New Zealand Society for Earthquake

New Zealand SocietyFor Earthquake

Engineering

Assessment and Improvementof the Structural Performance

of Buildings in Earthquakes

Section 10 RevisionSeismic Assessment of

Unreinforced Masonry Buildings

Recommendations of a NZSEE Project Technical GroupIn collaboration with SESOC and NZGS

Supported by MBIE and EQCJune 2006

Issued as part of Corrigendum No. 4ISBN 978-0-473-32278-6 (pdf version)

This document represents the current status of a review of the original Section 10 of the NZSEE Guidelines which was published in 2006. Any comments will be gratefully received. Please forward any comments to NZSEE Executive Officer at [email protected]. April 2015

SeUpd

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ection 10 - Seismdated 22 April 2015

Section Maso

10.1 Ge

10.

10.

10.

10.

10.

10.

10.2 Typ

10.

10.

10.

10.

10.

10.

10.

10.

10.

10.

10.

10.

10.

10.

10.

10.3 Ob

10.

10.

10.

10.

10.

10.

10.

10.

10.

10.4 Fac

10.

10.

10.5 As

10.

10.

10.

10.

10.6 On

10.

10.

10.

mic Assessmen

10 - Seinry Bui

eneral ..........

.1.1 Backg

.1.2 Scope

.1.3 Basis

.1.4 How t

.1.5 Notat

.1.6 Defin

pical URM B

.2.1 Gene

.2.2 Buildi

.2.3 Found

.2.4 Wall c

.2.5 Const

.2.6 Floor/

.2.7 Diaph

.2.8 Wall t

.2.9 Damp

.2.10 Built-i

.2.11 Bond

.2.12 Bed-j

.2.13 Lintel

.2.14 Secon

.2.15 Seism

bserved Seis

.3.1 Gene

.3.2 Buildi

.3.3 Diaph

.3.4 Conn

.3.5 Walls

.3.6 Walls

.3.7 Secon

.3.8 Poun

.3.9 Found

ctors Affect

.4.1 Numb

.4.2 Other

sessment A

.5.1 Gene

.5.2 Asses

.5.3 Asses

.5.4 Asses

n-site Invest

.6.1 Gene

.6.2 Form

.6.3 Diaph

t of Unreinforce

ismic Aildings .

....................

ground .........

e ..................

s of this secti

to use this se

tion ..............

itions ...........

Building Pra

eral ...............

ng forms .....

dations ........

construction .

tituent mater

/roof diaphra

hragm seatin

to wall conne

p-proof cours

in timber ......

beams ........

oint reinforce

s ..................

ndary compo

mic strengthe

smic Behav

eral ...............

ng configura

hragms .........

ections ........

s subjected to

s subjected to

ndary compo

ding .............

dations and

ting Seismic

ber of cycles

r key factors .

Approach ....

eral ...............

ssment proce

ssment of str

ssment of row

igations ......

eral ...............

and configu

hragm and co

ed Masonry Buil

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on ...............

ection ..........

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actices in Ne

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rials .............

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ng and conne

ections.........

se (DPC) .....

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ement ..........

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onents .........

ening method

iour of URM

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ation .............

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o face loads

o in-plane loa

onents/eleme

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geotechnica

c Performan

and duration

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ess ..............

rengthened b

w buildings .

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ldings

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ew Zealand .

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10-i473-26634-9

. 10-1

....... 10-1

....... 10-1

....... 10-2

....... 10-3

....... 10-4

....... 10-4

..... 10-11

..... 10-14

..... 10-14

..... 10-14

..... 10-18

..... 10-19

..... 10-25

..... 10-25

..... 10-28

..... 10-30

..... 10-30

..... 10-31

..... 10-32

..... 10-33

..... 10-34

..... 10-34

..... 10-35

..... 10-39

..... 10-39

..... 10-41

..... 10-42

..... 10-42

..... 10-45

..... 10-48

..... 10-52

..... 10-53

..... 10-54

..... 10-54

..... 10-54

..... 10-55

..... 10-58

..... 10-58

..... 10-60

..... 10-64

..... 10-66

..... 10-68

..... 10-68

..... 10-68

..... 10-68

s

i 9

Section 10 - SeUpdated 22 April 20

1

1

1

1

1

1

1

1

10.7 M

1

1

1

1

1

1

1

1

10.8 A

1

1

1

1

1

1

1

10.9 A

1

1

1

1

10.10 A

1

1

1

1

1

1

1

1

10.11 A

10.12 I

Refere

Sugge

eismic Assessm015

10.6.4 Loa

10.6.5 No

10.6.6 Co

10.6.7 Fou

10.6.8 Ge

10.6.9 Se

10.6.10 Se

10.6.11 Pre

Material Pro

10.7.1 Ge

10.7.2 Cla

10.7.3 Co

10.7.4 Dir

10.7.5 Dia

10.7.6 Mo

10.7.7 Tim

10.7.8 Ma

Assessmen

10.8.1 Ge

10.8.2 Str

10.8.3 Dia

10.8.4 Co

10.8.5 Wa

10.8.6 Wa

10.8.7 Oth

Assessmen

10.9.1 Ge

10.9.2 Glo

10.9.3 Glo

10.9.4 Glo

Assessmen

10.10.1 Ge

10.10.2 Pri

10.10.3 Pa

10.10.4 Ve

10.10.5 Fle

10.10.6 Rig

10.10.7 Co

10.10.8 Co

Assessmen

Improving S

ences ...........

ested Refere

ment of Unreinfo

ad-bearing w

on load-beari

oncrete .........

undations ....

eotechnical a

condary elem

ismic separa

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operties and

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ay bricks and

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rect tensile st

agonal tensile

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urther Readi

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asonry .........

f masonry ....

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ISBN 978-

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Masonry Buildi

1-0-473-26634

......... 10-69

......... 10-70

......... 10-70

......... 10-70

......... 10-70

......... 10-70

......... 10-71

......... 10-71

.......... 10-77

......... 10-77

......... 10-77

......... 10-78

......... 10-79

......... 10-79

......... 10-79

......... 10-79

......... 10-79

.......... 10-80

......... 10-80

......... 10-80

......... 10-80

......... 10-86

......... 10-89

....... 10-104

....... 10-119

........ 10-120

....... 10-120

....... 10-123

....... 10-125

....... 10-125

........ 10-128

....... 10-128

....... 10-128

....... 10-130

....... 10-130

....... 10-130

....... 10-132

....... 10-132

....... 10-132

........ 10-132

........ 10-133

........ 10-133

........ 10-137

ngs

0-ii 4-9

Seismic Assessment of Unreinforced Masonry Buildings

Section 10 - Seismic Assessment of Unreinforced Masonry Buildings 10-iii Updated 22 April 2015 ISBN 978-0-473-26634-9

Appendix 10A: On-site Testing

Appendix 10B: Derivation of Instability Deflection and Fundamental Period for Masonry Buildings

Appendix 10C: Charts for Assessment of Out-of-Plane Walls

SeUpd

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1

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ThIm20onAinapin Uoffechdipe Amtoatidto Thre

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he seismic esult in marg

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mic Assessmen

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panning vergly non-line02) using nuded testing

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Eqs 10B.5, 1

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Section 10 - Seismic Assessment of Unreinforced Masonry Buildings 10-5

Updated 22 April 2015 ISBN 978-0-473-26634-9

Symbol Meaning Comments

bsp Width of spandrel Eqs 10.35, 10.37, 10.39, 10.40, 10.41, 10.42, 10.43, 10.46, 10.47, 10.48, 10.49

c Masonry bed-joint cohesion, N/mm2 The ability of the mortar to work in conjunction with the bricks.

This is related to moisture absorption in the bricks. It depends less on the absorption qualities of individual brick types and is not greatly influenced by keying of the brick surface (e.g. holes, lattices or patterning).

Cohesion is relevant to the primary decision of whether to use cracked or un-cracked masonry properties for the analyses.

