Development and Evaluation of New Connection Systems for

Development and Evaluation of New Connection Systems forHybrid Truss BridgesKwang-Hoe Jung, Jang-Ho Jay Kim Jong-Wo Yi, , Sang-Hyu LeeJournal of Advanced Concrete Technology, volume ( ), pp.11 2013 61-79

Inelastic Performance of High-Strength Concrete Bridge Columns under EarthquakeHyunMock Shin, TaeHoon Kim MyungSeok Oh, , DaiJeong SeongJournal of Advanced Concrete Technology, volume ( ), pp.9 2011 205-220

Pseudo-Cracking Approach to Fatigue life Assessment of RC Bridge Decks in ServiceChikako Fujiyama, Xue Juan Tang Koichi Maekawa, , Xue Hui AnJournal of Advanced Concrete Technology, volume ( ), pp.11 2013 7-21

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Journal of Advanced Concrete Technology Vol. 11, 61-79, February 2013 / Copyright © 2013 Japan Concrete Institute 61

Scientific paper

Development and Evaluation of New Connection Systems for Hybrid Truss Bridges Kwang-Hoe Jung1, Jang-Ho Jay Kim2, Jong-Won Yi3, and Sang-Hyu Lee4

Received 27 August 2012, accepted 16 February 2013 doi:10.3151/jact.11.61

Abstract This study focused mainly on prestressed concrete box girder bridges with steel truss web members, also known as Hy-brid Truss Bridge (HTB). Previously, French engineers worked to improve the structural efficiency of prestressed con-crete box girders by using concrete-steel hybrid subcomponents in an effort to reduce the weight of the superstructure. As a result of these efforts, several HTBs have recently been constructed in Japan and Korea. All of these bridges have unique connection systems between the concrete slabs and web truss members, which dictate the safety as well as the failure behavior of the hybrid girders. Thus, evaluation of both construction and structural safety of the connection sys-tems regarded as an essential step in the design of HTB. In this study, four new connection systems using either hinge devices or T-type perfobonds have been proposed as improvements to the assemblage and to eliminate the need for welding during construction while satisfying load carrying capacity requirements. The static loading tests for six con-nection specimens in real scale were carried out in order to fully evaluate structural safety of the newly proposed con-nection systems. Also, a real scale railway bridge specimen with a 20m single span and an embedded hinge connection system was designed and constructed in a structural laboratory to perform for static and dynamic tests. Using this specimen, the possibility of using an embedded hinge connection system in a HTB, both for railway and general road-way usage, was evaluated.

1. Introduction

In order to enhance the structural efficiency of a PSC girder, the steel-concrete composite girders have been considered and examined by many researchers and en-gineers (Dezi et al. 2006; Chen et al. 2009; Ahn et al. 2010). Especially, to overcome the span length limita-tion caused by the relatively heavy self-weight of a PSC box girder bridge, various studies have been performed using steel web members, such as corrugated or truss steel webs (Brozzetti 2000; Ohyama et al. 2005; Ibra-him et al. 2006; Kiymaz et al. 2010; Jung et al. 2010; Jung et al. 2011; Xue et al. 2011). The theory behind this was that if the concrete webs in PSC box girders could be replaced by steel truss webs, the weight of the superstructure could be decreased by about 20%, ena-bling a span extension.

This PSC box girder bridge, composed of upper and

lower concrete slabs with steel truss members as web sections and external prestressing (PS) tendons, shown in Fig. 1, can be abbreviated as a Hybrid Truss Bridge (HTB). In recent years, multiple HTBs (i.e., Kinogawa Bridge, Shitsumi Ohashi Bridge, Sarutagawa Bridge, and Tomoegawa Bridge), as well as bridges incorporat-ing corrugated steel webs, have been constructed in Ja-pan (Aoki et al. 2005; Fujihara et al. 2005; Furuichi et al 2002; Minami et al. 2002; Tsujimura et al. 2002). In Korea, the construction of the Shinchun Bridge, an HTB type, is currently underway (Won et al. 2005; Sato et al. 2008). In addition to the Shinchun Bridge, the construc-tion of the Gangchun Bridge, another HTB type with a newly developed embedded hinge connection system in its superstructure, is recently completed. This bridge is a weir maintenance bridge crossing the Namhan River and is the first real application of this newly developed connection system. An added benefit for HTBs is their

1Chief Research Engineer, Research and Development Division, Hyundai Engineering and Construction Co. Ltd., Yongin-si, South Korea. 2Associate Professor, School of Civil and Environ-mental Engineering, Yonsei University, Concrete Struc-tural Engineering Laboratory, Seoul, South Korea. E-mail: [email protected] 3Senior Research Engineer, Ph. D., Institute Research and Development Division, Hyundai Engineering and Construction Co. Ltd., Yongin-si, South Korea. 4Research Engineer, Research and Development Division, Hyundai Engineering and Construction Co. Ltd., Yongin-si, South Korea. Fig. 1 Hybrid Truss Brige (HTB).

K.-H. Jung, J.-H. J. Kim, J.-W. Yi and S.-H. Lee / Journal of Advanced Concrete Technology Vol. 11, 61-79, 2013 62

see-through webs, which can better harmonize the bridge with the surrounding scenery, as compared to solid webs.

Also, Hybrid Truss Bridges require suitable connec-tion systems in order to ensure the safety and the stable failure behavior of the hybrid girders. Therefore, evalua-tion of the construction and structural safety of the con-nection systems between concrete slabs and web truss members have been regarded as the most essential step in the design of HTBs (Kutsuna et al. 2002; Miura et al. 2006; Sato et al. 2008; Shim et al. 2007a; Shim et al. 2007b; Josef and Martin 2009). Each connection system used in past HTBs has shown unique features and good structural performance. However, in order to improve assemblage convenience, eliminate welding during con-struction, and satisfy structural capacity, several new connection systems using hinge devices or T-type per-fobonds have been proposed and evaluated in this study.

2. Development of new connection systems

2.1 Conventional connection systems Among the main components of the HTBs, the connec-tion system is considered the most crucial component, critically affecting their safety and serviceability. Also, the basic design concept of the bridge ensures that the connection system will not fail before the concrete slab fails or the steel truss member yields. Therefore, struc-tural safety verification of the bridge connection system is an essential step. Because of this, many types of con-nection systems have been developed in an attempt to improve both connection capacity and construction effi-ciency. The connections that have been used up to now have been case specific, which provided sufficient struc-tural safety and serviceability to each bridge.

Figure 2 shows six conventional connection systems that have been implemented in actual HTBs. Figure 2(a) presents the punched box type connection system

(a) Punched box type system (b) Shear key system

(c) Double Pipes system (d) Double gusset plates systems

(e) Double Inverse T-type perfobond system (f) Group stud system

Fig. 2 Conventional connection systems.


used in the Kinogawa Bridge, which is composed of a punched box embedded in a longitudinal concrete girder along with tensile and compressive truss members. Since tensile forces are directly transferred to the con-crete slabs, the tensile truss members are welded to the top of the punched box. The compressive truss members have several ribs running around the cross-section in order to improve bond strength. These ribs are embed-ded in the punched box. The punched box plays an im-portant role in composite action, providing sufficient structural capacity (Furuichi et al 2002; Minami et al. 2002).

