.

The sleeve friction (or side friction) values measured during the CPT test near the location of test Site 6 are also plotted in Fig. 2b. Comparing side friction and torque data indicates that both sets of data showed similar trends, except that there was a lag of data of about 1.5 m (5 ft) between CPT side friction and torque data. This lag can be explained as the

length of sleeve of the CPT penetrometer is about 0.15 m (0.5 ft) while test piles ST61 and ST62 had double helices spaced at about 1.6 m (5.2 ft). Therefore, the lag between both sets of data is basically the size difference between the CPT sleeve and the portion of helical pile between the top and lower helices. This observation suggests that a correlation between side friction and torque measurement values may exist. This can be further explained by imagining the helical pile as a giant cone penetrometer, where similarities may exist. Therefore, in the absence of geotechnical information, installation torque records may provide a way of qualitatively assessing consistency of different soil layers.

It can be seen from Fig. 2c, for piles ST71 and ST72 at test site 7, that the measured torque during installation increased with depth up to about 8 m (26 ft), then decreased up to a depth of about 8.5 m (28 ft), and finally gradually increased up to a depth of about 13.5 m (44 ft). A considerable increase in torque values was observed at a depth of about 17 m (56 ft). The measured torque at the end of the installation for piles ST71 and ST72 was 338 kN.m (249.3 kip-ft) and the corresponding embedment depths were 18.5 m (60.7 ft) for both piles. As with Site 6, the torque-depth plot showed similar trend to the side friction-depth plot.

TEST SETUPThe axial compression and tensile load tests were carried out in accordance with ASTM standards D 1143-07 and D 3689-07. Since the main objective of the load tests was to determine the ultimate bearing capacity of the

(c)[FIG. 2] Measured Installation Torque vs. Depth for: (a) Site 5; (b) Site 6; and (c) Site 7

[FIG. 3] Typical Drilled Pilot Hole prior to Pile Installation at Site 5


pile, Procedure A (Quick Test) was adopted for eight of the tests wherein numerous small load increments were applied and maintained constant over short time intervals. Procedure B (Maintained Test), or the slow test, was used for two compression tests at Site 5 (piles ST7 and ST15) to provide a basis for comparison between both test methods and to study the creep effect on the axial load test results. Six axial compression pile load tests were carried out at test Sites 5 to 7. A total of four axial tension (uplift) tests were performed at test Sites 5 to 7 using Procedure A (Quick Test). Fig. 4 shows a typical load test setup using six reaction piles.

[FIG. 4] Oblique View for Axial Compression Load Test Setup using Six Reaction Piles

TEST PROCEDURES FOR AXIAL LOAD TESTS (COMPRESSION AND UPLIFT)The following specific test protocols using Procedure A for Quick Tests for piles under axial compressive or uplift loads were applied:

Apply test loads in increments equal to 5% 1. of the anticipated failure loads and maintain load constant for 5 minutes. Monitor movements using Linear Displacement Transducers (LDTs) at intervals of 30 seconds. Monitor movements using mechanical dial gauges at the end of each load increment.

Add load increments until reaching a failure 2. load, but do not exceed the safe structural capacity of the pile or reaction apparatus. Upon reaching the failure load, maintain the load for a longer period of time to monitor creep behaviour (about 10 to 15 minutes).

Unload the test pile in five increments and 3. hold for 5 minutes with same monitoring intervals used for loading. Upon reaching zero load continue monitoring the LDT

readings for 10 minutes to assess the rebound behaviour.

The following specific test protocols using Procedure B for Maintained Tests were applied:

Apply test loads in increments equal to 1. 25% of the anticipated design loads to a maximum maintained load of 200%. Maintain each load increment until the rate of axial movement does not exceed 0.25 mm (0.01 in) per hour. Monitor movements using LDTs at intervals of 30 seconds. Monitor movements using mechanical dial gauges at the beginning and at the end of each load increment.

After applying the maximum load and 2. reaching an overall test duration of at least 12 hours, begin unloading when the axial movement measured over a period of one hour does not exceed 0.25 mm (0.01 in) or a maximum increment of two hours, whichever occur first.

If failure occurs during loading, maintain 3. the failure load until the total axial movements equal 15% of the helix diameter. If failure does not occur, hold the full load for two hours.

Unload the test pile in four increments and 4. hold for 20 minutes between increments with the same monitoring intervals used for loading.

TEST RESULTS Axial Compressive Load Test Results

The load-displacement curves for compressive load tests at test Site 5 and Sites 6 and 7 are presented in Figs. 5 and 6 and were used to determine the load capacity of the piles. It should be noted that load-displacement curves were established using the loads and corresponding displacements at the end of each load increment to account for the creep effect. This interpretation technique provides a conservative estimate of the axial capacities and reduces the creep effect on the axial capacities. It can be seen from Figs. 5 and 6 that the load displacement curves are also characterized by three parts (a primary linear part up to a displacement of about 2 mm (0.08 in) , followed by a non-linear component up to a displacement of about 18 mm to 43 mm (0.7 in to 1.7 in), and a secondary linear component with less slope). No plunging failure was observed for all piles tested at Sites 5 to 7, and all piles continued to resist higher loads up to the end of testing.


(a)

(b)

[FIG. 5] Applied Loads at Pile Head vs. Displacement for Axial Compression Pile Load Tests at Site 5 for: (a) Piles ST6 and ST7, and (b), ST7, ST13 and ST15

The loads at the pile head at the end of the initial linear component at test Site 5 (Fig. 5) varied between about 250 kN (28 ton) for pile ST6 and about 500 kN (56 ton) for pile ST13. The loads at the pile head at the beginning of the second linear component varied between 1180 kN and 1600 kN (133 ton and 180 ton). A comparison between load displacement curves for piles ST6 with double helices, 762 mm (30 in) in diameter, and pile ST7 with a single helix, 762 mm (30 in) in diameter, and a similar embedment depth, is presented in Fig. 5a. It can be seen from Fig. 5a that both piles ST6 and ST7 showed similar results at the early stages of loading, up to a load of about 250 kN (28 ton), corresponding to a displacement of 1 mm (0.04 in). However, at higher displacement levels pile ST6, with double helices, exhibited resistance up to about 21% higher than pile ST7, with a single helix. Comparing piles ST7, ST13, and ST15 all with a single helix and similar embedment depths (Fig. 5b), indicates that the piles showed a similar trend. However, the load at the end of the initial linear portion of the load displacement curves increased from 250 kN (28 ton) for pile ST7,

with a shaft diameter of 324 mm (12.8 in) , to about 400 kN (45 ton) for pile ST15, with a shaft diameter of 508 mm (20 in).

[FIG. 6] Applied Load at Pile Head vs. Displacement for Axial Compression Pile Load Tests at Sites 6 and 7 for Pile ST61 and ST71

ST61 and ST71 at test Sites 6 and 7 (Fig. 6); both have a shaft diameter of 406 mm (16 in) and double helices 813 mm (32 in) in diameter. ST61 and ST71 have embedment depths of 14.1 (46.3 ft) and 18.5 m (60.7 ft), respectively. It can be seen from Fig. 6 that the load displacement curves for both ST61 and ST71 were very similar up to a displacement level of about 5 mm (0.2 in) , which corresponds to a load of about 800 kN (90 ton). However, at higher displacement levels, pile ST71 showed a stiffer response and continued to resist higher loads than did ST61 at the same displacement level until the end of test. The loads at the pile head at the end of the initial linear component for piles ST61 and ST71 were about 480 kN (54 ton) and 400 kN (45 ton), respectively. The loads at the pile head at the beginning of the second linear component for piles ST61 and ST71 were 1670 kN (188 ton) and 2290 kN (257 ton), respectively, and the corresponding displacements were 34 mm (1.3 in) and 43 mm (1.7 in).

Axial Compressive CapacityThere are a large number of failure criteria used to interpret the axial compressive capacities of piles from pile load test results; some of these include the Davisson criterion, Brinch Hansen, L1-L2 method, FHWA (5%) and ISSMFE (10%). Davisson’s criterion (1972) is the most widely used method for estimating the axial capacities of piles. In Davisson’s criterion the ultimate capacity is defined as the load corresponding to a total displacement equal to the sum of elastic

deflection of the pile and the offset as

identified in Equation [1] below:


[1]

where S = the displacement in m; P = load on pile; L = pile length in m;A = cross-sectional area; E = Yong’s modulus of pile material; and d = pile diameter in m.

It is worth mentioning that the Davisson cri-terion was initially developed for driven steel piles with small diameters up to 305 mm (12 in). The main shortcoming of applying Davis-son’s criterion for helical piles is that the offset limit was initially developed to satisfy the move-ments necessary to mobilize the toe resistance of driven steel piles with a small toe diameter. Davisson (1993) suggested that, for drilled shaft piles, the term that contains diameter, D, be multiplied by a factor of 2 to 6. Nesmith and Siegel (2009) and Kulhawy and Hirany (2009) argued against the use of the Davisson crite-rion with large-diameter, cast-in-place concrete piles. However helical piles derive most of their resistance from the helices and the end-bearing component and, therefore, the use of Davisson criterion is likely to yield lower capacities that do not reflect the actual capacities of helical piles.