Eqs 10.3, 10.33, 10.39, 10.47 10.36

c Probable cohesion, MPa Table 10.4

C(0) Section 10.10.5.1, Figure 10.78

C(T1) Elastic site hazard spectrum for horizontal loading

Eq 10.53

C(Td) Seismic coefficient at required height at period Td

Eq 10.54

Ch(0) Spectral shape factor for relevant soil determined from Clause 3.1.1, NZS 1170.5, g

Appendix 10C

Ch(T1) Spectral shape factor for relevant site subsoil type and period T1 as determined from Section 3, NZS 1170.5, g

Eq 10.53

Chc(Tp) Spectral shape factor for site subsoil type C and period Tp as determined from Section 3, NZS 1170.5, g

Eq 10.16

CHi Floor height coefficient for level i as defined in NZS 1170.5

Appendix 10C

Ci(Tp) Eq 10.16, Section 10.10.3

Cm Value of the seismic coefficient that would cause a mechanism to just form, g

Uniform acceleration to the entire panel is assumed in finding Cm

Eq 10.21, Section 10.8.5.2 Step 15 Note, Table 10.12, 10.27, 10B.20

Cp(0.75) Seismic coefficient for parts at 0.75 sec. Value of the seismic coefficient that would cause a mechanism to just form, g

Section 10.8.5.2 Step 13 & Step 15 Note

Cp(Tp) Design response coefficient for parts as defined by Section 8, NZS 1170.5, g

Section 18.8.5.2 Step 8, Eqs 10.18, 10.19, Section 10.8.5.2 Step 13

CSW Critical structural weakness

D Table 10.14

Dph Displacement response (demand) for a wall panel subject to an earthquake shaking as specified by Equation 10.18, mm

Eqs 10.18, 10.20

e Table 10.14





Em Young’s modulus of masonry, MPa, kN/m2 Eqs 10.4, 10.8, 10.51, 10.52

eb Eccentricity of the pivot at the bottom of the panel measured from the centroid of Wb, mm

Eq 10.12, Table 10.12, 10.15, Section 10.8.5.2 Step 4, Figure 10B.1

eo Eccentricity of the mid-height pivot measured from the centroid of Wb, mm

Eq 10.12, 10.15, Section 10.8.5.2 Step 4, Figure 10B.1

ep Eccentricity of P measured from the centroid of Wt, mm

Eq 10.12, 10.15, Table 10.12, Section 10.8.5.2 Step 4, Figure 10B.1

et Eccentricity of the mid-height pivot measured from the centroid of Wt, mm

Eqs 10.12, 10.15, Section 10.8.5.2 Step 4, Figure 10B.1

F Applied load on timber lintel Eq 10.42

Fi Equivalent static horizontal force at the level of the diaphragm (level i)

Section 10.10.5.1, Figure 10.78

f’b Compressive strength of bricks, N/mm2 Measured on the flat side

Section 10A.3.2

f’b Probable brick compressive strength, MPa Table 10.3, 10.5, Eqs 10.1, 10.2

f’j Normalised mortar compressive strength, N/mm2

Eq 10A.1

f’j Probable mortar compressive strength, MPa

Table 10.4, Eq 10.2, Table 10.5

f’ji Measured irregular mortar compressive strength, MPa

Eq 10A.1

f’m Compressive strength of masonry, MPa Eq 10.31

f’m Probable masonry compressive strength, MPa

Eq 10.2, Table 10.5, Eq 10.4

f'r Modulus of rupture of bricks, MPa Eq 10.1, Section 10.8.5.1

f't Equivalent tensile strength of masonry spandrel, MPa

Eqs 10.9, 10.35, 10.36

fa Axial compression stress on masonry due to gravity load, MPa

Eqs 10.3, 10.30, 10.31

fbt Probable brick tensile strength, MPa May be taken as 85% of the stress derived from splitting tests or as 50% of stress derived from bending tests

Table 10.3

fdt Diagonal tensile strength of masonry, MPa Eqs 10.3, 10.30, 10.38, 10.40, 10.48

fhm Compression strength of the masonry in the horizontal direction (0.5f’m), MPa

Eqs 10.37, 10.46

g Acceleration due to gravity, m/sec2 Eqs 10.15, 10.17, 10.18, 10.29

G’d Reduced diaphragm shear stiffness, kN/m Eqs 10.6, 10.7, 10.8

G’d,eff Effective diaphragm shear stiffness, kN/m Eqs 10.7, 10.54

Gd Shear stiffness of straight sheathed diaphragm, kN/m

Table 10.8, Eq 10.6

Gm Shear modulus of masonry, MPa Eqs 10.5, 10.51, 10.52





h Free height of a cantilever wall from its point of restraint or height of wall in between restraints in case of simply-supported face-loaded wall

The clear height can be taken at the centre-to-centre height between lines of horizontal restraint. In the case of concrete floors, the clear distance between floors will apply.

Eqs 10.13, 10.17

hi Average of the heights of point of support Section 10.8.5.2 Step 8

hi Height of attachment of the part Figure 10.78

Hl Height of wall below diaphragm, m Eq 10.8

heff Height of wall or pier between resultant forces

Table 10.13, Eqs 10.31, 10.32, Figures 10.65, 10.74, Table 10.14, Eq 10.51

hi Average of the heights of points of support Section 10.8.5.2 Step 8

hn Figure 10.78

hsp Height of spandrel excluding depth of timber lintel if present

Eqs 10.35, 10.37, 10.38, 10.39, 10.40, 10.41, 10.42, 10.43, 10.46, 10.47, 10.48, 10.49

htot Eq 10.46, Figure 10.69

Hu Heigh of wlall above diaphragm, m Eq 10.8

Ig Moment of inertia for the gross section representing uncracked behaviour

Eq 10.51

Ixx Mass moment of inertia about x-x axis, kgm2

Eq 10B.9

Iyy Mass moment of inertia about y-y axis, kgm2

Eq 10B.10

J Rotational inertia of the wall panel and attached masses, kgm2

Eqs 10.14, 10.15, 10.17, Table 10.12, Eqs 10.29, 10B.8, 10B.30

Janc Rotational inertia of ancillary masses, kgm2

Eqs 10.15, 10B.8

Jbo Rotational inertia of the bottom part of the panel about its centroid, kgm2

Eqs 10.15, 10B.11

Jbo Polar moment of inertia about centroid, kgm2

Eqs 10B.8, 10B.11

Jto Rotational inertia of the top part of the panel about its centroid

Eqs 10.15, 10B.8

k In-plane stiffness of walls and piers, N/mm Eqs 10.51, 10.52

KA Section 10.9.2

KR Seismic force reduction factor for in-plane seismic force

Coefficient proposed in lieu of Sp and K

Eq 10.53, Table 10.15

L Span of diaphragm, m Eq 10.8, 10.54, 10.55

l Length of header Section 10.8.5.1

lsp Clear length of spandrel between adjacent wall piers

Eqs 10.35, 10.37, 10.38, 10.40, 10.41, 10.42, 10.43, 10.44, 10.46, 10.48





Lw Length of wall Eqs 10.31, 10.33

M Moment capacity of the panel Eq 10.9

M.F Eq 10A.4

M1, Mi, Mn Moment imposed on wall/pier elements Figure 10.72

m Mass, kg Eq 10B.11

mi Seismic mass at the level of the diaphragm (level i)

Section 10.10.5.1, Figure 10.78

N(T1,D) Near fault factor determined from Clause 3.1.6, NZS 1170.5

Eq 10.53

n Number of recesses Eq 10.10

N1, Ni, Nn Axial loads on pier elements Figure 10.72

P Superimposed and dead load at top of wall/pier

Eqs 10.31, 10.32, 10.33, 10.34

P Load applied to the top of panel P is assumed to act through the pivot at the top of the wall

Section 10.8.5.1, 10.8.5.2 Step 2, 3 & 4, Eqs 10.9, 10.12, 10.13, 10.15, 10.28

p Depth of mortar recess, mm Eq 10.10, Figure 10.62

P-∆ P- delta Section 10.8

pp Mean axial stress due to superimposed and dead load in the adjacent wall piers

Eq 10.36

psp Axial stress in the spandrel Eqs 10.35, 10.37, 10.38, 10.39, 10.40, 10.41, 10.42, 10.43, 10.46, 10.47, 10.48, 10.49