Figure 2(b) includes the shear key connection system as used in the Shitsumi Ohashi Bridge. This system is composed of a cylindrical shear key device designed for truss members with male and female connection taps. The truss members are connected to each other through these male and female taps and the cylindrical shear key device is also inserted at the contact surface in the longi-tudinal direction to ensure a clear connection, as well as a load transferring capacity. In addition, fill concrete and inner ribs of each truss member play an important role in composite action, providing a sufficient struc-tural capacity (Fujihara et al. 2005; Tsujimura et al. 2002).

The Kinogawa and Shitsumi Ohashi Bridges have only one type of connection system throughout them; however, the Sarutagawa, Tomoegawa, and Shinchun Bridges incorporate two different types of connection systems in order to improve web connection efficiency. The Sarutagawa and Tomoegawa Bridges have a gusset plate welded to truss members, as well as a ring type slit device, as shown in Figs. 2(d) and 2(c), respectively. A gusset plate connection system and a ring type slit de-

vice are used in both the support and mid-span portions, respectively, since the ultimate strength of a gusset plate connection system is higher than that of a slit device connection system (Aoki et al. 2005).

The Shinchun Bridge uses two different connection systems for upper and lower slabs. The upper and lower slabs are installed with an inverse double T-type per-fobond welded to truss members over the entire longitu-dinal slab length. There is also a group of studs welded to a base plate connected to truss members using a gus-set plate, as shown in Figs. 2(e) and 2(f), respectively. The inverse double T-type perfobonds placed over the entire longitudinal length provide excellent structural composite action, resulting in an overall reduction in bridge height (Won et al. 2005; Sato et al. 2008). Also, superstructures using this connection system can be constructed using various erection methods, since a lift-ing element can be applied during construction.

2.2 Development of new connection systems The connection system used in HTBs has been based on task specific requirements and has demonstrated good performance in real structural settings. However, in or-der to enhance the assemblage convenience and elimi-nate the need for welding procedure in the construction site, several new connection systems using hinge de-vices or T-type perfobonds are proposed and their per-formances are evaluated in this study.

Figure 3 shows four new connection systems, which were developed in this study. Figure 3(a) is the T-GHT connection systems, composed of the T-type perfobond welded to a base plate connected to truss members using gusset plates. The T-GHT connection system is very similar to the stud grid connection system, as shown in

(a) T-GHT (b) EHT

(c) T-EHT (d) P-EHT

Fig. 3 New connection systems.


Fig. 2(f); however, it uses T-type perfobond instead of stud grids. Although the stud grid connection system used in the Shinchun Bridge was an effective system, highly precise concrete work within narrow spacing of studs and reinforcements may not be feasible. Also, in terms of mechanics, a connection system must be able to resist tensile forces since simultaneous tensile and horizontal shear forces can occur at a connection joint. Therefore, the T-type perfobond connection system was proposed to improve the stud grid connection system by simplifying complicated connection details and also to enhance tensile force resisting capacity at connection joints.

Both conventional connection systems (Fig. 2) and T-GHT (Fig. 3(a)) have shown sufficient structural safety in the past; however, numerous welding procedures used to align truss members at joints and small geometrical error adjustment difficulties caused by steel truss fabri-cation flaws have caused these construction difficulties in the past. Therefore, this study focuses on the devel-opment of new connection systems to improve assem-blage convenience and reduce welding procedure in joint construction while achieving structural safety.

The EHT connection system (Fig. 3(b)) is a hinge type device composed of connection plates and connec-tion bolts. This system requires no welding at the con-struction site since truss members are welded to the connection plates at the factory in advance. And the assembly of these truss units comprised of truss mem-bers with welded connection plates at the construction site requires only the use of connection bolts or long rods. This system allows position adjustment of truss members and easy amendment of small geometrical errors from steel truss fabrication to facilitate the instal-lation of the truss members. Figures 3(c) and 3(d) pre-sent the T-EHT and P-EHT connection systems, which are composed of T-type and general perfobond, respec-tively, using two connection bolts. Neither T-EHT nor P-EHT requires welding at prefabrication factories or construction sites. Utilizing the holes in T-type or gen-eral perfobond, the truss members are pin-joined by connecting bolts. It has been assumed that EHT, T-EHT, and P-EHT have sufficient structural capacity without requiring welding when confined in the top and bottom concrete slabs or girders. Therefore, the structural per-formance of EHT, T-EHT, and P-EHT connection sys-tems must be verified before they can be considered as possible options in HTB construction.

2.3 Design concept of new connection systems Generally, the connection joints of HTBs are subjected to tensile and compressive forces in their truss members, horizontal shear forces on the concrete girder, and local flexural moments at the connection joints, all of which are shown in Fig. 4. Therefore, the first design consid-eration of a connection system is that the substrata members must be able to fully resist these forces placed

upon it by its self-weight, as well as live loads. In par-ticular, a connection joint safety factor of over 1.3 is recommended in order to guarantee structural safety of HTBs both for service and ultimate loads.

Another important design consideration of connection systems is that the secondary local moment at the con-nection joint must be eliminated. Other than the secon-dary local moment caused by a fixed condition at the connection joint, the only additional local moment should be produced by the eccentricity between the cen-ter of the concrete slab and the crossing point of the two axes of the joining trusses. Figure 5 indicates the local moment caused by this eccentricity for various slab ax-ial load magnitudes and clearly shows that local mo-ment increases as the eccentricity and slab axial load increase. The EHT system produces no eccentricity since the center of the concrete slab and the crossing point of the two axes of truss members coincide due to a built-in free-rotation system. In contrast, T-GHT, T-EHT, and P-EHT can all have eccentricity due to their inabil-ity to rotate. Therefore, ETH has an advantage over the other three systems in that the eccentricity caused local moment does not have to be considered in the design. Despite its benefit, the free-rotation system may cause damage to the surrounding concrete, causing degrada-

Fig. 4 Member forces at the connection system.

-

50

100

150

200

250

300

0 50 100 150 200 250 300 350 400 450

Eccentricity (mm)

Lo

cal

Mo

men

t(kN

-m)

P=2000kN

P=1500kN

P=1000kN

P=500kN

Local Moment

Fig. 5 Local moment at the connection joint.


tion to the whole system. Therefore, a structural per-formance evaluation must be performed to verify the overall capacity of the proposed connection systems.