Kulhawy and Hirany (1989) developed the L1-L2 failure criterion for load displacement curves that exhibits three regions similar to the load test results of all piles considered in this study (i.e. linear, transient and final linear compo-nents). Point L1 corresponds to the load at the end of the first linear component that repre-sents the frictional resistance of the pile while point L2 corresponds to the load at the begin-

ning of the second linear component, beyond which a small increase in the load produces a significant increase in displacement.

Sakr (2009) defined the ultimate capacity of a helical pile as the load that corresponds to a displacement of 5% of the helix diameter. The end-bearing capacity is typically fully mobilized at relatively large displacement levels (about 10% of lower helix diameter). However the 10% displacement of a large-diameter helical pile is relatively large. For example for a helical pile with a helix diameter of 1016 mm (40 in), the displacements that produce the ultimate capac-ity using 10% and 5% criteria are about 102 mm (4 in) and 51 mm (2 in), respectively. Therefore, the 5% failure criterion provides more practical displacement levels since in many cases, design is controlled by the allowable displacements.

At test Site 5, the ultimate capacity of helical piles were estimated using the L1-L2 method, 5% criterion, and Davisson failure criterion and the results are presented in Table 4. It can be seen from Table 4, that the ultimate capacities of piles ST6, ST7, ST13, and ST15 using the 5% displacement criterion were about 1912 kN, 1540 kN, 2292 kN, and 2400 kN (215 ton, 173 ton, 258 ton and 270 ton), respectively, while the ultimate pile load capacities based on L1-L2 criterion were 1510 kN, 1250 kN, 1500 kN, and 1600 kN (170 ton, 140 ton, 169 ton and 180 ton), respectively. The axial capacities of piles estimated using the L1-L2 method were about 15% to 28% lower than that values estimated using the 5% criterion. The estimated capacities using the Davison failure criterion were about 45% to 32% lower than the 5% criterion and the corresponding displacements were about 13 mm to 15 mm (0.5 in to 0.6 in) which is about 27% to 34% of the displacements using the 5%

[TABLE 4] Summary of Axial Compressive Load Test Results

Site ID Test IDPile

Type

Shaft Dia.

mm

Helix Dia.

mm

No of helices

Ultimate Capacity (Davisson)

Ultimate Capacity L1-L2

Ultimate Capacity 5%

Load (kN)

Dispt. (mm)

Load (kN)

Dispt. (mm)

Load (kN)

Dispt. (mm)

Site 5

ST6 4 324 762 2 1050 13 1510 25 1912 38

ST7 3 324 762 1 880 13 1250 24 1540 38

ST13 5 406 914 1 1280 14.5 1500 21 2292 46

ST15 7 508 1016 1 1200 14 1600 26 2400 51

Site 6 ST61 8 406 813 2 1231 16 1587 35 1630 41

Site 7 ST71 8 406 813 2 1902 22 2291 34 2450 41


failure criterion. Comparing between the ulti-mate capacity of ST6 with double, 762 mm he-lices and pile ST7 with a single 762 mm (30 in) helix, indicates that ST6 capacities were about 25% higher than ST7. The axial compressive resistance of pile ST13 with a shaft diameter of 406 mm (16 in) and a single helix 914 mm (36 in) in diameter was higher than that of pile ST15 with a shaft diameter of 508 mm (20 in) and a single helix, 1016 mm (40 in) in diameter at low displacement levels (i.e. up to about 15 mm (0.6 in) of displacement). This odd behaviour is likely due to the differences of the testing pro-tocols used for both tests where procedure A (quick test) was used for pile ST13 while proce-dure B (maintained load test) was used for pile ST15. Pile ST15 crept considerably compared to pile ST13. Further discussion about creep effect is warranted in next section.

At test Sites 6 and 7, the axial capacities of piles ST61 and ST71 (Type 8) using the Davison failure criterion were about 1231 kN, and 1902 kN (138.4 ton and 213,8 ton), respectively and the corresponding displacements were 16 and 22 mm (0.63 in and 0.87 in). The axial capacities of piles ST61 and ST71 using the L1-L2 criterion were 1587 kN (178.4 ton), and 2291 kN (257.5 ton), respectively and the corresponding dis-placements were 35 mm (1.38 in) and 34 mm 1.34 in). The axial capacities of piles ST61 and ST71 using the 5% criterion were 1630 kN (183.2 ton), and 2450 kN (275.4 ton), respectively. The axial capacities of piles estimated using the L1-L2 method were about 3% to 7% lower than that values estimated using the 5% criterion. The estimated capacities using the Davison criterion were about 23% to 25% lower than the 5% crite-rion and the corresponding displacements were about 16 mm (0.63 in), and 22 mm (0.87 in), which is about 39% to 54% of the displacements using the 5% failure criterion. It is worth men-tioning that due to the longer piles at test Sites 6 and 7, their shaft resistances were consider-ably higher than that of piles tested at Site 5. The higher shaft resistance resulted in both the Davison and the L1-L2 failure criteria provid-ing closer correspondence to the pile capacities obtained using the 5% criterion.

Creep Effects for Cohesive Soils

In order to evaluate the creep effect on the pile load test results, incremental displacement and displacement rate versus time at load incre-ments of 1000 kN (112,4 ton) and 2000 kN

(224.8 ton) were plotted for pile ST7 and the results are presented in Figs. 7a and 7b. It can be seen from Fig. 7a that at a load level of 1000 kN, about 78% of incremental displacement was obtained at about 5 minutes of elapsed time. After which, the creep rate started to increase at a declining rate until the end of monitoring period (40 minutes). The final differential dis-placement at the end of the load increment of 1000 kN (112.4 ton) for pile ST7 was about 6.2 mm (0.24 in).

(a)

(b)

[FIG. 7] Evaluating Creep Effect for pile ST7 at Load Levels of: (a) 1000 kN and (b) 2000 kN

At high load level of 2000 kN (224.8 ton) (Fig. 7b), creep behaviour was more pronounced and differential displacement versus time curve showed three distinct regions; first linear component that extended to about 7 minutes of elapsed time followed by nonlinear component that continued until an elapsed time of about 170 minutes followed by steady creep to the end of loading increment at about 325 minutes.


The differential displacement at the end of the first linear component was about 7.8 mm (0.3 in) which is about 40% of the final differential displacement. The differential displacement at the beginning of the second linear component was about 17.6 mm (0.7 in) which is about 90% of the final differential displacement. The sec-ondary creep rate was 0.8 mm/hour (0.03 mm/hour). The differential displacement at the end of load increment of 2000 kN (224.8 ton) was about 19.6 mm (0.77 in).

It can be seen from Fig. 7 that the creep effect in cohesive soils similar to that at Site 5 is con-siderable especially at high load levels (Fig. 7b) and therefore it is expected that performing a pile load test using procedure B for main-tained load test result in softer response or higher displacements at same load level. In order to assess the creep effects on the load test results, the load-displacement curves of pile ST7 are plotted in Fig. 8 considering the following cases:

Displacement measured after 25 minutes 1. (i.e. equivalent to ASTM standard D 1143-07 Procedure A (quick test) where the with load increment about 5% of the ulti-mate load maintained for 10 minutes);

Displacement measured after 50 minutes 2. (i.e. equivalent to quick test with load incre-ment about 5% of the ultimate load main-tained for 20 minutes); and

Displacement measured at the end of each 3. load increment during the maintained test (procedure B).

It can be seen from Fig. 8 that up to a load of about 1000 kN (112.4 ton) and the correspond-ing displacement of about 16 mm (0.63 in), the creep was minor and did not affect the axial compressive capacity considerably. However at higher loads, the creep was considerable and the axial compressive resistance of the pile was lower for the case of the maintained pile load test than that for the case of quick test. For example, the ultimate capacities of pile ST7 using the 5% failure criterion and considering time intervals of 25 minutes (equivalent quick test), 50 minutes and slow test were about 1700 kN, 1620 kN, and 1540 kN (191.1 ton, 182.1 ton and 173.1 ton), respectively. Therefore, the test procedures have a considerable effect on the results for piles tested in similar soil con-ditions. As mentioned earlier this effect was clear when the results of piles ST13 and ST15

performed using quick and maintained load test procedures were compared. In order to reduce the creep effect on the test results for piles founded in cohesive soils and to maintain a practical time duration for the static load test, it is suggested to increase the time increments in the quick test (using ASTM standard D 1143-07 Procedure A) to about twenty minutes. By increasing the time interval to twenty minutes the load test may be performed in reasonable time and provide a more accurate prediction of pile capacity. If that is not achievable due to tight schedules, at least the time increments at higher load should be increased.