Pw Self-weight of wall and pier Eqs 10.31, 10.32, 10.33, Figure 10.65, EQ 10.34

Q Live load Section 10.10.5.2

Q1, Qi, Qn Shear in pier element Figure 10.72

R Return period factor, Ru determined from Clause 3.1.5, NZS 1170.5

Eq 10.16

ra Eq 10.44, Figure 10.69

ri Eqs 10.44, 10.45, Figure 10.69

ro Eq 10.45, Figure 10.69

RP Risk factor for parts as defined in NZS 1170.5

Eq 10.18, Section 10.8.5.2 Step 13

Ru Return period factor for ultimate limit state as defined in NZS 1170.5

Eq 10.53

Si Sway potential index Eq 10.50

Sp Structural performance factor in accordance with NZS 1170.5

Section 10.8.5.2, 10.10.2.1, 10.10.5.1, 10.10.8

SW Structural weakness





t Depth of header Section 10.8.5.1

t Effective thickness, mm Varies with position

Section 10.8.5.2 Step 2, 3 & 4, Eq 10B.22

T1 Fundamental period of the building, sec. Eq 10.53

Td Fundamental period of diaphragm, sec. Eqs 10.54, 10.55

tgross Overall thickness of wall, mm Varies with position

Eq 10.10

tl Effective thickness of walls below the diaphragm, m

Eq 10.8, Figure 10.59

tnom Nominal thickness of wall excluding pointing, mm

Varies with position

Eqs 10.9, 10.10, Section 10.8.5.2 Step 2 & 3, Eqs 10.33, 10B.22

Tp Effective period of parts, sec. Eqs 10.14, 10.16, 10.18, 10.19, 10.23, 10.24, 10B.15, 10B.16, 10B.17, 10B.18, 10B.24, 10B.25, 10B.31

tu Effective thickness of walls above the diaphragm, m

Eq 10.8, Figure 10.59

V Probable shear strength capacity

Vdpc Capacity of a slip plane for no slip Eq 10.34

Vdt In-plane diagonal tensile strength capacity of pier and wall

Eq 10.30

Vfl Peak flexural capacity of spandrel Figure 10.68, Eqs 10.35, 10.43

Vfl,r Residual flexural strength capacity Figure 10.68, Eqs 10.37, 10.46

(Vprob)global, base Figure 10.75

(Vprob)line, i Section 10.9.2

(Vprob)wall1,wall2 Figure 10.77

Vr In-plane rocking strength capacity of pier and wall

Eq 10.32, Figure 10.66

Vs In-plane bed-joint shear strength capacity of pier and wall

Eq 10.33, Figures 10.66, 10.68

Vs1 Eqs 10.39, 10.47

Vs2 Eqs 10.40, 10.48

Vs,r Residual spandrel shear strength capacity or residual wall sliding shear strength capacity

Eq 10.33, Figure 10.68, Eqs 10.41, 10.49

Vtc In-plane toe-crushing strength capacity of pier and wall

Eq 10.31

Vtc,r Figure 10.66

W Weight of the wall and pier Section 10.8.5.2 Step 3, Eqs 10.28, 10B.11

Wb Weight of the bottom part of the panel Section 10.8.5.1, 10.8.5.2 Step 2 & 14, Eqs 10.12, 10.13, 10.15, 10.17, 10.21





Wt Weight of the top part of the panel Section 10.8.5.1, 10.8.5.2 Step 2 & 14, Eqs 10.12, 10.13, 10.15, 10.17, 10.21

Wtrib Uniformly distributed tributary weight Eqs 10.54, 10.55

yb Height of the centroid of Wb from the pivot at the bottom of the panel

Eqs 10.12, 10.13, 10.15, 10.17, 10.21, Sections 10B.2.6, 10B.2.7, 10B.2.8, 10B.3.2, 10B.3.3

yt Height from the centroid of Wt to the pivot at the top of the panel

Eqs 10.12, 10.13, 10.15, 10.17, 10.21

Z Hazard factor as defined in NZS 1170.5 Eq 10.53

αa Arch half angle of embrace Eqs 10.43, 10.44, 10.45, 10.47, 10.48, 10.49

αht t/l ratio correction factor Eqs 10A.1, 10A.3

αtl t/l ratio correction factor Eqs 10A.1, 10A.2

αw Diaphragm stiffness modification factor taking into account boundary walls

Eqs 10.7, 10.8

β Factor to correct nonlinear stress distribution

Eq 10.30, Table 10.13

β1 Section 10.9.2

βs Spandrel aspect ratio Eq 10.38

Participation factor for rocking system This factor relates the deflection at the mid-height hinge to that obtained from the spectrum for a simple oscillator of the same effective period and damping

Eqs 10.17, 10.18, 10.25, 10B.21, 10B.32

Horizontal displacement, mm Eq 10B.16

d Horizontal displacement of diaphragm Eq 10.54

i Deflection that would cause instability of a face-loaded wall

Wb, Wt and P are the only forces applying for this calculation

Eqs 10.11, Table 10.12, Eqs 10B.6, 10B.16, 10B.30, Section 18.8.5.2 Step 6, Eq 10.20

m An assumed maximum useful deflection = 0.6 ∆i and 0.3∆i for simply-supported and cantilever walls respectively

Used for calculating deflection response capacity

Section 18.8.5.2 Step 6, Eqs 10.20, 10B.6

t An assumed maximum useful deflection = 0.6Δm and 0.8Δm for simply-supported and cantilever walls respectively

Used for calculating fundamental period of face-loaded rocking wall

Eq 10B.16

tc.r Deformation at the onset of toe crushing Section 10.8.6.2, Figure 10.66

y Yield displacement Section 10.8.6.2, Figure 10.66

y Yield rotation fo the spandrel Figure 10.68, Table 10.14

Structural ductility factor in accordance with NZS 1170.5

Sections 10.8.5.2, 10.10.2.1, 10.10.5.1, 10.10.8

dpc DPC coefficient of friction Eq 10.34

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Course A course refers to a row of units stacked on top of one another.

Cross wall An interior wall that extends from the floor to the underside of the floor above or to the ceiling, securely fastened to each and capable of resisting lateral forces.

Dead load The weight of the building materials that make up a building, including its structure, enclosure and architectural finishes. The dead load is supported by the structure (walls, floors and roof).

Diaphragm A horizontal structural element (usually suspended floor or ceiling or a braced roof structure) that is strongly connected to the walls around it and distributes earthquake lateral forces to vertical elements, such as walls, of the lateral force resisting system. Diaphragms can be classified as flexible or rigid.

Dimension When used alone to describe masonry units, means nominal dimension.

Ductility The ability of a structure to sustain its load-carrying capacity and dissipate energy when it is subjected to cyclic inelastic displacements during an earthquake.

Earthquake-Prone Building (EQP)

A legally defined category which describes a building that has been assessed as likely to have its ultimate limit state capacity exceeded in moderate earthquake shaking (which is defined in the regulations as being one third of the size of the shaking that a new building would be designed for on that site). A building having seismic capacity less than 34%NBS.

Earthquake Risk Building (ERB)

A building having seismic capacity less than 67%NBS.

Face-loaded walls Walls subjected to out-of-plane shaking. Also see Out-of-plane load.

Flexible diaphragm A diaphragm which for practical purposes is considered so flexible that it is unable to transfer the earthquake loads to shear walls even if the floors/roof are well connected to the walls. Floors and roofs constructed of timber, steel, or precast concrete without reinforced concrete topping fall in this category.

Gravity load The load applied in a vertical direction, including the weight of building materials (dead load), environmental loads such as snow, and moveable building contents (live load).

Gross area The total cross-sectional area of a section through an element bounded by its external perimeter faces without reduction for the area of cells and re-entrant spaces.

In-plane load Seismic load acting along the wall length.

In-plane walls Walls loaded along its length. Also referred as in-plane loaded wall.

Irregular building A building that has a sudden change in the shape of plan is considered to have a horizontal irregularity. A building that changes shape up its height (such as setbacks or overhangs) or is missing significant load-bearing walls is considered to have a vertical irregularity. In general, irregular buildings do not perform as well as regular buildings perform in earthquakes.