3. Structural capacity of new connection systems

3.1 Test specimens In order to compare the structural capacity and failure behavior of the newly proposed connection systems as part of an overall connection member composed of con-crete girders, steel trusses, and connection systems, a real scale static loading test was performed on the six connection systems. As indicated in Table 1, these in-cluded one-T-type perfobond system with gusset plate (T-GHT), one-embedded hinge device system with a concrete haunch (EHT), two-embedded T-type per-fobond systems with or without concrete haunches (T-EHT, T-EHT-1), and two-embedded type punched con-nection plate systems with or without concrete haunches (P-EHT, P-EHT-1). As shown in Fig. 3(a), T-GHT is composed of T-type perfobond welded to a base plate connected to truss members using gusset plates. It is very similar to the stud grid connection system, as shown in Fig. 2(f); however, it has T-type perfobond instead of stud grids. The EHT system is a hinge type

connection system composed of connection plates and a connection bolt, as shown in Fig. 3(b). It was developed by focusing on the elimination of welding at construc-tion sites in order to improve construction efficiency. Both T-EHT and P-EHT are composed of T-type per-fobond and general perfobond using two connections bolts, as shown in Figs. 3(c) and 3(d), respectively. They also do not require welding either at prefabrication factories or construction sites. Utilizing the holes in T-type perfobond or general perfobond, the truss members are joined with connecting bolts. Both T-EHT-1 and P-EHT-1 use the same connection system as T-EHT and P-EHT, respectively, but without a concrete haunch. In order to examine the role of the concrete haunch in these connection systems, static loading tests were per-formed on T-EHT-1 and P-EHT-1. The test results were compared to those of T-EHT and P-EHT. 3.2 Dimensions and loading test set-up All of the specimens had the same dimensions. Table 2 and Fig. 6 indicate the dimensions of the specimen, as well as the test set up. The length, width, and height of the concrete slab were 1,600mm, 2,200mm, and 250mm, respectively, which were determined based on the di-mensions of the Shinchun Bridge and the effective width of the wheel load. Also, the height of a concrete

Table 1 Test specimens of connection systems.

Index Connection System Specimens Concrete

Haunch

T-GHT T-type perfobond

+ Gusset plate

No

EHT Embedded hinge

+ Connection bolt

Yes

T-EHT T-type perfobond

+ Connection bolts

Yes

P-EHT Perfobond

+ Connection bolts

Yes

T-EHT-1 T-type perfobond

+ Connection bolts

No

P-EHT-1 Perfobond

+ Connection bolts

No


haunch on EHT was 250mm, which represented the minimum height required to fully embed the EHT con-nection system. Otherwise, the height of a concrete haunch on T-EHT and P-EHT was 150mm, which was also the minimum height required for the concrete haunch to fully resist the local moment at a concrete girder. Additionally, the diameter, thickness, and in-clined angle of the circular truss section were 318mm, 15mm, and 60°, respectively, which were also deter-mined based on the dimensions of the Shinchun Bridge.

Table 3 presents the material properties for all speci-mens. The concrete was mixed using Ordinary Portland Cement (OPC) and coarse aggregates with a maximum size of 19mm. The expected 28 day compressive strength of the concrete was 40MPa. Further, the steel truss members were made from SM490 with an allow-able tensile strength of 190MPa and all of the rein-forcements used were SD40 with a yield strength of 400MPa. Figure 7 presents the loading system for the static structural capacity test used in this study. This

Table 2 Dimension of the connection system specimen. Structural Member Element Dimension[mm]

Length 1,600 Width 2,200 Concrete Slab

Thickness 250 Concrete Haunch Height 250

Diameter 318 Thickness 15 Truss Members

Inclined angel 60° Effective width 233

Thickness 12 Connection plate (EHT) Diameter of connection bolt 42

Number of hole 6ea Thickness 6

Diameter of hole 45

Perfobond (T-GHT, T-EHT, P-EHT,

T-EHT-1, P-EHT-1) Diameter of through re-bar 13

Table 3 Material properties of the connection system specimen. Material Type Strength[MPa] Concrete OPC 40

Steel Truss Pipe SM490 490 [360] (190) Reinforcement SD40 400

[ ]: the yield strength, ( ): the allowable strength

Fig. 6 Dimension of test specimen.

Fig. 7 Loading test system.


loading configuration was selected to simultaneously apply a tensile force on one truss member and a com-pressive force on the other truss member with horizontal forces applied to a concrete slab center.

The design ultimate load for all specimens was ap-proximately 1,000kN, and therefore, a 2,500kN force was required as a loading capacity since ultimate loads are generally twice that of yield loads. Therefore, two 2,000kN actuators (a total of 4,000kN) were used in this loading system to ensure sufficient loading capacity and to evenly balance the applied loading. As shown in Fig. 7, the two actuators were fixed to a strong wall of the laboratory and connected by a cross beam. When the pushing forces of the actuators were induced in this sys-tem, the load was applied to the specimen by way of this cross beam. The height for applying horizontal forces from the base of the laboratory was 2.0 m, which coin-cides with the center of concrete slab. To ensure a fixed condition at the ends of truss members during the test, two truss members were firmly connected by a H-type steel girder in the concrete base block, which was fixed to the strong floor of the laboratory using ten ultra strength steel rods. Also, to ensure safety during the test, the applied load was stroke-controlled with a loading rate of 0.02mm/sec.

3.3 Design of specimens The design of the specimens can be modified depending on the connection system; however, fundamental con-cepts of the connection systems focus on resisting the following four types of forces (Fig. 3): tensile and com-pressive forces of truss members, horizontal shear forces on the longitudinal concrete girder, and local moments at the connection joint. The design required that the joint was not to yield before the concrete slabs or before the steel members reached their yielding state. Therefore, the first design step involved ensuring that the two truss members at the connection joint had suffi-cient strength to prevent yielding or buckling under an ultimate or service load. At the joints of the truss mem-bers, the axial forces and the local moments always took place because the joints were not a hinge structure. Therefore, the maximum local stress _ maxtrf was calcu-lated using the following equation,

_ max ( )2

tr tr trtr

tr tr

P M df

A I= + (1)

where trP and trM were the maximum axial force and the maximum moment of truss members, respectively;

trd , trA , and trI were diameter, area, and second mo-ment of inertia of the truss member, respectively. The maximum local stress occurring at the end of truss members should have been smaller than the allowable stress of truss members. In this test, the truss members were designed to have sufficient strength to resist a de-sign ultimate load of approximately 1,000kN, predicted based on the actual forces on the Shinchun Bridge. When such a load was applied to the specimen, the ten-

sile and compressive forces in the truss members were equivalent to 1,000kN. Also, the maximum local stress _ maxtrf was always required to be smaller than the allowable tensile stress taf or the allowable compressive stress caf as proposed by the design specification (MLTM 2008). In this case, the tensile stress which oc-curred in the tensile truss member was about 80MPa, which was smaller than the allowable tensile stress taf of 190MPa. Further, the compressive stress in the com-pressive truss member was about 80MPa, which was smaller than the allowable compressive stress caf of 181.5MPa. The second design requirement was that connection devices, such as hinge connections and per-fobonds were to have sufficient structural capacity to resist tensile and horizontal shear forces. EHT has con-nection plates directly connected to truss members, so the applied loads at the joint could be directly trans-ferred to the connection plates, as well as to the truss members. Therefore, the connection plates used in EHT were mainly subjected to tensile forces of the truss members rather than horizontal shear forces. Tensile stress of the connection plates was calculated using the following equation:

( )tr tr

plateplate plate plate

P Pf

A w t= =

⋅ (2)

where plateA was the cross section area of connection palates; platew and platet were the effective width and thickness of connection plates, respectively. To make sure the ultimate strength of the connection plates in EHT was higher than that of truss members, the connec-tion plates were designed so that they would not reach the ultimate state before the truss member yielded. As such, the effective width and thickness of the connection plates in EHT were designed to be 233mm and 12mm, respectively, ensuring the tensile stress of the connec-tion plates platef would be 357.7MPa when a design ultimate load of 1,000kN was applied to the specimen. This value of 357.7MPa was approximately the yield strength of SM 490 of 360MPa, but was 4.5 folds larger than the stress of the truss member’s yield strength of 80MPa. Also, the design ultimate load of EHT ( uP ) was determined by the following equation and, in this case, was 1006.6kN.