[FIG. 8] Assessment of Creep Effects on Load Displacement Curve for Pile ST7 at Site 5

Axial Tensile (Uplift) Load Test Results

The load-displacement curves for piles ST5 and ST14 tested at Site 5 and for piles ST62 and ST72 at Sites 6 and 7 are presented in Figs. 9 and 10 and used to determine the axial tensile load capacities for piles. It can be seen from Figs. 9 and 10 that the load displacement curves consisted of typical three components including first linear part up to a displacement of about 2 mm, followed by a non-linear component up to displacements of about 18 mm to 20 mm (0.7 in to 08 in), and a secondary linear component with flatter slope.

[FIG. 9] Applied Loads at Pile Head vs. Displacement for Axial Tension (Uplift) Pile Load Tests at Site 5 for Piles ST5 and ST14


[FIG. 10] Applied Loads at Pile Head vs. Displacement for Axial Tension (Uplift) Pile Load Tests at Sites 6 and 7

The loads at the pile head at the end of the first linear component for piles ST5 and ST14 at Site 5 were about 400 kN (45 ton) for both piles. The secondary slope of the load displacement curve for pile ST5 with double helices was steeper than that of pile ST14 with a single helix despite its larger shaft and helix size. This observa-tion confirms the favourable effects of using a double-helix pile opposed to a single helix pile in term of reducing the displacements at high load levels or in other words increasing their axial tensile capacities.

The loads at the pile head at the end of the first linear component for piles ST62 and ST72 at Sites 6 and 7 (Fig. 10) were about 520 kN and 800 kN (58.5 ton and 90.0 ton), respectively, and the corresponding displacements were about 2 mm (0.08 in) and 5 mm (0.2 in), which corresponds to the elastic deformation of the pile shaft. Both piles ST62 and ST72 reached plunging failure (true failure where pile contin-ue to move under sustained load) at displace-ments of about 62 and 69 mm (2.44 and 2.72 in), respectively, and the corresponding loads of about 1505 and 2220 kN (169 and 250 ton).

Axial Tensile Capacity

The ultimate tensile capacities of helical piles tested at Sites 5, 6 and 7 were also estimated using the L1-L2 method, 5% and Davisson failure criteria and the results are presented in Table 5. It can be seen from Table 5 that the ultimate capacities of piles ST5 (Type 4) and ST14 (Type 5) using the 5% failure criterion were about 1195 kN (134.3 ton) and 1680 kN (188.8 ton), respectively. The ultimate capacities using the L1-L2 limits for piles ST5 and ST14 were 800 kN (90 ton) and 1260 kN (141.6 ton), respectively. The axial uplift capacities of piles ST5 and ST14 using the Davisson failure cri-terion were 630 kN (71 ton) and 1056 kN (120 ton), respectively. As expected, the axial uplift capacities of both piles (ST5 and ST14) using the Davison criterion were about 60% of their capacities using the 5% failure criterion and the corresponding displacements were about 12 mm and 14 mm (0.47 in and 0.55 in). The mea-sured displacements were about 32% and 30% of the displacement using 5% criterion. The ulti-mate capacities estimated using the L1-L2 limits were about 25% to 33% lower than that values estimated using the 5% criterion.

At test Sites 6 and 7, the ultimate tensile capaci-ties (using 5% criterion) of piles ST62 and ST72 (Type 8) with double helices were 1420 kN (160 ton) and 2100 kN (236 ton), respectively. The ultimate tensile capacities of piles ST62 and ST72 using the L1-L2 limits were about 1280 kN (144 ton) and 1840 kN (207 ton) which is about 10% to 12% lower than that capacities measured using the 5% criterion. The axial uplift capaci-ties of piles ST62 and ST72 using the Davisson criterion were 1140 kN (128 ton), and 1730 kN (194 ton), respectively, which is about 18% to 20% lower than 5% criterion.

[TABLE 5] Summary of Axial Tension (Uplift) Test Results

Site ID Test IDPile

Type

Shaft Dia.

mm

Helix Dia.

mm

No of helices

Ultimate Capacity (Davisson)

Ultimate Capacity L1-L2

Ultimate Capacity 5%

Load (kN)

Dispt. (mm)

Load (kN)

Dispt. (mm)

Load (kN)

Dispt. (mm)

Site 5ST5 4 324 762 2 630 12 800 18.7 1195 38

ST14 5 406 914 1 1056 14 1260 20 1680 46

Site 6 ST62 8 406 813 2 1140 15 1280 25 1420 41

Site 7 ST72 8 406 813 2 1730 21 1840 25 2100 41


COMPARISON BETWEEN MEASURED AND ESTIMATED PILE CAPACITIES

Axial Compressive Load Test Results

For the case of compressive loading for heli-cal piles tested at Sites 5 and 7 and founded in clay/clay till soils, the total axial compressive resistance can be estimated as follows:

Qc= ∑ A

H C

u N

c + π d H

effα C

u [2a]

whereN

c = dimensionless bearing capacity factors;

(Nc = 9 for helices smaller than 0.5m; = 7

for helices between 0.5 m and 1m; and = 6 for helixes larger than 1 m in diameter) (ref. Canadian Foundation Engineering Manual, 4th Edition, pp 266)A

H = net surface area of bearing helix (helix

area – shaft area), m2;C

u = undrained shear strength of soil layer,

kPa;d = diameter of the shaft, (m); H

eff = effective length of pile, H

eff = H – D, (m);

H = depth to top helix, (m); and α = Adhesion factor

For the case of compressive loading for helical piles with double helices tested at Site 6 and founded in sand soils, the total axial compres-sive resistance can be estimated as follows:

Qh= A (γH

bN

q+0.5γDNγ)+AH CuNc+πd H

effα C

u [2b]

whereA = projected helix areaγ = Unit weight of the soil;H

b = depth to helical bearing plate

D = diameter of helical plate

[TABLE 6] Comparison between Measured and Estimated Axial Compressive Capacities

Site ID Test ID Pile Type

Ultimate Capacity

Estimated

Ultimate Capacity (5%)

Measured Prediction Ratio

Shaft

kN

Helix(es)

kN

Total

kN

Shaft

kN

Helix(es)

kN

Total

kN

Site 5

ST6 4 287 1799 2086 360 1352 1912 1.09

ST7 3 287 1276 1563 250 1290 1540 1.01

ST13 5 341 1746 2087 500 1792 2292 0.91

ST15 7 339 1946 2285 400 1800 2400 0.95

Site 6 ST61 8 731 1070 1801 490 1140 1630 1.1

Site 7 ST71 8 1055 1125 2180 800 1650 2450 0.89

Nq and Nγ = bearing capacity factors for local

shear conditions.

Adhesion factor (α) can be evaluated from the following equation (CFEM, 2006):

Paα = 0.21 + 0.26 ( ) ≤ 1 [3] Cu

where Pa is the atmospheric pressure (P

a = 101

kPa) and undrained shear strength Cu in kPa.

The estimated axial compressive capacities for different piles tested at Sites 5, 6 and 7 are pre-sented in Table 6. The axial compressive capaci-ties were estimated using Eqn. 2 along with soil parameters presented in Table 1. The estimated pile capacities were compared to the measured capacities based on 5% criterion. Measured and estimated axial compressive capacities agreed reasonably within ±11%.

Axial Tensile Load Test Results

For predicting the ultimate uplift capacities of helical piles in cohesive soils encountered at test Sites 5, 6, and 7, the following expression may be used (Das, B.M. and Seeley, 1975):

Qt= ∑ AH (C

u N

u + γ’ H ) + π d H

eff α C

u [4]

HnNU = 1.2 ( ) ≤ 9 [5] D

where:Q

t = ultimate screw pile uplift capacity, (kN);

γ’ = Average effective unit weight of soil above the top helix, (kN/m3);N

u = Dimensionless uplift bearing capacity

factor for cohesive soils; andH

n = depth to helical bearing plate number n


The estimated axial uplift capacities of piles ST5, ST14, ST62, and ST72 tested in tension at Sites 5, 6 and 7 are presented in Table 7. The axial capacities were estimated using Eqn. 4 along with soil parameters presented in Table 1. The estimated capacities were also compared to the measured capacities using the 5% criterion. The estimated capacities agreed reasonably with the measured capacities with the exception of pile ST5 with double helices where its estimated capacity was about 47% higher than the mea-sured capacity. The measured low capacity of pile ST5 is likely due to the predrilling process that extended to a depth of about 5.2 m (17 ft) deeper than the upper helix level) and there-fore the upper helix was embedded into the disturbed soil zone which would considerably reduce its uplift capacity. The measured and estimated capacities for pile ST14 agreed well.