Lateral load Load acting in the horizontal direction, which can be due to wind or earthquake effects.

Leaf See Wythe.

Load path A path through which vertical or seismic forces travel from the point of their origin to the foundation and, ultimately, to the supporting soil.

Load See Action.

Low-strength masonry Masonry laid in weak mortar; such as weak cement/sand or lime/sand mortar.

Masonry unit A preformed component intended for use in masonry construction.

Masonry Any construction in units of clay, stone or concrete laid to a bond and joined together with mortar.




Mortar The cement/lime/sand mix in which masonry units are bedded.

Mullion A vertical member, as of stone or wood, between the lights of a window, the panels in wainscoting, or the like.

Net area The gross cross-sectional area of the wall less the area of un-grouted areas or penetrations.

Out-of-plane load Seismic load (earthquake shaking) acting normally (perpendicular) or at right angles to the wall surface. Walls subjected to out-of-plane shaking are also known as face-loaded walls. Walls are weaker and less stable under out-of-plane than under in-plane seismic loads.

Partition A non-load-bearing wall which is separated so as not to be part of the seismic resisting structure.

Party wall A party wall (occasionally party-wall or parting wall) is a dividing partition between two adjoining buildings or units that is shared by the tenants of each residence or business.

Pier A portion of wall between doors, windows or similar structures.

Pointing (masonry) Troweling mortar into a masonry joint after the masonry units have been laid. Higher quality mortar is used than for the brickwork.

Primary element An element which is relied on as part of the seismic force resisting system.

Regular building see Irregular building.

Return wall A short wall usually perpendicular to, and at the end of, a freestanding wall to increase its structural stability.

Rigid diaphragm A suspended floor, roof or ceiling structure that is able to transfer lateral loads to the walls with negligible horizontal deformation of the diaphragm. Floors or roofs made from reinforced concrete, such as reinforced concrete slabs, fall into this category.

Running or stretcher bond

The unit set out when the units of each course overlap the units in the preceding course by between 25% and 75% of the length of the units.

Seismic hazard The potential for damage caused by earthquakes. The level of hazard depends on the magnitude of probable earthquakes, the type of fault, the distance from faults associated with those earthquakes, and the type of soil at the site.

Seismic system That portion of the structure which is considered to provide the earthquake resistance to the entire structure.

Shear wall A wall which is subjected to lateral loads due to wind or earthquakes acting parallel to the direction of an earthquake load being considered (also known as an in-plane wall). Walls are stronger and stiffer in plane than out of plane.

Special study A procedure for justifying a departure from these guidelines or determining information not covered by them. Special studies are outside the scope of these guidelines.

Stack bond The unit set out when the units of each course do not overlap the units of the preceding course by the amount specified for running or stretcher bond.

Strength, design The nominal strength multiplied by the appropriate strength reduction factor.

Strength, probable The theoretical strength of a component section, calculated using the section dimensions as detailed and the theoretical characteristic material strengths as defined in these guidelines.

Strength, required The strength of a component section required to resist combinations of actions for ultimate limit states as specified in AS/NZS 1170 Part 0.

Structural element Component of a building that provides gravity and lateral load resistance and is part of a continuous load path. Walls are key structural elements in all masonry buildings.


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(a) Common bond (b) Running bond

(c) English bond (d) Flemish bond

Figure 10.6: Different types of brick masonry bonds

Other bond patterns used in New Zealand include Running bond (Figure 10.6(b)) and Flemish bond (Figure 10.6(d)). Running bond (stretcher courses only) often indicates the presence of a cavity wall. Flemish bond (alternating headers and stretchers in every course) is the least common bond pattern and is generally found between openings on an upper storey; for example, on piers between windows.

Stone masonry

Stone masonry buildings in New Zealand are mainly built with igneous rocks such as basalt and scoria, or sedimentary rocks such as limestone. Greywacke, which is closely related to schist, is also used in some parts of the country. Trachyte, dolerite, and combinations of these are also used.

Wall texture

Wall texture describes the disposition of the stone courses and vertical joints. There are three different categories (Figure 10.7): ashlar (squared stone); rubble (broken stone); and cobble stones (field stone), which is less common.

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It is usually not possible to establish the cross section characteristics of a stone masonry wall from the bond pattern. More detailed inspection is required to identify any connections between the wythes; determine what material the core is composed of; and locate any voids, a cavity, or the presence of other elements such as steel ties. All of these contribute to determining the wall’s structural properties.

(a) Dressed stone in outer

leaves and “rubble” fill (b) Stone facing and brickwork backing

(c) Stone facing and concrete core

Figure 10.9: Stone masonry cross sections in New Zealand. Representative cases observed in Christchurch after the Canterbury earthquakes (Marta Giaretton)

Concrete block masonry

Although solid concrete masonry was used in New Zealand from the 1880s, hollow concrete block masonry was not used widely until the late 1950s. Masonry was usually constructed in running bond, but stacked bond was sometimes used for architectural effect. From the 1960s onwards, masonry was usually constructed with one wythe 190 mm thick, although this was sometimes 140 mm thick. Cavity construction, involving two wythes with a cavity between, was mostly used for residential or commercial office construction but occasionally for industrial buildings. The external wythe was usually 90 mm thick and the interior wythe was either 90 mm or 140 mm. Cavity construction was often used for infills, with a bounding frame of either concrete or encased steelwork. To begin with, reinforcement in concrete masonry was usually quite sparse, with vertical bars tending to be placed at window and door openings and wall ends, corners and intersections, and horizontal bars at sill and heads and the tops of walls or at floor levels. Early on, it was common to fill just the reinforced cells. Later, when the depressed web open-ended bond beam blocks became more available, more closely spaced vertical reinforcement became more practicable. When the depressed web open-ended bond beam blocks (style 20.16) became available without excessive distortion from drying shrinkage, these tended to replace the standard hollow blocks for construction of the whole wall (with specials at ends, lintels and the like). Wire reinforcement formed into a ladder structure (“Bloklok” or a similar proprietary product) was common in cavity construction. Two wires ran in the mortar in bed joints, joined across the cavity by another wire at regular centres and acting as cavity ties.

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diaphragm strengthening: plywood overlay floor or roof sarking, plywood ceiling, plywood/light gauge steel composite, plasterboard ceiling, thin concrete overlay/topping, elastic cross bracing, semi-ductile cross bracing (e.g. Proving ring), replacement floor over/below with new diaphragm

in-plane wall strengthening/ new primary strengthening elements: sprayed concrete overlay, vertical post-tensioning, internal horizontal reinforcement or external horizontal post-tensioning, bed-joint reinforcement, composite reinforced concrete boundary or local reinforcement elements, composite FRP boundary or local reinforcement elements, nominally ductile concrete walls or punched wall/frame or reinforced concrete masonry walls, nominally ductile steel concentric or cross bracing, limited ductility steel moment frame or concrete frame or concrete walls or timber walls, ductile eccentrically braced frame/K-frames, ductile concrete coupled or rocking walls, or tie to new adjacent (new) structure

reinforcement at wall intersections in plan: removal and rebuilding of bricks with inter-bonding, bed-joint ties, drilled and grouted ties, metalwork reinforcing internal corner, grouting of crack

foundation strengthening: mass underpinning, grout injection, concentric/balanced re-piling, eccentric re-piling with foundation beams, mini piling/ground anchors

façade wythe ties: helical steel mechanical engagement – small diameter, steel mechanical engagement – medium diameter, epoxied steel rods/gauze sleeve, epoxied composite/non-metallic rods, brick header strengthening

canopies: reinforce or recast existing hanger embedment, new steel/cast iron posts, new cantilevered beams, deck reinforcement to mitigate overhead hazard, conversion to accessible balcony, base isolation.

Figures 10.23 to 10.27 illustrate some of these techniques. A detailed table (Table 10.2) is included in Section 10.6.11. This table lists common strengthening techniques and particular features or issues to check for each method.




Figure 10.23: Bracing of wall against face load (Dunning Thornton)

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Failures of URM buildings (summarised in Figure 10.28) can be broadly categorised as:

local failures – these include the toppling of parapets, walls not carrying joists or beams under face load, and materials falling from damaged in-plane walls. These local failures could cause significant life-safety hazards, although buildings may still survive these failures.

global failures – these include failure modes leading to total collapse of a building due to such factors as loss of load path and deficient configuration.