( )u yield plate yield plate plateP f A f w t= ⋅ = ⋅ ⋅ (3)

where yieldf was the yield strength of SM490 (360MPa). Also the shear strength of a connection bolt should be evaluated to prevent the shear failure. A 42mm high tension connection bolt with a yield strength of 900MPa was selected to insure sufficient shear strength com-pared to a yield strength of connection plate equal to 357.7MPa.

The T-GHT had a T-type perfobond welded on the base plate; however, T-EHT and P-EHT, as well as T-EHT-1 and P-EHT-1, had a T-type perfobond and a gen-eral perfobond embedded in the concrete slab where


these perfobond type connection systems in the horizon-tal direction were mainly subjected to horizontal shear forces rather than tensile forces of truss members. Re-gardless of whether they were welded on the base plate or embedded in the concrete slab, perfobonds were only able to resist horizontal shear forces. The ultimate shear strength per hole in perfobond had been proposed by several researchers in the past. In this study, Eqs. (4a) and (4b), which were used in the design of the per-fobond for the Tanigawa Bridge (Yuasa et al. 2003), were used to determine ultimate shear strength. Depend-ing on whether or not perfobond had cross reinforce-ments, Eqs. (4a) or (4b) were applied, respectively.

2 1/ 2 33.38 ( / ) 39.0 10u ckQ d t d f= − × (4a)

( ){ }2 2 2 31.45 26.1 10u ck yQ d f fφ φ= − + − × (4b)

where t is the thickness of the perfobond; d andφ were the diameter of the hole in perfobond and cross rein-forcements, respectively. In these specimens, a per-fobond had six holes and six cross reinforcements, where the diameter of each hole and each cross rein-forcement were 45mm and 13mm, respectively. The ultimate strength of one perfobond, as calculated by Eq. (4b), was 1,077kN, which was similar to the design ul-timate load of 1,000kN.

The last design requirement was that the longitudinal concrete girder or a concrete haunch had to be designed to resist the local moment caused by the eccentricity between the concrete slab center to the crossing point on the axes of truss members as shown in Fig 4. With re-spect to EHT, it had no eccentricity; however, T-GHT had the eccentricity of 250mm and T-EHT, P-EHT, T-EHT-1, and P-EHT-1 all had the same eccentricity of 352mm, as shown in Fig. 8. When a design ultimate

(a) T- GHT (b) EHT

(c) T- EHT (d) P-EHT

(e) T- GHT-1 (f) P-EHT-1

Fig. 8 Eccentricity b/w center of concrete slab & axes of truss members.


load of 1,000kN was applied to the specimens, these eccentricities induced a local moment of approximately 125kN-m at the connection joint, as shown in Fig. 5. The minimum height of the concrete haunch to resist this local moment was calculated as 150mm. In this study, the minimum height of the concrete haunch was used in the design of T-EHT and P-EHT but not in that of T-EHT-1 or P-EHT-1.

3.4 Test results 3.4.1 Ultimate load and failure mode To verify the structural capacity of newly developed connection systems, the tested specimens’ ultimate loads

and the failure modes were compared and have been presented in Table 4. Figure 9 presents the failure mode for each specimen in the ultimate state. The ultimate loads for all of the specimens, except for T-EHT-1 and P-EHT-1, were larger than the design ultimate load of 1,000kN. In particular, the ultimate loads of EHT and T-GHT were approximately 119% and 150% of the design ultimate load, respectively. Additionally, the ultimate loads of T-EHT and P-EHT were approximately 121-163% of the calculated design ultimate load. In the test, when the ultimate load was reached, the radial cracks were formed on the upper surface of the slab as the shear failure of the concrete slab occurred. In contrast,

Table 4 Failure mechanism of specimens.

Specimens Ultimate Load (kN)

Resisting Member to the Local Moment Failure Mode

T-GHT 1496 Steel base plate Crack on the slab EHT 1191 Concrete haunch Crack on the haunch

T-EHT 1626 Concrete haunch Shear failure of slab P-EHT 1217 Concrete haunch Shear failure of slab

T-EHT-1 695 None Punching failure of slab P- EHT-1 654 None Punching failure of slab

(a) T-GHT (b) EHT

(c) T-EHT (d) P-EHT

(e) T-EHT-1 (f) P-EHT-1

Fig. 9 Failure mechanism of specimens.


the ultimate loads of T-EHT-1 and P-EHT-1 were ap-proximately 65-69% of the design ultimate load and displayed only brittle punching failure of the concrete slab. From these results, the local moment resisting members, such as the concrete haunch from EHT, T-EHT, and P-EHT, as well as the steel base plate from T-GHT, must be considered in the design of connection system for HTBs. Also, results indicated that a T-type perfobond had better structural capacity than a general perfobond, since the ultimate strength of T-EHT was larger than that of P-EHT. 3.4.2 Load-displacement relation In order to look more precisely into the structural be-havior of each specimen, the load-displacement rela-tions obtained from the experiment were analyzed. Fig. 10 presents the load-displacement relation of the six connection systems proposed in the current study. The figure indicates that the initial stiffness of T-EHT and P-EHT were nearly the same in the linear elastic region (from 0kN to about 1,000kN); however, the behaviors of the two specimens differed according to the connection systems when loads exceeded 1,000kN. Similarly, Fig. 10 also indicates that the initial stiffness of T-GHT and EHT were almost the same in the linear elastic region (from 0kN to about 800kN), but differed when loads exceeded 800kN. Further, it shows that the initial stiff-ness of T-EHT-1 and P-EHT-1 were nearly the same in the linear elastic region (from 0kN to about 400kN), but differed when loads exceeded 400kN. The initial stiff-ness of T-EHT and P-EHT were higher than those of T-GHT and EHT, but the initial stiffness of T-GHT and EHT were nearly the same as those of T-EHT-1 and P-EHT-1 in linear elastic region (from 0kN to about 400kN). Both T-EHT and P-EHT were able to have a larger cross section than other connection systems due to the addition of a concrete haunch. The test results demonstrate that a concrete haunch enhances the hori-zontal shear force resisting ability and increases initial stiffness. Although EHT, T-EHT, and P-EHT also had concrete haunches, the concrete haunch of EHT was discontinuously installed in the longitudinal direction and only played a role in protecting the connection sys-tem. On the other hand, the concrete haunches of T-EHT and P-EHT were continuously installed in the longitudi-nal direction as a structural member and played an es-sential role in resisting horizontal shear force and local moment. 3.4.3 Stress levels of truss members The axial tensile strain in the truss member of each specimen was measured at the middle point of truss members. Figure 11 presents measured axial tensile strains of truss members. Because this study focused on the ultimate strength and structural capacity of each connection system, truss members were designed so that they would not yield as each specimen reached an ulti-mate state. Under this design provision, when each