Comparison between Compressive and Uplift Load Test Results

The results of both compressive and uplift load tests for piles ST61 and ST62 tested at test Site 6 and piles ST71 and ST72 tested at Site 7 are compared in Figs. 11a and 11b. All piles were type 8 with similar configurations. Piles ST61 and ST62 were embedded to depths of 14.1 m and 14.3 m (46.3 ft and 46.9 ft), respectively; while both piles ST71 and ST72 were embed-ded to depth of 18.5 m (60.7 ft). It can be seen from Figs. 11a and 11b that load-displacement relationship was almost identical for each pile-pairs (i.e. piles ST61 and ST62 and ST71 and ST72) s at the early stages of loadings up to displacements of about 9 mm and 11 mm (0.35 in to 0.43 in). The first linear part of the load displacement curve generally is believed to rep-

resent the pile shaft resistance due to frictional resistance. Therefore, the frictional resistances of piles tested in compression and uplift (ten-sion) loading were similar. This observation supports that the same adhesion factor (α) in compression and tension as per equation 3 may be used.

(a)

(b)

[FIG. 11] Comparison between Axial Compressive and Tensile Load Tests for: (a) Piles ST61 and ST62 at Site 6 and (b) Piles ST71 and ST72 at Site 7

At higher displacement levels, piles tested in compression offered higher resistance com-pared to piles tested in tension. Comparing

[TABLE 7] Comparison between Measured and Estimated Axial Tensile (Uplift) Capacities

Site ID Test ID Pile Type

Ultimate Capacity

Estimated

Ultimate Capacity (5%)

Measured Prediction Ratio

Shaft

kN

Helix(es)

kN

Total

kN

Shaft

kN

Helix(es)

kN

Total

kN

Site 5

ST5 4 216 1546 1762 200 995 1195 1.47

ST14 5 256 1198 1454 400 1280 1680 0.87

ST62 8 559 840 1399 560 860 1420 0.99

ST72 8 790 1260 2050 800 1300 2100 0.98

Site 6 ST61 8 731 1070 1801 490 1140 1630 1.1

Site 7 ST71 8 1055 1125 2180 800 1650 2450 0.89


between axial compressive and tensile capaci-ties of tested piles at Sites 6 and 7, indicates that the compressive capacities of piles ST61 and ST71 were about 13% to 14% higher than the uplift capacities of piles ST62 and ST72. The lower end bearing resistance of helical piles loaded in tension is likely due to the smaller projected area of the upper bearing surface of the bottom helix compared to the growth area of bottom helix (including soil plug area) for piles loaded in compression.

Another comparison between piles tested in compression and tension can be made using the stiffness at the pile head defined as the slope of load-displacement curves. The stiffness of the pile loaded in compression and tension (Figs. 10a and b) was similar at the early stages of loading. However, at high loads toward the end of loading, the stiffness of the pile in compres-sion was higher than that of the pile loaded in tension. The slope of the secondary linear part of load displacement curves of piles ST61 and ST71 loaded in compression was steeper than those of piles ST62 and ST72 tested in tension which exhibited near horizontal line (i.e. plung-ing failure).

Installation Torque – Pile Capacity Relationship

Empirical relationship between measured torque during installation and pile capacity is widely used in the industry in North America especially for small size helical piles. The empir-

ical relationship can be expressed as (Hoyt and Clemence, 1989; CFEM, 2006, Perko 2009):

Qt = K

tT [6]

whereK

t = empirical factor; and

T = average installation torque at the last 300 mm (1 ft) of pile installation

Torque-load correlation factor, Kt for compres-

sion and uplift loading found in this study are presented in Table 8. It can be seen from Table 8 that the ratio of torque to compressive ca-pacities for piles with single and double helices varied between 9.0 m-1 (2.7 ft-1) and 5.5 m-1 (1.7 ft-1). The values of K

t for piles tested in tension

(uplift) varied between 6.2 m-1 (1.9 ft-1) and 4.8 m-1 (1.4 ft-1). Therefore there is a considerable variance between torque factors K

t in compres-

sion and in tension. The values of Kt for pile

ST6 with a shaft diameter of 324 mm (12.8 in) and double helices, and pile ST7 with a single helix (and similar shaft diameter) were 9 m-1 (2.7 ft-1) and 7.3 m-1 (2.2 ft-1). The difference in torque factor of both piles was about 24%. As per the Canadian Engineering Foundation Manual (2006), it was suggested to use K

t of 3

m-1 (0.9 ft-1), which is considerably lower than the measured values in the current study.

Based on the test results presented in this study, a simplified torque-pile capacity relation-ship was not attainable. There are many factors that contribute to the exerted torque during

[TABLE 8] Summary of Torque-Capacity Factors

Test ID Pile TypeShaft

Diameter

mm

No of Helices

Installation Torque at end of installation

kN.m

Embedment Depth

m

Axial Capacity

kN

Kt

m-1

Site 5

ST5 4 324 2 211.5 5.9 1195 5.7

ST6 4 324 2 211.5 5.7 1912 9.0

ST7 3 324 1 211.5 5.7 1540 7.3

ST13 5 406 1 338.3 5.8 2292 6.8

ST14 5 406 1 338.3 5.6 1680 5.0

ST15 7 508 1 338.3 5.4 2400 7.1

Site 6ST61 8 406 2 297.7 14.1 1630 5.5

ST62 8 406 2 297.7 14.3 1420 4.8

Site 7ST71 8 406 2 338.3 18.5 2450 7.2

ST72 8 406 2 338.3 18.5 2100 6.2


installation. Some of which include pile configu-ration, soil conditions operator skill level, and accuracy of measurements. Pile configurations such as shaft size, shape of pile shaft, number of helices, diameter of helix, and pitch size are some parameters that affect the torque mea-surements. Soil conditions and ground water level have a considerable effect on pile instal-lation and recorded torque values. Examples include the presence of cobbles or boulders during installation which results in a sharp rise in torque values which is not necessarily an indication of better soil conditions. Installation procedure such as applying down pressure on pile, use of predrilling process and speed of rotary head are other factors that impact torque measurements. Method of measuring torque using either a differential hydraulic pressure along with a mechanical device to assess torque, or using an electronic load cell attached to pile head and frequency of calibrating torque measurement devise are other parameters that affect the quality and reliability of torque mea-surements.

Perko (2009) proposed an empirical relation-ship between K

t and effective shaft diameter

(deff

) based on exponential regression analysis of over than 300 load tests in both compres-sion and tension. The empirical equation can be expressed as:

Kt = [7]

d0.92 eff

where

= fitting factor equal to 1433 mm0.92/m

deff

= shaft diameter for round shaft, mm

Torque factor Kt for piles with shaft diameters

of 324 mm, 406 mm and 508 mm (12.8 in, 16 in and 20 in) considered in the present study were 3.9, 3.2 and 2.6 m-1 (1.2, 1.0 and 0.8 ft-1), respectively. The values obtained from equation 6 were considerably lower than the measured values in Table 8. Moreover K

t values obtained

from Equation 6 did not differentiate between compression and tension loading. Therefore, in the absence of a reliable torque-pile capacity correlation that can be used to assess the axial capacities of large diameter helical piles, it is suggested to use torque values only as a quali-tative measurements.

CONCLUSIONSA full scale pile load testing program was car-ried out for large-diameter helical piles installed into cohesive soils to validate their intstallabil-ity into very stiff to very hard clay/clay till material and to investigate their performance under axial compressive and uplift loading conditions. A total of ten axial load tests were performed including six compression tests and four tension tests. Eight tests were carried out using quick test method while two tests were performed using maintained test method to evaluate the creep characteristics. The finding of this study can be summarized in the follow-ing conclusions:

Helical piles with a relatively large shaft and 1. helix diameters were successfully installed into very stiff to very hard clay and clay till materials.

Predrilling process was successfully used to 2. facilitate pile penetration through very hard soil layers. However the diameter of the pi-lot hole should be at smaller than the shaft diameter to ensure intimate contact between pile shaft and surrounding soils and reduce soil disturbance.

The load-displacement curves of piles tested 3. in compression displayed typical trends including an initial linear segment, followed by a highly non-linear segment and a sec-ondary near linear segment with less slope. Helical piles with double helices provided about 25% higher resistance compared to piles with a single helix. Therefore, the use of an additional helix is a cost-effective method for increasing the axial capacities of helical piles.

Creep had a considerable effect on the 4. axial compressive load test results for piles installed in cohesive soils presented in this study. In order to reduce the creep effects on the axial capacities of piles tested, it is suggested to modify the ASTM Standards for testing piles by increasing time increments to 20 minutes for the quick test method.

The load-displacement curves for piles sub-5. jected to uplift loads indicate that a typical trend consisted of an initial linear segment followed by nonlinear segment followed by asymptote (plateau). The axial uplift capaci-ties of helical piles are typically lower than the axial compressive capacities.


Davisson failure criterion provided a con-6. servative pile capacity estimate compared to 5% criterion especially for short piles. The 5% failure criterion provided more reason-able estimates for the axial capacities and were in closer agreement with the plunging failure (true failure) for axial tensile tests.