Figure 10.28: Failure modes of URM buildings

In URM buildings, in-plane demands on walls decrease up the height of the walls. In-plane capacity also decreases with height as the vertical load decreases. In contrast, out-of-plane demands are greatest at the upper level of walls (Figure 10.29), but out-of plane capacity is lowest in these areas due to a lack of vertical load on them. Hence, the toppling of walls starts from the top unless these are tied to the diaphragm.

Figure 10.29: Out-of-plane vibration of masonry walls are most pronounced at the top floor

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Figure 10.38: Effect of types of diaphragm on face-loaded walls – a) inferior wall-to-wall

connection and no diaphragm, b) good wall-to-wall connection and ring beam with flexible diaphragm, c) good wall-to-wall connection and rigid diaphragm

Figures 10.39 and 10.40 show images of damage to masonry buildings due to collapse of walls under face load.

Figure 10.39: Out-of-plane instability of wall under face load due to a lack of ties between

the face-loaded wall and rest of the structure (Richard Sharpe)

Gable end walls sit at the top of walls at the end of buildings with pitched roofs. If this triangular portion of the wall is not adequately attached to the roof or ceiling, it will rock as a free cantilever (similar to a chimney or parapet) so is vulnerable to collapse. This is one of the common types of out-of-plane failure of gable walls (Figure 10.40).

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Figure 10.43: Rocking and delamination of bricks of a one-storey unreinforced brick

masonry building with reinforced concrete roof slab (Bothara & Hiçyılmaz, 2008)

Sliding shear can occur along a distinctly defined mortar course (Figure 10.44(a)) or over a limited length of several adjacent courses, with the length that slides increasing with height (Figure 10.44(b)). This can often be mistaken for diagonal tension failure, which is less common in walls with moderate to low axial forces.

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Figure 10.44: Sliding shear failure in a brick masonry building

Alternatively, masonry piers subjected to shear forces can experience diagonal tension cracking, also known as X-cracking (Figure 10.45). Diagonal cracks develop when tensile stresses in the pier exceed the masonry tensile strength, which is inherently very low. This type of damage is typically observed in long and squat piers and on the bottom storey of buildings, where gravity loads are relatively large and the mortar is excessively strong.

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Step 1 Gather documentation

Collect relevant information and documents about the building including drawings, design feature reports, calculations and specifications, and any historical material test results and inspection reports (if available).

If the building has been previously altered or strengthened, collect all available drawings, calculations and specifications of this work.

Study this information before proceeding with the on-site investigation.

Step 2 Consider building complexity

Determine an assessment strategy based on an initial appraisal of the complexity of the building. This can be reviewed as the assessment progresses.

Although all aspects will need to be considered for all buildings, simplifications can be made for basic buildings e.g. one or two storey commercial, rectangular in plan. For these buildings the default material strengths are expected to be adequate without further consideration so that on-site testing, other than scratch testing of the bed joints to ascertain mortar type and quality, is not considered necessary. Foundation rotations are also not expected to have a significant effect so can be ignored.

Concentration of effort should be on assessing the score for face-loaded walls, connections from the walls to the diaphragms and the diaphragms (lateral deflection between supported walls). The score for the walls in plane will depend on the ability (stiffness) of the diaphragm to transfer the shears but the calculations required are likely to be simple irrespective of whether the diaphragms are rigid (concrete) or flexible (timber, steel braced). Behaviour can be assumed to be linear-elastic (ie ignore any non-linear behaviour).

Complexity is likely to be increased if a building has previously been retrofitted. Not all issues with the building will necessarily have been addressed in historical retrofits. Stiffness compatibility issues will often not have been considered or fully addressed.

Step 3 Investigate on-site

Refer Section 10.6.

Evaluate how well the documentation describes the “as constructed” and, where appropriate, the “as strengthened” building.

Carry out a condition assessment of the existing building.

Complete any on-site retrieval of samples and test these.

Identify any site conditions that could potentially affect the building performance. (Refer Section 14).

Step 4 Assign material properties

Start by using the probable material properties that are provided in Section 10.8, or establish actual probable values through intrusive testing (this may be a step you come back to depending on the outcome of your assessment).

Recognise that for basic buildings obtaining building-specific material strengths through testing may not be necessary to complete an assessment.




Step 5 Identify potential structural weaknesses and relative vulnerability

The first step is to identify all of the various components in the building and then to identify potential SWs related to these.

The identification of potential SWs in this type of building requires a good understanding of the issues discussed in Sections 10.2, 10.3 and 10.4.

Early recognition of SWs and their relative vulnerability and interdependence is likely to reduce assessment costs and focus the assessment effort.

Prior experience is considered essential when identifying the SWs in complex buildings.

Separate the various components into those that are part of the primary lateral load resisting system and those that are not (secondary components). Some components may be categorised as having both a primary lateral load resisting function (e.g. in-plane walls and shear connections to diaphragms) and a secondary function (e.g. face-loaded walls and supporting connections).

The relative vulnerability of various components in typical URM buildings is likely to be (refer also Figure 10.54):

- Inadequately restrained elements located at height; such as street-facing façades, unrestrained parapets, chimneys, ornaments and gable end walls. Collapse of these components may not lead to building collapse but they are potential life-safety hazards and therefore their performance must be reflected in the overall building score.

- Inadequate connection between face-loaded walls and floors/roof; little or no connection capacity will mean that the walls will not be laterally supported when the inertial wall forces are in a direction away from the building and then it can be easily concluded that the walls and/or connections will be unlikely to score above 34%NBS, except perhaps in low-seismic regions. If observations indicate reasonable diaphragm action from the floors and/or roof, adequate connections will mean that the out-of-plane capacity of the face-loaded walls may now become the limiting aspect.

- Out-of-plane instability of face-loaded walls. If the wall capacity is sufficient to meet the requirements set out for face-loaded walls, then the capacity of the diaphragms becomes important as the diaphragms are required to transfer the seismic loads from the face-loaded walls into the in-plane walls.

- The in-plane capacity of walls: these are usually the least vulnerable components.

Step 6 Assess component capacities

Calculate the seismic capacities from the most to the least vulnerable component, in turn. There may be little point in expending effort on refining existing capacities only to find that the capacity is significantly influenced by a more vulnerable item that will require addressing to meet earthquake-prone requirements or target performance levels. Connections from brick walls to floors/roof diaphragms are an example of this. Lack of ties in moderate to high seismic areas will invariably result in an earthquake-prone status for the masonry wall and therefore it may be more appropriate and useful to assess the wall as < 34%NBS and also calculate a capacity assuming ties are in place. This will inform on the likely effect of retrofit measures.

A component may consist of a number of individual elements. For example, the capacity of a penetrated wall (a component) loaded in-plane will need to consider the




likely behaviour of each of the piers and the spandrel regions between and above and below the openings respectively (the elements). For some components the capacity will be a function of the capacity of individual elements and the way in which the elements act together. To establish the capacity of a component may therefore require structural analysis of the component to determine the manner in which actions in the elements develop.

For each component assess whether or not exceeding its capacity (this may be more easily conceptualised as failure for these purposes) would lead to a life safety issue. If it is determined that it will not, then that component can be neglected in the assessment of the expected seismic performance of the structure. The same decisions may need to be made regarding the performance of elements within a component.

Step 7 Analyse the global structure

In general, the complexity and extent of the analysis should reflect the complexity of the building.

Start with analyses of low sophistication, progressing to greater sophistication only as necessary.

An analysis of the primary lateral load resisting structure will be required to determine the relationship between the global capacity and the individual component actions.

The analysis undertaken will need to recognise that the capacity of components will not be limited to consideration of elastic behaviour. Elastic linear analysis will likely be the easiest to carry out but the assessor must recognise that restricting to elastic behaviour will likely lead to a conservatively low assessment score.

The analysis will need to consider the likely impacts of plan eccentricities (mass, stiffness and/or strength).