specimen reached the ultimate state, the measured stresses of both tensile and compressive truss members were smaller than 360MPa, the yield strength of SM490 (approximately equivalent to 1,800με). The test results indicated that higher strain occurred in EHT than the other connection systems, because the load transmission path of EHT was different than that of other specimens. In EHT, the applied load at the connection joint was directly transferred to the steel truss members without causing eccentricity between the center of slab and the crossing point of two truss axes as shown in Fig. 8. On the other hand, all specimens except for EHT had an indirect load transmission path using perfobonds with eccentricity at the connection joint as shown in Fig. 8. 3.4.4 Stress levels of connection plate and per-fobonds Figure 12 presents the measured strains of tensile and compressive connection plates in EHT. These two con-nection plates, as well as the truss members of EHT, did not yield when EHT reached the ultimate state because the measured tensile and compressive strains were smaller than 1,800με, the yield strain of SM490, which was approximately equivalent to the yield strength of 360MPa. However, the measured strain of a tensile plate reached the SM490`s yield strain at the ultimate state and reached the SM490`s allowable strain of approxi-mately 950με (equivalent to the allowable strength of 190MPa) when the applied load reached the design ul-

0

200

400

600

800

1,000

1,200

1,400

1,600

1,800

2,000

0 10 20 30 40 50 60 70 80

Lo

ad

(kN

)

Displacement (mm)

T-GHT

EHT

T-EHT

P-EHT

T-EHT-1

P-EHT-1

T-EHT-1

EHT

T-GHT T-EHT

P-EHT

P-EHT-1

Fig. 10 Load-displacement.

0

200

400

600

800

1,000

1,200

1,400

1,600

1,800

2,000

0 500 1,000 1,500 2,000 2,500 3,000

Lo

ad

(kN

)

Strain (με)

T-GHT

EHT

T-EHT

P-EHT

T-GHT T-EHT

P-EHT

EHT

Fig. 11 Measured axial tensile strain at the middle point of truss members.


timate load of 1,000kN. On the other hand, the meas-ured strain of a compressive plate was smaller than the SM490`s allowable strain of approximately 950με and SM490`s yield strain of 1,800με. Table 5 tabulates the load distribution ratio (LDR) of each connection plate, which is the ratio of the transferred force to the applied load for each connection plate. If the EHT connection system was an ideal hinge joint without a concrete slab, the applied load would have been the same as the tensile and compressive forces of the truss members. However, since the applied load was transferred to both the con-crete and connection plate, EHT is not an ideal joint. The load distribution ratio of the compressive connec-tion plate was relatively low and slightly changed from 8.0% to 10.7%, but that of the tensile connection plate was relatively high and drastically changed from 25.3% to 81.1% when the applied load increased. From this result, it is clear that the compressive force was resisted both by concrete slabs and compressive connection plates; however, the tensile force was mainly resisted by tensile connection plates since the concrete slab could only resist compressive force. Therefore, tensile connec-tion plates must be designed to have a sufficient strength to withstand the tensile forces exerted in the EHT con-nection system.

Figures 13 and 14 present the actual strains of T-type or general perfobonds and rebars, respectively, used in T-GHT, T-EHT, and P-EHT. As shown in Fig. 14, the behaviors of T-type and general perfobonds were similar to the perfectly plastic behavior where strains are very small and close to zero until reaching the ultimate state. However, strain suddenly increased after each specimen reached the ultimate state. On the other hand, the behav-ior of rebar was similar to the bilinear behavior where strain was very minimal and close to zero until the cracking load was reached. It then increased linearly until the ultimate state was reached, after which, it sud-denly increased, except for T-GHT, as shown in Fig. 14. From this result, the load resisting sequence can be di-vided into three parts: one in which the concrete resists the applied load in the initial state, the next in which the through rebar resists it after the cracking load, and the last in which the perfobond resists it after the ultimate load.

Table 5 Load distribution ratio (LDR) of connection plates. Compressive Connection Plate Tensile Connection Plate

Load [kN] Strain

[με] Stress [MPa]

Force [kN]

*LDR [%]

Strain [με]

Stress [MPa]

Force [kN]

*LDR [%]

200 -28.7 -5.7 -16.1 8.0 90.9 18.2 50.8 25.3 400 -58.4 -11.7 -32.6 8.1 227.8 45.6 127.4 31.7 600 -109.1 -21.8 -58.0 9.7 490.9 98.2 274.5 45.7 800 -141.6 -28.3 -79.2 9.8 713.9 142.8 399.2 49.5

1,000 -184.7 -36.9 -103.5 10.3 958.9 191.8 536.2 53.6 1,100 -202.9 -40.6 -113.5 10.3 1,113.9 222.8 622.9 56.6 1,190 -228.7 -45.7 -127.9 10.7 1,726.4 345.3 965.1 81.1

*LDR= Force/Load

0

200

400

600

800

1,000

1,200

1,400

1,600

1,800

2,000

-500 0 500 1,000 1,500 2,000 2,500 3,000 3,500

Lo

ad

(kN

)

Strain (με)

T-GHT

T-EHT

P-EHT

T-GHT

T-EHT

P-EHT

Strain

gauge

Fig. 13 Measured strain of the perfobond.

0

200

400

600

800

1,000

1,200

1,400

1,600

1,800

2,000

-500 0 500 1,000 1,500 2,000 2,500 3,000 3,500

Lo

ad

(kN

)

Strain (με)

T-GHT

T-EHT

P-EHT

T-EHT

T-GHT P-EHT

Strain gauge

Through

Re-bar

Fig. 14 Measured strain of the through rebar in the per-fobond.

0

200

400

600

800

1,000

1,200

1,400

-2,000 -1,500 -1,000 -500 0 500 1,000 1,500 2,000

Strain (μm)

Lo

ad

(kN

)

Compression

Tension

Compressive Connection Plate Tensile Connection Plate

Fig. 12 Measured strain of the connection plates in EHT.


4. Structural safety evaluation of new connection systems

4.1 Bridge specimen 4.1.1 Outline of bridge specimen Based on the static loading tests on newly developed connection systems, it was concluded that all connection systems (T-GHT, EHT, T-EHT, and P-EHT) had suffi-cient structural capacity as long as they included a local moment resisting member, such as a concrete haunch, to resist. However, it must be ensured that HTBs incorpo-rating these connection systems can behave in a ductile manner. Also, serviceability issues such as deflection and vibration must comply with general design limits for railway and roadway bridges.