A comparison between compression and 7. uplift load tests suggested that similar unit skin friction was developed for both com-pression and uplift load tests. Therefore similar adhesion factors may be used for both axial compression and tension load conditions. However, the end bearing com-ponent of the uplift capacities of the helical piles was controlled by the soils above the top helix.

The capacity of helical piles may be esti-8. mated based on the bearing capacity theory using the individual bearing method where, the ultimate axial capacity of the pile is the sum of the individual bearing capacities of all helices and the shaft resistance.

A simple and reliable correlation between 9. torque and axial capacity of helical piles is not presently avialable. Moreover, the results of the study indicate that torque fac-tors for both axial compression and uplift load conditions were different.

A correlation between the sleeve friction of 10. the the CPT soundings and torque measure-ments may exist. However a simple empiri-cal torque-axial capacities relationship may be difficult to achieve due to several factors that contribute to torque measurement such as method of pile installation, type of torque measurements, variability in soil conditions and pile configurations.

ACKNOWLEDGEMENTSThe author would like to thank Almita Piling Inc., for allowing use of the test data and Imperial Oil for the opportunity to carry out the load testing program at their site. The author would like to thank Ms. Kimberly Steward of Imperial Oil, Mr. Chris Palanque of FLUOR Canada Ltd. and Almita field staff for their patience, attention to details, as well as their careful installation and load testing methods for piles considered in this study. In particular the author would like to thank Messrs, Mathew Young and Scott Blackie.

REFERENCESASTM D 1143/D 1143M – 07. 2007. 1. Standard test methods for deep foundations under static axial compressive load. Annual Book of ASTM Standards.

ASTM D 3689-07. 2007. Standard test 2. methods for deep foundations under static axial tensile load. Annual Book of ASTM Standards.

Bobbitt, D.E. and Clemence, S.P. 1987. 3. Helical Anchors: Application and Design Criteria, Proceedings of the 9th Southeast Asian Geotechnical Conference, Bangkok, Thailand: 6-105 to 6-120.

CFEM. 2006. Canadian Foundation 4. Engineering Manual. 4th Edition. Canadian Geotechnical Society, Technical Committee on Foundations, BiTech Publishers Ltd., Richmond, BC.

Das, B.M. 1990. Earth anchors. Elsevier, 5. Amsterdam.

Das, B.M. and Seeley G. R. 1975. Breakout 6. resistance of horizontal anchors. Journal of Geotechnical Engineering Division, ASCE, 101(9): 999–1003.

Hoyt, R.M., and Clemence, S.P. 1989. 7. Uplift capacity of helical anchors in soil. Proceedings of the 12th International Conference on Soil Mechanics and Foundation Engineering, Rio de Janerio, Brazil, Vol. 2, pp. 1019-1022.

Hoyt, R., Seider, G., Reese, L. C., and Wang, 8. S. T. 1995. Buckling of helical anchors used for underpinning: Foundation upgrading and repair for infrastructure improvement. Edited by William F. K. and Thaney, J. M. Geotechnical Special Publication No. 50, ASCE, pp. 89-108.

Kulhawy, F.H. and Hirany, A. 2009. 9. Interpreted failure load for drilled shafts via Davisson and L1-L2. Proceedings of the 2009 International Foundation Congress and Equipment Expo, Orlando, Florida, GSP No. 185, pp. 127-134.

Kulhawy, F.H. and Hirany, A. 1989. 10. Interpretation of load tests on drilled shafts; Part 2: Axial uplift. Proceedings of Foundation Engineering: Current Principles and Practices, ASCE, Vol. 2, pp. 1150-1159.


Meyerhof, G.G., and Adams, J.I. 1968. The 11. Ultimate uplift capacity of foundations. Canadian Geotechnical Journal, V(4): 225–244.

Mitsch, M.P., and Clemence, S.P. 1985. The 12. Uplift capacity of helix anchors in sand. Proceedings of Uplift Behavior of Anchor Foundations in Sand, Detroit, Michigan, ASCE, pp. 26-47.

Perko, H. A. 2009. Helical Piles: A Practical 13. Guide to Design and Installation. John Wiley & Sons. New York, N.Y.

Sakr, M. 2011. Installation and performance 14. characteristics of high. capacity helical piles in cohesionless soils. DFI Journal, 5(1): 20-38.

Sakr, M. 2009. Axial and lateral behaviour 15. of helical piles in oil sand. Canadian Geotechnical Journal, 46(9): 1046-1061.

Sakr, M., Mitchells, R., and Kenzie, J. 2009. 16. “Pile load testing of helical piles and driven steel piles in Anchorage, Alaska”. Proceedings of the Uplift Behavior of anchor Foundations in Soil. Submitted to DFI 34th Annual Conference On Deep Foundations, October 21 – 23, 2009, Kansas City, Missouri, USA.

Vesic, A.S. 1971. Breakout resistance of 17. objects embedded in ocean bottom. Journal of Soil Mechanics and Foundation Division, ASCE, 97(SM 9): 1183–1205.

Zhang, D.J.W. 1999. Predicting capacity of 18. helical screw piles in Alberta soils. MSc. Thesis, The University of Alberta, Edmonton, Alberta, Canada.


Results of Dynamic and Static Load Tests on Helical Piles in the varved clay of MassachusettsJorge Beim, Pile Dynamics, Inc., Cleveland, OH, USA, Ph: 440-996-0629, [email protected]

Severino Carlos Luna, Geotechnical Consultants, Inc., Boston, MA, USA

ABSTRACTHelical Piles are increasingly being used to support and rehabilitate structures subjected to compressive axial loads. Traditionally, empirical correlations between installation torque and load capacity of helical screw piles are used as the field quality control. Load tests, however, are always recommended to validate the installation process and design methodology.

This paper describes a test program performed on helical piles installed at the National Geotechnical Experimentation Site of the University of Massachusetts - Amherst campus (UMass-Amherst), well known for its deep varved clay deposit. Seven of these piles were submitted to Dynamic Load Tests using the Pile Driving Analyzer® (PDA), and also to Static Load Tests. Good agreements were obtained between the results of the dynamic and static load tests, encouraging the use of the former testing method as a viable alternative to the latter for determining the compressive load capacity of helical piles installed in cohesive soils. The results of the load tests were also compared to the predictions based on torque correlations, resulting in correlation coefficients which were higher than the ones traditionally used.

Hanna (1991) and Perko (2000), among others, established the theoretical background for the design of Helical Piles. To verify the bearing capacity of those piles the industry relies mainly on torque correlations. The use of empirical torque correlations, however, is viewed with reservation by some engineers, who see a dependency of the procedures adopted by the installer on the results. Also, some engineers have misgivings about determining the capacity of helical piles using only torque measurements, without taking into consideration geotechnical parameters (Cannon, 2000).

It is recognized that the use of traditional geotechnical design methods, such as individual bearing and cylindrical shear methods, can predict the helical pile capacity (Perko, 2009). Load tests, however, are necessary to establish the pile capacity to torque ratio (Kt), used to control the capacity of the production piles.

This requirement motivated the present research project, which submitted Helical Piles to both Dynamic Load Tests (DLT) and to Static Load Tests (SLT). The results of the two kinds of load tests were compared, and the correlation of the load tests with torque measurements was also investigated.

INTRODUCTIONHelical piles were invented by Alexander Mitchell in 1836 and have been used mainly to resist tension as anchor foundation elements. In the last decade, however, they have increasingly been used to support and rehabilitate structures subjected to compressive axial loads.

Helical piles consist of one or more circular helical plates affixed to a central shaft of smaller diameter. They are fabricated from steel and the helices are generally attached to the shaft by welding and then galvanized for extra protection against corrosion.

There are several advantages associated with the use of helical piles. For example, they can be installed using lightweight, sometimes hand-operated equipment, in small and limited spaces; also, the installation does not produce residues, excessive vibrations, or disruptive noise, and is unaffected by the groundwater table. The ease and flexibility of their installation and the cost effectiveness of the solution spurred the growth of the helical pile industry.

The works of Wilson (1950), Robison & Taylor (1969), Hoyt & Clemence (1989), Ghaly &


crust that was formed as a result of surface erosion, desiccation, ground water fluctuations and other physical and chemical processes (Lutenegger and Dearth, 2004).

Fig. 1 shows the results of seismic piezocone tests with dissipations, termed SCPTù (NCHRP, 2007), performed in the upper 15 m at the NGES of the University of Massachusetts – Amherst campus.

The CPTU profile in Fig. 1 clearly shows the upper silty-clay fill and the underlying stiff clay crust to approximately 4 m (13 ft). A sharp decrease in soil strength can be seen below this depth, which marks the boundary with the underlying soft lacustrine varved clay. Typical index properties of the natural soft clay are: liquid limit ≈ 58, plasticity index ≈ 32, natural water content ≈ 65, clay fraction (CF < 2μ) ≈ 50 and OCR ≈ 2 (Mayne et al 1999). The mean annual ground water level in the NGES site is about 1.3 m (4.3 ft) (Lutenegger and Dearth 2004). The exact water level depth at the time of the tests was not measured.