Step 8 Assess global capacity

From the structural analyses determine the global capacity of the building. This will be the capacity of the building as a whole determined at the point that the most critical component of the primary lateral load resisting system reaches its determined capacity.

It may also be useful to determine the global capacity assuming successive critical components are addressed (retrofitted). This will inform on the extent of retrofit that would be required to achieve a target score.

Step 9 Determine the demands and %NBS

Determine the global demand for the building from Section 5 and assess the global %NBS (global capacity/ global demand x 100).

Assess the demands on secondary components and parts of the building and assess %NBS for each (capacity/demand x 100).

List the %NBS values in a table.

The CSW will be the item in the table with the lowest %NBS score and that %NBS becomes the score for the building.

Review the items in the %NBS table to confirm that all relate to elements, the failure of which would lead to a life safety issue. If not, revise the assessment to remove the non-life safety element from consideration.


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Structural Mechanism

Technique Comments/Issues

Steel capping spanning between abutting frames or walls

Anchorage depth down into mass of parapet to clamp down loose upper bricks

Internal Post-tensioning Anchorage depth down into mass of parapet to clamp down loose upper bricks

External post-tensioning Anchorage depth down into mass of parapet to clamp down loose upper bricks

Internal bonded reinforcement Anchorage depth down into mass of parapet to clamp down loose upper bricks

Near Surface Mounted (NSM) composite strips

Parapet responds differently to different directions of load

UV degradation

Face-loaded walls

Vertical steel mullions (Figure 10.23) Stiffness vs out-of-plane rocking/displacement capability important

Regularity/robustness of attachment to wall is important

Vertical timber mullions Stiffness vs out-of-plane rocking/displacement capability important

Regularity/robustness of attachment to wall is important

Horizontal transoms spanning between abutting frames or walls

Stiffness and attachment requirements need to consider wall above which gives clamping action to masonry at level of attachment

Internal post-tensioning Durability

Anchorage level and fixity

Level of pre-stress to allow rocking without brittle crushing

External post-tensioning As above

Internal bonded reinforcement Maximum quantity to ensure ductile failure

Anchorage beyond cracking points, and consider short un-bonded lengths

Composite fibre overlay Preparation to give planar surface very involved

Near Surface Mounted (NSM) composite strips

Wall responds differently to different directions of load

Bond important if in-plane capacity is not to be weakened

Reinforced concrete overlay Wall responds differently to different directions of load

Reinforced cementitious overlay Wall responds differently to different directions of load

Ductility of reinforcement important for deflection capacity

Grout saturation/injection Elastic improvement only: more suitable for low seismic zones and very weak materials






Connection of walls to diaphragms

Steel angle with grouted bars (Figure 10.24(a))

Bar anchorage

Diaphragm/bar eccentricity must be resolved

Steel angle with bolts/external plate (Figure 10.24(b))

Diaphragm/bar eccentricity must be resolved

Timber joist/ribbon plate with grouted bars

Bar anchorage

Diaphragm/bolt eccentricity causes bending of timber across grain - a potential point of weakness

Timber joist/ribbon plate with bolts/external plate

Diaphragm/bolt eccentricity causes bending of timber across grain - a potential point of weakness

Blocking between joists notched into masonry

Joist weak axis bending must be checked

Tightness of fit of joists into pockets

Degradation of joists

External pinning to timber beam end Quality assurance/buildability of epoxy in timber

Concentrated localised load

Development in masonry (external plate preferred for high loads)

External pinning to concrete beam or floor

Development in masonry (external plate preferred for high loads)

Concrete floor type (hollow pots, clinker concrete)

Through rods with external plates Elastic elongation

Concentrated localised load

New isolated padstones Tightness of fit

Resolution of eccentricity between masonry bearing and diaphragm connection

New bond beams High degree of intervention

Diaphragm strengthening

Plywood overlay floor or roof sparking (Figure 10.25)

Flexibility

Requires continuous chord members and primary resistance elements

Plywood ceiling As above, plus existing ceiling battening/fixings may not be robust or may be decayed

Plywood/light gauge steel composite Stiffer but less ductile than ply-only

Eccentricities between thin plate and connections must be resolved

Plasterboard ceiling As ply ceiling but less ductile

Prevention of future modification/removal

Thin concrete overlay/topping Thickness for adequate reinforcement Additional mass

Ductility capacity of non-traditional reinforcement

Buckling restraint/bond to existing structure






Elastic cross bracing Stiffness relative to wall out-of-plane capacity

Edge distribution members and chords critical

Concentration of loads at connections

Semi-ductile cross bracing (e.g. Proving ring)

As elastic

Energy absorption benefit not easily quantified without sophisticated analysis

Replacement floor over/below with new diaphragm

Design as new structure

In-plane wall strengthening

New primary strengthening elements (Figure 10.26)

Sprayed concrete overlay Restraint to existing floor/ roof structure

Out-of-plane capacity of wall

Ductility capacity if used very dependent on aspect ratio

Chords

Foundation capacity needs to be checked (uplift/rocking)

Internal vertical post-tensioning Ensure pre-stress limited to ensure no brittle failure

See out-of-plane issues also

External vertical post-tensioning Ensure pre-stress limited to ensure no brittle failure

See out-of-plane issues also

Internal horizontal reinforcement Coring/drilling difficult

Stressing horizontally requires good vertical (perpendicular) mortar placement and quality

External horizontal post-tensioning Stressing horizontally requires good vertical (perpendicular) mortar placement and quality

Bed-joint reinforcement Workmanship critical

Low quantities of reinforcement only possible

Composite reinforced concrete boundary or local reinforcement elements

Development at ends/nodes

Bond to existing

Composite FRP boundary or local reinforcement elements

As above plus stiffness compatibility with existing

Nominally ductile concrete walls or punched wall/frame

High foundation loads result

Nominally ductile reinforced concrete masonry walls

Stiffness compatibility considering geometry (including foundation movement) important

Nominally ductile steel concentric or cross bracing

Stiffness compatibility assessment critical considering element flexibility, plan position and diaphragm stiffness

Drag beams usually required

Limited ductility steel moment Frame Flexibility/stiffness compatibility very important

Limited ductility concrete frame Flexibility/stiffness compatibility important






Limited ductility concrete walls Assess effectiveness of ductility, including foundation movements

Ensure compatibility with any elements cast against

Drag beams often required

Limited ductility timber walls Flexibility/stiffness compatibility very important

Drag beams often required

Ductile EBF/K-frames Element ductility demand vs building ductility assessment important

Drag beams usually required

Ductile concrete coupled or rocking walls

Element ductility demand vs building ductility assessment important

Ensure compatibility with any elements cast against drag beams often required

Tie to new adjacent (new) structure Elastic elongation and robustness of ties to be considered

Higher level of strengthening likely to be required

Reinforcement at wall intersections in plan

Removal and rebuilding of bricks with inter-bonding

Shear connection only with capacity reduced considering adhesion and tightness of fit

Disturbance of bond to adjacent bricks

Bed joint ties Small reinforcement only practical but can be well distributed

Care with resolving resultant thrust at any bends

Drilled and grouted ties Tension only: consider shear capacity

Depth to develop capacity typically large

Compatibility with face-load spanning of wall

Metalwork reinforcing internal corner Attachment to masonry

Small end-distance in abutting wall can mean negligible tension capacity

Grouting of crack Shear friction only: tension mechanism also required

Stabilises any dilation but does not allow recovery

Foundation strengthening

Mass underpinning Creates hard point in softer/swellable soils

Even support critical

Grout injection Creates hard point in softer/swellable soils

Difficult to quantify accurately

Concentric/balanced re-piling Localised “needles” through walls must provide sufficient bearing for masonry

Eccentric re-piling with foundation beams

Stiffness of found beams important to not rotate walls out-of-plane

Mini piling/ground anchors Cyclic bond less than static bond

Testing – only static practical

Vulnerable to bucking if liquefaction






Pile type: vertical stiffness and pre-loading

Pre-loading dictates load position

Pre-loading important if new foundations less stiff than existing

Dynamic distribution between new and old likely different than static

Effects of liquefaction must be considered: may create limiting upper bound to strengthening level