In this study, the railway bridge specimen which used an EHT connection system was designed and con-structed in a structural laboratory in order to perform static and the dynamic loading tests. This type of con-nection system is expected to behave better than the current connection systems since an additional local moment is not generated at the connection joint from the center of concrete slab coinciding with the crossing point of two truss axes. Further, the load transmission is simpler where the applied load is directly transferred to truss members and there is no need for welding, making

construction much simpler at construction sites. Addi-tionally, the web trusses are adjustable during construc-tion since the joint is composed of a simple hinge con-nection using connection plates and bolts. Finally, the long transverse steel rod can be placed utilizing a lifting device, as well as an assembling guild to replace a con-nection bolt with a long steel rod in the transverse direc-tion, as shown in Fig. 15.

The span length, height, and width of the bridge specimen were 20.0m, 2.0m, and 3.0m, respectively, as shown in Fig. 16. The dimensions and material proper-ties of the connection system are the same as those used in the EHT connection system specimen, as shown in Tables 2 and 3. The only difference was a change from a connection bolt to a long transverse steel rod to im-prove the construction efficiency, as shown in Fig. 15. Also, in order to satisfy the required flexural capacity, three PS tendons composed of ten SWPC7B strands with 15.2mm diameters were used. Two straight tendons were internally installed in the lower slab and one ten-don was externally installed in the chevron profile where both ends were anchored at the top of the end diaphragms. The center was anchored at the lower slab, as shown in Fig. 17. Figure 18 shows the external ten-don of this bridge specimen before and after prestress-ing.

Fig. 15 EHT with a steel rod in the transverse direction.

Fig. 16 Railway bridge specimen (Span:20m).


4.1.2 Ultimate design load The ultimate design moment of the cross section of this bridge specimen nM can be calculated using the fol-lowing equation, which is commonly used to calculate the nominal moment for a conventional prestressed con-crete section.

( ) ( )2 2n p ps p s y ysa a

M A f d A f d A f d d′ ′= − + − + −⎛ ⎞⎜ ⎟⎝ ⎠

(5)

where sA and sA′ are the areas of the tensile and com-pressive reinforcements, respectively; pA is the area of the tendons; d and d ′ are the distances from the ex-treme compression fiber to the centroids of the tensile and compressive reinforcements, respectively; pd is the distance from the extreme compression fiber to the cen-troid of the tendons; and yf is the yield strength of the reinforcements. The stress of the prestressing tendon psf is given by:

1

1 p pu y y

ps pu p

ck p ck ck

f f fdf f

f d f f

γρ ρ ρ

β′= − + −

⎡ ⎡ ⎤⎤⎛ ⎞ ⎡ ⎛ ⎞ ⎛ ⎞⎤⎢ ⎢ ⎥⎥⎜ ⎟ ⎜ ⎟ ⎜ ⎟⎢ ⎥

⎝ ⎠ ⎣ ⎝ ⎠ ⎝ ⎠⎦⎣ ⎣ ⎦⎦ (6)

where ρ and ρ′ are the ratios of the tensile and com-pressive reinforcements, respectively; pρ is the ratio of the tendons; puf is the ultimate tensile strength of the tendons; ckf is the compressive strength of the con-

crete; pγ and 1β are the values determined using Table 6, which is based on the concrete structure design speci-fication in Korea (KCI 2007). The depth of equivalent compressive region a is given by:

0.85p ps s y s y

ck

A f A f A fa

f b

′+ −= (7)

where b is the width of the compressive region. Finally, it is founded that the ultimate design moment of the cross section of this bridge specimen nM can be calcu-lated as 13,041kN-m, and the ultimate design load of it

nP on the 3-point bending test can be obtained as 2,557kN using Eq. (8).

24( ) 4( ( / 8))n d n dn

M M M w LP

L L− −

= = (8)

where dM and dw is the moment and the unit weight caused by the self-weight, respectively; L is the span length of this bridge specimen. 4.2 Static loading test 4.2.1 Loading test setup Figure 19 shows the loading system for a 3-point bend-ing test performed in this study. Since the design ulti-mate load for this bridge specimen was approximately

(a) Before prestressing (b) After prestressing

Fig. 18 External tendon.

Fig. 17 Tendon profiles of the railway bridge specimen (Span: 20m).

Table 6 The value of pγ and 1β .

Index Condition Value Index Condition Value / 0.90py puf f ≥ 0.28 28ckf < MPa 0.85 / 0.85py puf f ≥ 0.40 pγ / 0.80py puf f ≥ 0.55

1β 28ckf ≥ MPa 0.85 0.007( 28)ckf− −


2,557kN, a 3,000kN hydraulic actuator was used as a loading cell. In this test setup, long loading frames with lengths of over 6.3m were used. To ensure safety during the test, these frames had sufficiently high levels of stiffness. For this test, a rectangular rigid frame was erected at center of a span, and a 3,000kN hydraulic jack and a load cell were used in the loading system. 4.2.2 Global behavior and failure mode Figure 20 presents the load-displacement curve meas-ured from the middle span (LVDT-1) and the quarter span (LVDT-2, LVDT-3) from the both ends as shown in Fig. 16. Fig. 21 presents deflection shapes according to a loading step of every 250kN. As shown in Figs. 20 and 21, it is clear that the behavior of this bridge speci-men showed a flexural behavior similar to a general prestressed concrete member. Also, the ultimate load obtained from the test was 2,588kN, which was similar to the design ultimate load of 2,557kN calculated with-out considering the deviator failure. But this bridge specimen reached the ultimate state due to an unex-pected failure of the concrete deviator block in the mid-dle span as shown in Fig. 22. However, this specimen had sufficient strength since the unexpected deviator failure occurred right after the applied load reached the ultimate load. If the concrete deviator block had not failed, the ultimate load would have been much higher than the design ultimate load of 2,557kN. 4.2.3 Neutral axis The variation in the neutral axis in the cross section ob-tained by the measured strains on the concrete surfaces

and the embedded rebars of the upper and lower slabs according to a loading step of every 250kN is shown in Fig. 23. Until the applied load reached 1,500kN, the strain in each cross-section of the upper and lower slab

Fig. 19 Static loading test set-up.

0

500

1,000

1,500

2,000

2,500

3,000

0 10 20 30 40 50 60 70 80 90Displacement (mm)

Load

(kN

)

LVDT-1

LVDT-2

LVDT-3

Fig. 20 Load-displacement relation.

0

10

20

30

40

50

60

70

80

90

-10 -5 0 5 10

250kN 500kN 750kN 1000kN 1250kN 1500kN 1750kN2000kN 2250kN 2500kN

LVDT-2 LVDT-1 LVDT-3

Fig. 21 Deflection shape according to loading steps.