DESCRIPTION OF THE PILESAll piles have the same helices and shaft configuration. The shaft is a circular steel tube with 73.0 mm (2-7/8 in) outside diameter and 5.5 mm (0.217 in) wall thickness. The 2.13 m (7 ft) long bottom sections have three circular pitched bearing plates (helices), the first one

DESCRIPTION OF THE SITEThe National Geotechnical Experimentation Site (NGES) of the University of Massachusetts - Amherst campus (UMass-Amherst) is situated in the Connecticut River Valley and within the region of the former glacial Lake Hitchcock. The lacustrine sediment deposits originated as a result of an ice-wall dam, which formed across the valley in northern Connecticut creating seasonal deposition and settling of fine-grained particles over the coarser glacial till for a period of approximately 4000 years. This soil is locally known as Connecticut Valley Varved Clay (CVVC) and extends from Northern Vermont to Central Connecticut in the present Connecticut River Valley. The individual varves of the CVVC are on the order of 2 to 8 mm (0.08 to 0.31 in) in thickness and are generally horizontally layered.

One of the reasons for choosing this particular site was that its stratigraphy and soil properties have been extensively studied and are very well documented (Lutenegger, 2000). The tests were performed in Area B, which consists of approximately 1.5 m (5 ft) of stiff silty-clay fill overlaying a thick 30-m (98 ft) deposit of late Pleistocene lacustrine varved clay. The fill consists of CVVC placed about 30 years ago after excavations at the Town of Amherst Wastewater Treatment plant, which is adjacent to the site. Below the fill, the CVVC has a well-developed stiff over-consolidated

0

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[FIG 1] Seismic piezocone results at NGES in Amherst, Massachusetts (after DeGroot and Lutenegger 1994)


to facilitate the helices in “biting” into the soil and advancing the downward movement of the pile. To minimize the disturbance to the soil during installation, the helical pile should be advanced into the ground at a rate of one pitch distance per revolution, and multiple helices should be spaced along the shaft in multiples of the pitch, such that subsequent helices follow the same path as the initial helix when penetrat-ing the ground (Ghaly et al., 1991).

A field control installation log was recorded re-lating the pile depth to the torque and the num-ber of revolutions for every 0.3 m (1 ft) penetra-tion; this is the typical inspection procedure for helical pile projects. Fig. 3 shows the graphs of the torque measured during installation of the test piles (including that of control pile HP-4), versus depth.

[FIG. 3] Torque measured during installation versus depth

TORQUE CORRELATIONSThe concept of correlating installation torque to axial capacity for helical piles is analogous to the relationship between the pile driving effort and pile capacity. Indeed, several authors have attempted to express such an empirical torque to capacity relationship (Hoyt and Clemence, 1989; Narasimha Rao et al., 1989; Ghaly and Hanna, 1991; Perko, 2000; 2009). Those torque formulas have been used in the helical pile in-dustry for many years; however, they do not ex-plicitly consider soil profile or soil parameters and therefore are too crude for advanced design requirements. Once calibrated by load testing, however, the torque formulas are a convenient and accepted Quality Control procedure for production piles.

The widely used acceptance criteria for heli-cal piles AC-358 (ICC Evaluation Service, 2007), includes a formula, initially proposed by Hoyt and Clemence (1989), relating the final instal-lation torque (T) to the ultimate axial capacity

with a diameter of 203.2 mm (8 in) welded 152 mm (6 in) above the tip, the second with a diameter of 254 mm (10 in) welded 610 mm (24 in) above the first, and the third one with a diameter of 305 mm (12 in) welded 762 mm (30 in) above the second. The space between the helices is therefore three times their diameter, as is usual practice. The pitch distance of the helices is 76.2 mm (3 in), and the thickness of the helix plates is 12.7 mm (1/2 in). The extensions are 2.13 m (7 ft) long, with 0.15 m (6 in) overlap connections, resulting in 1.98 m (6.5 ft) effective extension length. Because of the temporary function of the piles the steel was not galvanized, as would be the usual practice.

Three piles (HP-5, HP-10 and HP-15) were installed to a depth of 3.7 m (12 ft) below grade, and consisted of one bottom section and one extension. Five piles (HP-4, HP-7, HP-9, HP-12 and HP-14) were installed to a depth of 5.5 m (18 ft) and consisted of one bottom section and two extensions. It was expected that the shorter piles would show more resistance than the longer ones, due to the drop in soil resistance at 4 m (13 ft).

INSTALLATIONA total of 18 piles were installed at the NGES of UMass-Amherst on the 2nd and 3rd of July of 2010. Ten piles were used to verify differ-ent procedures of installation and to serve as reaction piles. Eight piles were designated as test piles: one of them was used as a control pile with only a SLT performed, and the other 7 helical piles were subjected to both DLT and SLT. The relative location of the different piles is shown in the pile installation plan of Fig. 2.

[FIG. 2] Pile installation plan

The installation of helical piles consists of em-bedding them into the soil by applying a torque to the head of the central shaft, which causes the helix or helices to penetrate the ground in a “screwing” motion. A downward force may also be applied to the helical pile during installation


(Q), and defines a torque correlation factor (Kt) such that:

Q = Kt T [1]

The Kt factor is an empirically developed quan-

tity which typically ranges from 23 m-1 to 32.8 m-1 (7 ft-1 to 10 ft-1). For the 73.0 mm (2-7/8 in) outside diameter round shaft (considered a “conforming” system) AC-358 suggests a K

t of

29.5 m-1 (9 ft-1); however, AC-358 also states that the parameter K

t shall be verified by full-scale

field installation and load tests.

It should also be noted that only tension tests were analyzed by Hoyt and Clemence (1989), as a basis for the torque correlation currently in use. More compressive tests on piles installed in various types of soil are necessary so that statistical analyses can be performed to obtain correlations between the tension and compres-sion capacity of helical piles.

DYNAMIC LOAD TESTS DLTs are performed by installing strain gages and accelerometers close to the top of the pile, and recording the resulting force and veloc-ity signals while the pile is impacted by a pile driving hammer or similar device (Rausche et at., 1972). It is now extensively used with most kinds of foundation piles, and has the advantage over SLTs of being much faster and of lower cost. Although numerous good agree-ments between the results of SLTs and DLTs on other kinds of piles have been published (Likins, Rausche, 2004), the use of DLTs with helical piles had little application to date. One of the objectives of this work was therefore to verify the viability of using DLTs as a routine way of determining the bearing capacity of helical piles.

The DLTs were performed on September 15, 2010, about 2 ½ months after installation of the piles. Because of the small diameter of the shaft, the conventional bolted strain transduc-ers normally used for dynamic pile testing could not be employed in this case. A calibrated 1.07 m (3.5 ft)-long piece of extension rod with bonded strain-gages was used instead. Two conventional piezo-resistive (PR) accelerom-eters were bolted to the rod, and the strain and acceleration signals were sent to a Pile Driving Analyzer (PDA) by means of radio transmitters. Fig. 4 shows the instrumented rod with sensors and radio transmitters attached to the top of the pile. It also shows the type of bolted overlap connector used to assemble the rod string.

[FIG. 4] Dynamic Load Test sensors arrangement

The tests consisted of a few blows applied by a truck-mounted cathead SPT rig fitted with a 1336 N (300 lbs) weight, dropped from a maximum height of 0.9 m (3 ft). Fig. 5 shows the setup used. The ideal ram weight and drop height is a function of the test load, shaft size and failure criterion, among other parameters, and can be determined for example by wave equation analyses (Rausche, 2000).

[FIG. 5] Dynamic Load Test setup

A CAPWAP® analysis (Rausche et al., 2010) was performed for each pile, using the data from the blow with the highest energy. CAPWAP uses signal matching procedure, based on a modified and extended Smith soil model, to determine the static and dynamic soil resistance param-eters; it also provides a simulated static pile-top load versus displacement curve which can be directly compared with the corresponding curve from a SLT.

The pile was modeled as a uniform rod, with a cross-sectional area corresponding to that of the 73.0 mm (2-7/8 in) shaft. The impedance increases caused by the helices were neglected, since they would be very small for the 1 m (3.3 ft) long pile segments of the numerical model. Small reductions in impedance or the use of “slacks” were necessary to model the joints

Strain gages

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instrumented rod

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along the shaft. No additional end bear-ing was modeled to account for the upper two helices. The good signal matching and good agreements of DLT and SLT results obtained using this model suggest that, at least for this type of soil, there was mini-mal soil resistance developed at the bot-tom of the two upper helices. Practically all soil resistance above the lower helix apparently originated from shearing of the soil plug trapped between the helices. This is in agreement with the method for predicting the axial compressive capacity of helical piles proposed by Livneh and El Naggar (2008).