Façade wythe ties

Helical steel mechanical engagement – small diameter

Low tension capacity, especially if cracked

Steel mechanical engagement – medium diameter

Some vierendeel action between wythes

Durability

Epoxied steel rods/gauze sleeve Some vierendeel action between wythes

Epoxied composite/non-metallic rods Stiffness

Brick header strengthening Additional new headers still brittle; can become overstressed under thermal/seasonal or foundation loadings in combination

Canopies Reinforce or recast existing hanger embedment

Degradation of steel

Depth of embedment to ensure sufficient mass of bricks to prevent pull-out

New steel/cast iron posts Propping of canopy can mitigate hazard from masonry falling to pavement

Props in addition to hangars are not so critical with regard to traffic damage

New cantilevered beams Co-ordination with clerestory/bressumer beam

Backspan reaction on floor

Deck reinforcement to mitigate overhead hazard

Sacrificial/crushable layer to mitigate pavement hazard

Conversion to accessible balcony Likely to achieves all of the above objectives for canopies and also has natural robustness as designed for additional live load. Hazard still exists for balcony occupants

Base isolation A lack of sufficient gap around the building

Vertically re-founding the building

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Figure 10.58: Orthogonal diaphragm response due to joist orientation

It is assumed here that the diaphragm is adequately secured to all perimeter walls via pocketing and/or anchorages to ensure that diaphragm deformation occurs rather than global sliding of the diaphragm on a ledge. It is also assumed that the URM boundary walls deform out-of-plane in collaboration with deformation of the flexible timber diaphragm. For non-rectangular diaphragms, use the mean dimensions of the two opposing edges of the diaphragm to establish the appropriate dimensions of an equivalent rectangular diaphragm. Note:

Timber roofs of unreinforced masonry buildings were often built with both a roof and ceiling lining. As a result, roof diaphragms are likely to be significantly stiffer than the mid-height floor diaphragms if there are no ceilings on the mid-floors. Diagonal sarking in the roof diaphragm will also further increase its relative stiffness compared to the floor diaphragms. If the diaphragm you are assessing has an overlay or underlay (e.g. of plywood or pressed metal sheeting), consult the stiffness and strength criteria for improved diaphragms. You will still need to consider stiffness and ductility compatibility between the two. For example, it is likely that a stiff, brittle timber lath-and-plaster ceiling will delaminate before any straight sarking in the roof above can be fully mobilised. While the flooring, sarking and sheathing provide a shear load path across the diaphragm, it is necessary to consider the connections to the surrounding walls (refer to Section 10.8.4) and any drag or chord members. A solid URM wall may be able to act as a chord as it has sufficient in-plane capacity to transfer the chord loads directly to the ground. However, a punched URM wall with lintels only over the openings will have little tension capacity and may be the critical element in the assessment. Timber trusses and purlins, by their nature, only occur in finite lengths: their connections/splices designed for gravity loads may have little tie capacity.




Probable strength capacity

The probable strength capacity of a timber diaphragm should be assessed in accordance with Section 11 of these guidelines.

Probable deformation capacity

Deformations in timber diaphragms should be assessed using the effective diaphragm stiffness defined below. The probable deformation capacity should be taken as the lower of the following, assessed for each direction:

L/33 for loading oriented perpendicular to the joists or L/53 for loading oriented parallel to the joists

Deformation limit to provide adequate support to face-loaded walls. Refer Section 10.8.3.2.

Deformation required to meet global inter-storey drift limit of 2.5% in accordance with NZS 1170.5. Refer Section 10.8.3.1.

Effective diaphragm stiffness

To determine the effective stiffness of a timber diaphragm, first assess the condition of the diaphragm using the information in Table 10.7. Table 10.7: Diaphragm condition assessment criteria (Giongo et al., 2014)

Condition rating Condition description

Poor Considerable borer; floorboard separation greater than 3 mm; water damage evident; nail rust extensive; significant timber degradation surrounding nails; floorboard joist connection appears loose and able to wobble

Fair Little or no borer; less than 3 mm of floorboard separation; little or no signs of past water damage; some nail rust but integrity still fair; floorboard-to-joist connection has some but little movement; small degree of timber wear surrounding nails

Good Timber free of borer; little separation of floorboards; no signs of past water damage; little or no nail rust; floorboard-to-joist connection tight, coherent and unable to wobble

Next, select the diaphragm stiffness using Table 10.8 and accounting for both loading orientations. Note:

While other diaphragm characteristics such as timber species, floor board width and thickness, and joist spacing and depth are known to influence diaphragm stiffness, their effects on stiffness can be neglected for the purposes of this assessment.

Pretesting has indicated that re-nailing vintage timber floors using modern nail guns can provide a 20% increase in stiffness.




Table 10.8: Shear stiffness values† for straight sheathed vintage flexible timber floor diaphragms (Giongo et al., 2014)

Direction of loading Joist continuity Condition rating Shear stiffness†, Gd (kN/m)

Parallel to joists Continuous or discontinuous joists Good 350

Fair 285

Poor 225

Perpendicular to joists††

Continuous joist, or discontinuous joist with reliable mechanical anchorage

Good 265

Fair 215

Poor 170

Discontinuous joist without reliable mechanical anchorage

Good 210

Fair 170

Poor 135

Note: † Values may be amplified by 20% when the diaphragm has been renailed using modern nails and nail guns †† Values should be interpolated when there is mixed continuity of joists or to account for continuous sheathing at joist

splice

For diaphragms constructed using other than straight sheathing, multiply the diaphragm stiffness by the values given in Table 10.9. If roof linings and ceiling linings are both assumed to be effective in providing stiffness, add their contributions. Table 10.9: Stiffness multipliers for other forms of flexible timber diaphragms (derived from ASCE, 2013)

Type of diaphragm sheathing Multipliers to account for other sheathing types

Single straight sheathing x 1.0

Double straight sheathing

Chorded x 7.5

Unchorded x 3.5

Single diagonal sheathing Chorded x 4.0

Unchorded x 2.0

Double diagonal sheathing or straight sheathing above diagonal sheathing

Chorded x 9.0

Unchorded x 4.5

For typically-sized diaphragm penetrations (usually less than 10% of gross area) the reduced diaphragm shear stiffness, G’d, is given by Equation 10.6:

/ …10.6

where Anet and Agross refer to the net and the gross diaphragm plan area (in square metres). For non-typical sizes of diaphragm penetration, a special study should be undertaken to determine the influence of diaphragm penetration on diaphragm stiffness and strength. The




effective diaphragm stiffness must be modified further to account for stiffness of the URM boundary walls deforming in collaboration with the flexible timber diaphragm. Hence:

, kN/m …10.7

where αw may be determined using any rational procedure to account for the stiffness and incompatibility of deformation modes arising from collaborative deformation of the URM walls displacing out-of-plane as fixed end flexure beams and the diaphragm deforming as a shear beam. In lieu of a special study, prior elastic analysis has suggested that Equation 10.8 provides adequate values for αw:

α ≅ 1 ℓ

ℓ …10.8

where

ℓ = effective thickness of walls below the diaphragm, m = effective l thickness of walls above the diaphragm, m ℓ = height of wall below diaphragm, m = height of wall above diaphragm, m

Em = Young’s modulus of masonry, MPa = depth of diaphragm, m. = span of diaphragm perpendicular to loading, m.

Refer to Figure 10.59 for definition of the above terms. For scenarios where the URM end walls are likely to provide no supplementary stiffness to the diaphragm, αw = 1.0 should be adopted.

Figure 10.59: Schematics showing dimensions of diaphragm

B

Direction of loading

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Table 10.10: Default anchor probable shear strength capacities for anchors into masonry units only1.