Fig. 22 Failure mode of the bridge specimen.


did not reach the yield state. But starting at a load of 1,750kN, the tensile rebar in the lower slab reached yield state and the strain rate of the tensile rebar rapidly increased. However, the compressive rebar in the upper slab did not reach yielding until the ultimate load of 2,500kN was reached. Based on this result, it was clear that the cross section of this bridge specimen with an EHT connection system behaved as a general prestressed concrete section in spite of having an open truss web, thereby fully ensuring a typical flexural be-havior. 4.2.4 Stress level of truss members and con-nection plates Figures 24 and 25 present the measured strain of truss members (T1, T2 in Fig. 16) and connection plates (J1, J2 in Fig. 16), respectively. From these two figures, the stress levels of truss members and connection plates were estimated during the test. The maximum stress of the truss members was estimated to be approximately 80MPa with the maximum compressive and tensile stresses equivalent to 400με strain. This value was about 42% of the SM490`s allowable yield strength of 190MPa. In order to comply with the strict deflection limit requirements (L/1,600) for railway bridges in Ko-rea, the stiffness of this bridge specimen was designed conservatively by selecting a larger-sized truss member, thereby decreasing the stress level of the truss member. The maximum stress of the compressive connection plates was also measured to be smaller than 80MPa (equivalent to 400με strain); Also, the maximum stress of the tensile connection plates were approximately 260MPa (equivalent to 1,300με strain), which is equiva-lent to 36.8% and 72% of the SM490`s allowable yield strength of 190MPa and the ultimate strength of 360MPa, respectively. The design condition of the con-nection plates followed the philosophy of the connec-tion plates not reaching its yield state when the connec-tion system reaches its ultimate state. It was difficult to directly compare the strain levels of Figs. 24 and 25 to those of Figs. 11 and 12 in Chapter 3, since the loading condition and the boundary condition were different. The y-axis of Fig. 12 presents the horizontally applied force at the connection system, which is nearly as same as the axial forces of tensile and compressive truss members since the truss angle is approximately 60 de-gree and both ends of truss are fixed. But the y-axis of Fig. 25 presents the vertically applied load on the 3-point bending test for this bridge specimen, which is not same as the tensile and compressive axial forces (J1, J2 in Fig. 16) of truss members. And the axial force of the tensile truss member is different than that of the com-pressive truss member due to the inclined trusses in the cross section as well as the difference in boundary con-dition. A comparison of the stress level of the local con-nection system (Fig. 12) to that of the global bridge specimen (Fig. 25) showed that the stress level of the tensile connection plates was high, exceeding the

SM490`s allowable strength. However, the stress level did not reach yield state when each specimen reached its ultimate state. Finally, to ensure the safety of the tensile connection in a real bridge application, the thickness of the tensile connection plates should be increased to re-duce the stress level below that of the allowable strength. In conclusion, there is no problem in applying this EHT connection system to a real HTB since the load induced stresses are less than the allowable strength of all steel members and systems in the real bridge. 4.2.5 Stress level of tendons Figure 26 presents the stress variation of tendons of this bridge specimen. In this figure, PT-1 was the external

-1,250

-1,000

-750

-500

-250

0

250

500

750

1,000

1,250

-3,000 -2,000 -1,000 0 1,000 2,000 3,000 4,000 5,000 6,000

250kN 500kN 750kN 1000kN 1250kN 1500kN1750kN 2000kN 2250kN 2500kN

Upper Slab

Lower Slab

Reinforcement Yield Reinforcement Yield

Fig. 23 Variation of the neutral axis.

0

500

1,000

1,500

2,000

2,500

3,000

-500 -400 -300 -200 -100 0 100 200 300

Strain (με)

Load

(kN

)

Compressive Truss Member

Tensile Truss Member

Fig. 24 Measured axial strains of the truss members (T1, T2 in Fig. 16).

0

500

1,000

1,500

2,000

2,500

3,000

-2,000 -1,500 -1,000 -500 0 500 1,000 1,500 2,00

Strain (με)

Load

(kN

)

Compression

Tension

Compressive Connection Plates Tensile Connection Plates

Fig. 25 Measured strains of the connection plates (J1, J2 in Fig. 16).


tendon bending in the middle span, and PT-2 and PT-3 were the internal tendons embedded in the lower slab. In this test, in order to examine the variation of tendon stresses during the test, the load cells were installed on each tendon anchorage plate and an equivalent applied load of approximately 70% of the ultimate tensile strength of the prestressing tendon ( puf ) of 1,317MPa (equivalent to 1,280kN) was applied to the prestressing tendon. This figure showed that the stresses of three prestressing tendons did not change until the applied load was approximately 500kN. However, the stresses of three prestressing tendons gradually increased after 500kN. As the applied load increased over the yield strength of prestressing tendons ( pyf ), the stress varia-tion rate of the external tendon (PT-1) was higher than that of the internal tendons (PT-2 and PT-3). Finally, in the ultimate state, the stresses of the external (PT-1) and internal (PT-2 and PT-3) tendons were 1,791MPa and 1,696MPa, respectively.

The stress of prestressing tendon ( psf ) calculated by Eq. (6) would be valid in the ultimate state, only if it was larger than the yield strength of the prestressing tendon ( pyf ) of 1,601MPa and smaller than the ultimate tensile strength of the prestressing tendon ( puf ) of 1,882MPa. From these results, the stress of prestressing tendons ( psf ) can change depending on tendon type and installation geometry. But the design specifications in Korea (KCI 2007; KSCE, 2005; 2008) have no detail calculation methods for the stress of prestressing ten-dons ( psf ) according to the tendon type and installation geometry. On the other hand, the calculation methods for both bonded and unbounded prestressing tendon stresses were illustrated in ACI 318-08 and AASHTO LRFD Bridge Design Specification (ACI 2008; AASHTO 2007).

The stress of the bonded prestressing tendon ( psf ) in ACI specification is shown in Eq. (6) and the stress of the unbonded prestressing tendon ( psf ) in ACI is given by:

70100

ckps pe

p

ff f

ρ= + + or 400pef + , if / 35L h ≤ (9a)

70300

ckps pe

p

ff f

ρ= + + or 210pef + , if / 35L h ≥ (9b)

where pρ is the ratio of the prestressing tendons; pef is the effective stress of the prestressing tendon under con-sideration after all losses; ckf is the compressive strength of the concrete; L and h are the length and height of the prestressed concrete girder, respectively. On the other hand, the stress of the boned prestressing tendon ( psf ) in AASHTO is given by:

1 2 1.04 pyps pu

pu p

f cf ff d

⎡ ⎤⎛ ⎞⎛ ⎞= − −⎢ ⎥⎜ ⎟⎜ ⎟⎜ ⎟⎜ ⎟⎢ ⎥⎝ ⎠⎝ ⎠⎣ ⎦

(10)

where pyf and puf are the yield and ultimate strengths of the prestressing tendon, respectively; pd is the distance from extreme compression fiber to the cen-troid of the prestressing tendons. The distance between the neutral axis and the compressive face c is given by:

10.85 2 1.04

p pu s y s y

py puck p

pu p

A f A f A fc

f ff b A

f dβ

′+ −=

⎛ ⎞+ −⎜ ⎟⎜ ⎟

⎝ ⎠

(11)

where sA and sA′ are the area of the tensile and com-pressive reinforcements, respectively; pA is the area of the prestressing tendons; ckf and yf are the compres-sive strength of the concrete and the yield strengths of the reinforcements, respectively; 1β is the value deter-mined using Table 6; b is the width of the compres-sive region.