A shaft radiation damping model (Likins et al., 1996) was used in the CAPWAP analyses. In order to obtain good agree-ment with the static load test results, how-ever, lower values for the damper had to be used, compared to those recommended in the CAPWAP documentation. This was attributed to the fact that the effective circumference of the skin friction behavior, controlled by the helix diameters, was larger than the value entered of the outside perim-eter of the 73.0 mm (2-7/8 in) pipe.

Fig. 6 shows the summary table of CAPWAP results for pile HP-7, taken as an example, and Fig. 7 shows the measured force and velocity records, together with a diagram of the pile model showing the im-pedance decreases at the joint locations.

STATIC L OAD TESTSThe SLT is the traditional way of verifying the bearing capacity and displacement behavior of deep foundations; however, it requires a rela-tively large reaction system and long prepara-tion time, which may have unfavorable economi-cal impact on the construction. On the other hand, SLTs are still the standard with which other test methods are compared, so they were performed in this case for comparison purposes.

The SLTs were performed between October 10 and 15, 2010, i.e., 25 to 30 days after the DLTs. They were carried out according to Procedure A (Quick Test) specified by ASTM D1143/D1143M - 07, with measurement of displacements for load increments of 5% of the anticipated failure load. Each increment was kept constant for 15 minutes, until what was perceived as plunging failure was achieved in all cases. Fig. 8 shows

Force

Velocity

[FIG. 8] Static Load Test setup

[FIG. 6] CAPWAP Summary Table for HP-7

[FIG. 7] Force-Velocity record for pile HP-7, and pile model showing impedance decreases at joint locations

the typical SLT setup, including reference beam, dial gages and hydraulic pressure gage as em-ployed for all piles.

The results of the SLTs are discussed below, together with those of the DLTs.


RESULTS Figs. 9 to 15 show the comparison for each pile between the load-displacement curve of the SLT and the corresponding curve simulated by the CAPWAP program. The figures also show the Davisson offset limit (Davisson, 1972). This method defines the failure load as that cor-responding to the movement that exceeds the elastic deformation of the pile by a value of 3.8 mm (0.15 inch) plus the diameter of the pile (in the present case taken as that of the largest helix, that is, 305 mm or 12 inches) divided by 120. The elastic deformation of the pile is calcu-lated using the expression:

D = PL / AE [2]

Where D is the deformation for the applied load P, L is the length, A is the cross-sectional area of the shaft and E is the elastic modulus of the pile material. Table 1 summarizes all results in numerical form.

[FIG. 9] Results for HP5 (3.7 m)








The Davisson criterion was chosen since most comparisons of DLT and SLT results available in the literature were obtained using this method

(Likins and Rausche 2004), and it is also one of the most widely used methods in North America (Fellenius, 2001). In order to guarantee the best comparisons between DLT and SLT results interpreted by the Davisson criterion, the usual DLT practice requires that the energy applied is sufficiently large to produce a set per blow of 2.5 mm (0.1 in) or more (Likins et al, 2000). For the shorter higher capacity piles, however, the sets were of the order of only about 2 mm (0.08 in) , and the results shown in Figs. 9 to 11 are clearly indicating that the energy applied was not sufficient to mobilize the full static capacity. This explains the somewhat conservative DLT results for piles HP10 and HP15. The SLT simulation curves predicted by the CAPWAP analysis of the DLT data, however, are in good agreement with the actual SLT, up to the point of the maximum DLT displacement achieved. A higher applied energy would have been recommended for the 3.7 m (12 ft) piles, by using a heavier weight and/or increasing the drop height.

Besides the Davisson criterion, several other definitions of pile capacity evaluated from load-movement records of SLTs can be found in the literature (Fellenius, 2001). A criterion specifically for helical piles has been proposed by Livneh and El Naggar (2008), defining the failure load as that corresponding to the movement that exceeds the elastic deformation of the pile (calculated using expression 2 above) by 8% of the largest helical diameter. For drilled shafts, the Federal Highway Administration (FHWA) for example defines the failure load as that corresponding to a gross settlement of 5% of the diameter (O’Neill and Reese, 1999). Fig. 16 below shows the Load-Displacement curves

[TABLE 1] Summary of results

Pile HP5 HP10 HP15 HP4 HP7 HP9 HP12 HP14

Depth (m) 3.7 3.7 3.7 5.5 5.5 5.5 5.5 5.5

Torque (kN-m) 1.9 2.1 2.3 1.5 1.2 1.2 1.2 1.5

SLT - Davisson (kN) 102.8 97.9 101.9 50.0 38.3 48.5 52.0 56.5

DLT - Davisson (kN) 100.5 85.9 81.0 - 38.3 49.8 49.4 56.5

SLT/DLT 1.02 1.14 1.26 - 1.00 0.97 1.05 1.00

Kt static (m-1) 53.1 46.8 44.2 33.3 32.5 41.2 43.1 37.9

Kt dynamic (m-1) 51.9 41.0 35.2 - 32.5 42.3 40.9 37.9

Average Kt static (m-1) 48.0 37.6


of the SLTs, and the thresholds corresponding to three failure criteria: Davisson, FHWA (5% of largest helical diameter) and Livneh-El Naggar.

It can be seen that in this case the Davisson criterion produced more conservative results than the two other criteria. For longer piles and/or smaller diameter helices, it is possible that the FHWA criterion produces similar or even more conservative results than the Davisson criterion. The Livneh-El Naggar criterion, on the other hand, calls for extending the load application to much larger displacements. In any event, if the results of the SLTs are to be analyzed using a less conservative criterion, like the Livneh-El Naggar, the energies applied in the DLT would have to be increased, so that permanent sets larger than 2.5 mm (0.1 in) are produced.

The question of the failure criterion that should be adopted for DLTs on helical piles is one that requires further investigation. For drilled shafts and augercast piles, for example, an alternative criterion to the 2.5 mm (0.1 in) minimum set has been proposed based on the total accumulated toe displacement caused by successive blows with increasing energy (Rausche et al., 2008). A suitable criterion will have to be developed for helical piles, considering their particular characteristics and installation method.

Additional research is also necessary to verify the agreement of SLT and DLT results for piles in non-cohesive soils. At least one case study presented in the literature (Cannon, 2000) suggests that for moderately dense to dense medium sands the skin friction shows resistance concentrations at the helix locations,

whereas the model used in the analyses presented herein only shows a resistance concentration at the location of the leading helix (included in the end bearing as, for example, shown in the CAPWAP summary table of Fig. 6).

CONCLUSIONS Dynamic and Static Load Tests were performed on seven helical piles installed in the NGES at Amherst, Massachusetts. Good agreements were obtained between the results of the SLTs interpreted by the Davisson criterion and the results of the DLTs, and the general shapes of the load-displacement curves are similar for both

tests. This shows that DLT is a viable alternative to determine the compressive load capacity of Helical Piles, at least in cohesive soils, with added advantages of lower cost and execution time.

Although AC-358 suggests a torque to capacity correlation factor K

t of 29.5 m-1 (9.0 ft-1) for the

size of piles tested, the results obtained in this research show that for compressive loads on CVVC this factor proved to be too low, leading to overly-conservative design. This confirms the necessity of executing one or more load tests on each site for determining the best correlating value of K

t. It should also be noted that torque

correlations depend on the particular pile size and configuration and also on the type and condition of the soil where the piles are installed. This was also shown in the present study, where substantially different values of K

t

were obtained for piles installed in the different bearing strata.

ACKNOWLEDGMENTS This research had the support of Ideal Foundation Systems who supplied the helical piles. Terratec Construction, Inc. and Monponset Erectors, Inc. provided installation and field support. Umass-Amherst and Dr. Alan Lutenegger granted the access of the site and provided some of the references for this study.

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[FIG. 16] Results of the Static Load Tests


ASTM D4945-08. Standard Test Method 2. for High-Strain Dynamic Testing of Deep Foundations. ASTM International, 100 Barr Harbor Drive, West Conshohocken, PA 19428-2959, United States

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Perko, H. A., 2009. Helical Piles: A Practical 20. Guide to Design and Installation. 1st Edition by John Wiley & Sons Ed., 512 p.


Rausche, F., Moses, F., Goble, G. G., 1972. 21. Soil Resistance Predictions From Pile Dynamics. Journal of the Soil Mechanics and Foundations Division, American Society of Civil Engineers. Reprinted in Current Practices and Future Trends in Deep Foundations, Geotechnical Special Publication No. 125, DiMaggio, J. A., and Hussein, M. H., Eds, August, 2004. American Society of Civil Engineers: Reston, VA, pp. 418-440.

Rausche, F., 2000. Wave Mechanics and the 22. Wave Equation. The Design and Installation of Higher Capacity Driven Piles, Pile Driving Contractors Association, Annapolis, MD.

Raushe, F., Likins, G.E., Hussein, M.H., 2008. 23. Analysis of Post-Installation Dynamic Load Test Data for Capacity Evaluation of Deep Foundations. From Research to Practice in Geotechnical Engineering

Rausche, F., Likins, G. E., Liang, L., Hussein, 24. M.H., 2010. Static and Dynamic Models for CAPWAP Signal Matching. The Art of Foundation Engineering Practice, Geotechnical Special Publication No. 198, Hussein, M. H., J. B. Anderson, W. M. Camp Eds, American Society of Civil Engineers: Reston, VA, pp. 534-553.