And the stress of the unbonded prestressing tendon ( psf ) in AASHTO is given by:

6300 pps pe

e

d cf f

l−⎛ ⎞

= + ⎜ ⎟⎝ ⎠

(12)

where pef is the effective stress of the prestressing tendon under consideration after all losses; pd is the distance from extreme compression fiber to the centroid of the prestressing tendons. The distance c from ex-treme compression fiber to the neutral axis assuming that the prestressing tendon yielded is given by:

10.85p ps s y s y

ck

A f A f A fc

f bβ

′+ −= (13)

where sA and sA′ are the area of the tensile and com-pressive reinforcements, respectively; pA is the area of the prestressing tendons; ckf and yf are the compres-sive strength of the concrete and the yield strength of the reinforcements, respectively; 1β is the value deter-mined using Table 6; b is the width of the compres-sive region. A first estimate of the average stress in un-bonded tendon may be made as:

0

500

1,000

1,500

2,000

2,500

3,000

1,200 1,300 1,400 1,500 1,600 1,700 1,800 1,900

Load

(kN

)

Stress Variation of Tendons (MPa)

PT1 (External Tendon) PT2 (Internal Tendon) PT3 (Internal Tendon)

fpufpy

Fig. 26 Stress variation of tendons.


103ps pef f= + (14)

Also, el is the effective tendon length is given by:

22

ie

s

ll

N⎛ ⎞

= ⎜ ⎟+⎝ ⎠ (15)

where il is the length of tendon between anchorages; sN is the number of support hinges crossed by tendon

between anchorages. Table 7 presents the measured stresses of the

prestressing tendons in this static loading test and the stresses of them calculated by Eqs. (6), (9), (10) and (12). This Table shows that the stresses of the un-bounded prestressing tendons calculated by Eq. (9) came from ACI were close to the measured stresses of the prestressing tendons. However, in order to predict the exact behavior of these bridges in the ultimate state, the calculation methods for the stress of the prestressing tendon ( psf ) should be classified and proposed as a function of tendon type and installation geometry based on the further test results obtained from testing many

prestressed concrete girders with external tendons.

4.3 Dynamic loading test The serviceability issues of railway bridges such as de-flection, vibration, and cracking capacity are very im-portant factors in a real bridge application. Therefore, the stiffness of a bridge should be designed to satisfy the deflection limit with respect to vibration, where the natural frequency of a bridge should be within a rec-ommended frequency range. In order to determine the natural frequency of the bridge specimen, a dynamic loading test using a vibration exciter was conducted, as shown in Fig. 27. Also, an eigenvalue anlaysis using a 3D FEM model was performed, as shown in Fig. 28.

Figure 29 presents the natural frequency limit of rail-way bridges with respect to the design specifications for railway bridges in Korea (MLTM 2004). It has been recommended that the natural frequency of railway bridges be between upper and lower limits according to the span length, which can be checked using UIC776-1R Appendix 102. The experimental and analytical natural frequencies of this bridge specimen were 8.69Hz and 8.5Hz, respectively, which were within the upper

Table 7 Stresses of the prestressing tendons [Unit : MPa].

Stress of the prestressing tendons ( psf ) ACI AASHTO Index

(Type)

Yield strength ( pyf )

Ultimate strength ( puf ) bond unbond bond unbond

Test results

PT-1 (External) 1,601 1,882 1,860 1,713 1,865 1,865 1,791

PT-2 (Internal) 1,601 1,882 1,860 1,723 1,865 1,877 1,710

PT-3 (Internal) 1,601 1,882 1,860 1,719 1.865 1,873 1,682

Fig. 27 Dynamic test set-up and vibration exciter.

Fig. 28 3D FEM model.


and lower limits, as shown in Fig. 29. Therefore, it can be safe to conclude that this bridge specimen would be safe from vibration loading if HTB is used for a railway bridge, HTB vibration resistant capacity should be fur-ther checked using 3D FEM analysis.

5. Conclusions

The following conclusions can be drawn based on the investigation of the paper: 1. The performance of four new connection systems

(T-GHT, EHT, T-EHT, P-EHT) using hinge devices or T-type perfobonds were evaluated in order to improve assemblage convenience, to maximize construction efficiency, and to eliminate welding during construction. The evaluation procedure in-cluded design, fabrication, and static loading tests for all connection specimens. Results demonstrated that all specimens fully satisfied structural safety requirements.

2. The static loading tests of the newly developed connection specimens in real scale showed that their structural capacities were excellent. However, the local moment resisting members at the connec-tion joints such as concrete haunches (EHT, T-EHT and P-EHT) and a steel base plates (T-GHT) re-quire careful selection. The use of concrete haunches may enhance the local moment resisting ability and increase the initial stiffness. Further, T-type perfobond showed better structural capacity than general perfobond.

3. With respect to EHT, the applied load at the con-nection joint was directly transferred to the steel truss members, eliminating eccentricity due to dis-cordance between the center of the slab and the cross point of two truss axes. All specimens except for EHT had an indirect load transmission path us-ing perfobonds and an eccentricity at the connec-tion joint. Therefore, the local moment caused by this eccentricity should be considered in the joint design.

4. With respect to EHT, compressive stress was re-sisted both by concrete slabs and compressive con-

nection plates; however, the tensile stress was mainly resisted by tensile connection plates since the concrete slabs could only resist a compressive stress. Therefore, tensile connection plates should be designed to have sufficient strength to withstand tensile stresses.

5. The tests results revealed that the load resisting sequence for T-GHT, T-EHT, and P-EHT was as follows. Concrete resists the applied load in the initial state, which was followed by rebar resisting the load after the cracking, and finally by per-fobond resisting the load after the ultimate state.

6. The structural safety of the real scale 20m railway bridge specimen with an EHT connection system was evaluated by static and dynamic loading tests. From the static loading test, the cross-section of this bridge specimen with an EHT connection sys-tem behaved as a general prestressed concrete sec-tion, despite having an open truss web, thereby in-ducing typical flexural behavior. Dynamic loading verification for this bridge resulted in experimental and analytical natural frequencies of 8.69Hz and 8.5Hz, respectively. Since these values complied with the upper and lower limits recommended by the design specification in Korea, it was concluded that this bridge could be used as a railway bridge.

Acknowledgements This study were supported by a research pro-ject(10ADIT07) and grant (05 Construction Core C14) from the Construction Core Technology Program by the Ministry of Land, Transport, and Maritime Affairs of the Korean government.. The authors wish to express their gratitude for this financial support. The corresponding and second author of this paper would like to thank En-gineering Research Center (ERC) of the National Re-search Foundation of Korea (NRF) grant funded by the Korea government (MEST) for partial financial support (No. 2011-0030846). The opinions, findings, and con-clusions of the paper are the authors’ and do not neces-sarily reflect the views of the sponsors. References AASHTO, (2007). “AASHTO LRFD bridge design

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