O’Neill, M.W., Reese, L.C., August 1999. 25. Drilled shafts: construction procedures and design methods, FHWA-IF-99-025. Federal Highway Administration, McLean, Va.

Robison, K. E. and Taylor, H., 1969. 26. Selection and Performance of Anchors for Guyed Transmission Towers. Canadian Geotechnical Journal, Vol. 6, pp. 119-135.

Wilson, G., 1950. The bearing capacity 27. of screw piles and screwcrete cylinders. Journal of Institution of Civil Engineers. Vol. 34, pp. 4-73.


Underwriters: Gold

Underwriters: Silver


DFI Journal Paper Review Process

The peer review process for documents considered for publication in the DFI Journal is still evolving. The following is a description of the current process, however, the publication is still in its infancy and the review process is still in a state of flux. DFI reserves the right to alter the procedures as necessary.

Paper SubmittalPapers may be submitted at any time. Authors wishing to submit their papers for consideration of publication in the DFI Journal are invited to access www.dfi-journal.org. The website will ask for a login or, for new submitters, will ask for creation of an account. Once logged in the author must upload a full paper in MS Word format as well as any ancillary files such as figures, photos and other graphics which are included in the paper. The paper is then converted to a PDF file which the author must approve before the paper will be released to the publisher and journal editors for viewing. The journal editors preliminarily review the paper for relevancy to the Journal mission.

Paper Review The journal editors assign those papers deemed to be worthy of consideration for Journal publication to the appropriate editorial board member, which currently consists of DFI technical committee chairmen and other industry leaders, so that appropriate reviewers for the paper topic can be obtained. Reviewers are chosen based on their knowledge, areas of expertise, and qualifications to act as a reviewer on the particular subject matter of the paper in question. At least three reviewers will be assigned to each paper.

After the reviewers are selected, they are provided with instructions and a password for entry into the website where they can view the paper PDF and submit their evaluation. The criteria on which they base their review fall under two areas: technical content and quality of paper presentation. The criteria for technical content include relevancy, originality, appropriate references to support statements, significance of results and exclusion of personal opinion and commercialism. The criteria for paper presentation include quality of figures, quality of English language, paper organization and completeness. The reviewers enter their evaluation by responding to a number of questions rating the paper as well as entry of comments to authors. They are also required to make a recommendation to the journal editors of: accept as is; accept with mandatory changes; or reject. The author is advised by automatic email of the posting of reviews and he/she can access the reviews and respond and/or modify the paper to satisfy comments by the reviewers. A second round review can then take place if necessary, ultimately leading to second round reviewer recommendations. The publisher and editors, acting as a final review committee, make the decision, based on the reviewers’ recommendations, as to acceptance of the paper for publication in the next issue of the journal or in a subsequent issue.

Throughout the process, automatic emails are sent out to reviewers when papers are ready for their review and to the authors to keep them aware of the progress of their paper.

Paper Finalization Upon acceptance, the final paper submission by the author and all graphic files are downloaded by the publisher for processing and formatting for publication. The publisher is provided with proofs by the production house and these are edited to ensure acceptable layout, the absence of typos, clarity of figures, etc. In most cases the author(s) are provided with a final PDF for their review and approval.


DFI Journal Call for Papers

The Deep Foundations Institute compiles and publishes a Journal of practical and technically rigorous papers on a bi-annual schedule. The DFI Journal is distributed to 2,500+ DFI members plus non-member subscribers.

The DFI Journal content is subject to quality technical review, and must meet a standard in quality on practical subjects dealing with case studies, deep foundations history, design, construction, testing, innovations and research in the field.

Each journal consists of at least five documents collected from technical papers that are invited or selected from papers submitted by international industry members based on this call. Papers presented at the DFI Annual Conference and Specialty Seminars may be included if expanded to the Journal standard and review process.

The editors are herein sending out a call for original papers for consideration of inclusion in the upcoming journals. Full draft papers up to 15 pages in length are to be submitted to: http://www.dfijournal.org for review. Authors will be required to create a login account and will be notified via email on the status of their submission.

Papers are solicited on the following topics:

Case studies involving foundation systems with technical data support• Historical evolution of deep foundations• Relationship between use of design, construction and equipment• Quality control, quality assurance and non-destructive testing• Innovation in all aspects of deep foundations and earth retention• Practice-oriented research•

The Publisher and the Journal Editorial Board will review submitted papers for acceptability for publication in the current or future issues of the Journal, subject to full peer reviews as described on the preceding page entitled "DFI Journal Paper Review Process". Authors of papers accepted for publication will be required to sign a copyright licence agreement.


2012 DFI Journal Subscriptions Electronic Subscriptions for the two-issue volume (July & November 2012) are provided to all DFI members at no cost – the electronic version is included as a benefit of annual membership dues.

DFI Members may order print versions of the DFI Journal. Non-Members can join DFI to receive the electronic version and print version at member rates OR can purchase a subscription only of either electronic, print or both.

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Deep Foundations Institute was incorporated in 1976 in the State of New Jersey as a non-profit educational activity. DFI is a technical association of firms and individuals in the field of designing and constructing deep foundations and excavations. DFI covers the gamut of deep foundation construction and earth retention systems.

Although the bulk of the membership is in North America, the Institute is worldwide.

DFI’s strengths are:

• Communication of information concerning the state-of-the-art and state of the practice of deep foundation technologies

• Offering networking opportunities for our members

• Offering opportunities for members to improve the industry through publications produced by volunteer committees

• Offering educational conferences, seminars and workshops in the industry

The core strength of DFI is the broad spectrum of its membership. All disciplines participate on an equal footing, be they contractors, engineers, owners, academicians, equipment manufacturers and distributors or materials manufacturers and suppliers. All types of foundation systems are represented, whether installed by driving, drilling or other means. This diversity and openness without bias provides a forum for the free exchange of knowledge and a platform for the development of new technology and opportunity.

DFI is:

• An international network of heavy construction professionals dedicated to quality and economy in foundation design and construction

• A forum open to all construction professionals across disciplines and borders.

• A technological association devoted to gathering, storing and disseminating practical information

• A resource for identifying and locating the specialists and sources of expertise.

• An initiator and participant in research

Deep Foundations Institute Sustaining Members are Corporate Members of DFI who have voluntarily granted funding to the Institute for expanded support of the Industry. The fund is managed by the DFI Educational Trust.

DFI Sustaining MembersAECOM USA INC.AMERICAN EQUIPMENT & FABRICATING CORP.ANDERSON DRILLINGAPE/J&MBAUER - PILECO INC.BEN C. GERWICK INC.BERKEL & COMPANY CONTRACTORS INC.BRASFOND FUNDAÇÕES ESPECIAIS S/ABRAYMAN CONSTRUCTION CORPORATIONCAJUN DEEP FOUNDATIONS LLCCASE FOUNDATION COMPANYCIPORT S.A.DEAN CONSTRUCTION CO. LTD.DEWITT CONSTRUCTION INC.DOSDOURIAN ENTERPRISES INC.FOUNDATION CONSTRUCTORS INC.FOUNDATION SUPPORTWORKS INC.FOUNDATION TECHNOLOGIES INC.GEOKON INC.HAYWARD BAKER INC.HJ FOUNDATION COMPANYKIEWIT CONSTRUCTION GROUP INC.KLEINFELDERL.G. BARCUS & SONS INC.LANGAN ENGINEERING AND ENVIRONMENTAL SERVICESMACTEC ENGINEERING & CONSULTING INC.MCKINNEY DRILLING COMPANYMENARDMORETRENCHMUESER RUTLEDGE CONSULTING ENGINEERSNICHOLSON CONSTRUCTION COMPANYNORTH AMERICAN CONSTRUCTION GROUPO.C.I. DIVISION / GLOBAL DRILLING SUPPLIERS INC.PND ENGINEERS INC.RICHARD GOETTLE INC.SAS STRESSTEEL INC.SCHNABEL FOUNDATION COMPANYTEI ROCK DRILLS INC.THATCHER ENGINEERING CORPORATIONURBAN FOUNDATION/ENGINEERING LLCWILLIAM F. LOFTUS ASSOCIATES FOUNDATION ENGINEERSWURSTER ENGINEERING & CONSTRUCTION INC.

DFI JOURNALThe Journal of the Deep Foundations Institute

Deep Foundations Institute326 Lafayette AvenueHawthorne, New Jersey 07506 USATel: 973-423-4030Fax: 973-423-4031www.dfi .org

International Standard Serial Number (ISSN): 1937-5247

Documents

DFI JOURNAL · Jet grouting has increasingly become a ground improvement technology used to address seepage concerns and provide strength improvement to soils underlying dams. The