Anchorage type Rod size Probable shear strength

capacity2,

(kN)

Bolts/steel rods fixed through and bearing against a timber member1,2

M12 8.5

M16 15

M20 18.5

Bolts/steel rods fixed through a steel member (washer) having a thickness of 6 mm or greater

M16 20

Note: 1. Anchors into mortar bed joints will have significantly lower shear capacities 2. Timber member to be at least 50 mm thick and MSG8 grade or better 3. For adhesive connectors embedment should be at least 200 mm into solid masonry

Table 10.11: Default anchor probable tension pull-out capacities for 0m, >0.3m and > 3m of wall above the embedment)

Mortar hardness Single-wythe wall

(kN)

Embedment 160 mm1 into two-wythe wall

(kN)

Embedment 250 mm1 into three-wythe wall

(kN)

0 >0.3 m(3) >3 m 0 >0.3 m(3) >3 m 0 >0.3 m(3) >3 m

Very soft 0.3 0.5 1 1 1.5 4 1.5 3 8

Soft 1 1.5 3 2.5 4 9 5 8 18

Medium 1.5 2.5 6 4 6.5 15 8 14 31

Hard 2.5 3.5 8 6 9 21 11 19 43

Very hard >2.5(4) >4(4) >8(4) >6(4) >10(4) >21 >11(4) >20(4) >43(4)

Notes: 1. Representative value only: assumes drilling within 50 mm of far face of wall 2. Simultaneous application of tension and shear loading need not be considered 3. These values are intended to be used until there is >3 m of wall above the embedment. 4. Values for very hard mortar may be substantiated by calculation but can be assumed to be at least those shown.

The values in Table 10.11 are based on the pull-out of a region of brick, assuming cohesion or adhesion strength of the mortar on the faces of the bricks perpendicular to the application of the load factored by 0.5 and friction on the top and/or bottom faces (refer Figure 10.60), depending on the height of wall above the embedment as follows:

0 m (ie at the top of the wall) - adhesion only on the bottom and side faces

>0.3 m but < 3 m – adhesion on the top, bottom and side faces, friction on the top and bottom faces

>3 m – cohesion on the top, bottom and side faces, friction on the top and bottom faces. A factor of 0.5 has been included in these values to reflect the general reliability of mechanisms involving cohesion/adhesion and friction.


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The direct tensile strength, f’t, should be ignored in capacity calculations unless there is no sign of pre-cracking in the wall at the section being considered and the demand is assessed assuming fully elastic behaviour and taking Sp =2 (synonymous with applying a 0.5 factor to the capacity) and cracking of the brickwork in the region of the section is not expected for loading in-plane.

Inelastic displacement-based analysis for walls spanning vertically between supports

Follow the steps below to assess the displacement response capability and displacement demand in order to determine the adequacy of the walls. Note:

Appendix 10B provides some guidance on methods for determining key parameters. Refer to Figure 10B.1 for the notation employed.

We have also provided some approximations you can use (listed after these steps) if wall panels are uniform within a storey (approximately rectangular in vertical and horizontal section and without openings).

Charts are provided in Appendix 10C that allow assessment of %NBS for regular walls (vertically spanning and vertical cantilever) in terms of height to thickness ratio of the wall, gravity load on the wall and parameters defining the demand on the wall. The wall panel is assumed to form hinge lines at the points where effective horizontal restraint is assumed to be applied. The centre of compression on each of these hinge lines is assumed to form a pivot point. The height between these pivot points is the effective panel height h (in mm). At mid-height between these pivots, height h/2 from either, a third pivot point is assumed to form. The recommended Steps for assessment of walls following the displacement-based method are discussed below:

Step1

Divide the wall panel into two parts: a top part bounded by the upper pivot and the mid height between the top and bottom pivots; and a bottom part bounded by the mid-height pivot and the bottom pivot.

Note:

This division into two parts is based on the assumption that a significant crack will form at the mid height of the wall, where an effective hinge will form. The two parts are then assumed to remain effectively rigid. While this assumption is not always correct, the errors introduced by the resulting approximations are not significant.

One example is that significant deformation occurs in the upper part of top-storey walls. In particular, where the tensile strength of the mortar is small the third hinge will not necessarily form at the mid height.




Step 2

Calculate the weight of the wall parts: Wb (in N) of the bottom part and Wt (in N) of the top part, and the weight acting at the top of the storey, P (in N).

Note:

The weight of the wall should include any render and linings, but these should not be included in tnom or t (in mm) unless the renderings are integral with the wall. The weight acting on the top of the wall should include all roofs, floors (including partitions and ceilings and the seismic live load) and other features that are tributary to the wall.

Step 3

From the nominal thickness of the wall, tnom, calculate the effective thickness, t.

Note:

The effective thickness is the actual thickness minus the depth of the equivalent rectangular stress block. The reduction in thickness is intended to reflect that the walls will not rock about their edge but about the centre of the compressive stress block.

The depth of the equivalent rectangular stress block should be calculated with caution, as the depth determined for static loads may increase under earthquake excitation. Appendix 10B suggests a reasonable value based on experiments, t = tnom (0.975-0.025 P/W). The thickness calculated by this formula may be assumed to apply to any type of mortar, provided it is cohesive. For weaker (and softer) mortars, greater damping will compensate for any error in the calculated t.

Step 4

Assess the maximum distance, ep, from the centroid of the top part of the wall to the line of action of P. Refer to Figure 10B.1 for definition of eb, et and eo. Usually, the eccentricities eb and ep will each vary between 0 and t/2 (where t is the effective thickness of the wall). Exceptionally they may be negative, i.e. where P promotes instability due to its placement. When considering the restraint available from walls on foundations assume the foundation is the same width as the wall and use the following values for eb:

0 if the factor of safety for bearing under the foundation, for dead load only (FOS), is equal to 1

t/3 if FOS = 3 (commonly the case) t/4 if FOS = 2

Note:

Figure 10B.2 shows the positive directions for the eccentricities for the assumed direction of rotation (angle A at the bottom of the wall is positive for anti-clockwise rotation).

The walls do not need to be rigidly attached or continuous with a very stiff section of wall beyond to qualify for an assumption of full flexural restraint.




Care should be taken not to assign the full value of eccentricity at the bottom of the wall if the foundations are indifferent and may themselves rock at moments less than those causing rocking in the wall. In this case, the wall might be considered to extend down to the supporting soil where a cautious appraisal should then establish the eccentricity. The eccentricity is then related to the centroid of the lower block in the usual way.

Step 5

Calculate the mid-height deflection, i, that would cause instability under static conditions. The following formula may be used to calculate this deflection.

…10.11

where:

…10.12

and:

…10.13

Note:

The deflection that would cause instability in the walls is most directly determined from virtual work expressions, as noted in Appendix 10B.

Step 6

Assign the maximum usable deflection, m (in mm), as 0.6 i.

Note:

The lower value of the deflection for calculation of instability limits reflects that response predictions become difficult as the theoretical limit is approached. In particular, the response becomes overly dependent on the characteristics of the earthquake, and minor perturbances lead quickly to instability and collapse.

Step 7

Calculate the period of the wall, Tp, as four times the duration for the wall to return from a displaced position measured by t (in mm) to the vertical. The value of Δt is less than Δm. Research indicates that t = 0.6m =0.36i for the calculation of an effective period for use in an analysis using a linear response spectrum provides a close approximation to the results of more detailed methods. The period may be calculated from the following equation:

4.07 …10.14




where J is the rotational inertia of the masses associated with Wb, Wt and P and any ancillary masses, and is given by the following equation:

…10.15

where Jbo and Jto are mass moment of inertia of the bottom and top parts about their centroids, and Janc is the inertia of any ancillary masses, such as veneers, that are not integral with the wall but that contribute to the inertia. When treating cavity walls, make the following provisions:

When the veneer is much thinner than the main wythe, the veneer can be treated as an appendage. For inelastic analysis, the veneers can be accounted through Janc.

If both wythes are a one - brick (110 mm) thick, then these could be treated as independent walls. Allocate appropriate proportion of overburden on them and solve the problem in the usual way.

Where an accurate solution is the objective, solve the general problem with the kinematic constraint that the two walls deflect the same.

Note:

The equations are derived in Appendix 10B. You can use the method in this appendix to assess less common configurations as necessary.

Step 8

Calculate the design response coefficient Cp(Tp) in accordance with Section 8 NZS 1170.5 taking p =1 and substituting C (Tp):

…10.16

where: Chc(Tp) = the spectral shape factor ordinate, Ch(Tp), from NZS 1170.5 for

Ground Class C and period Tp, provided that, solely for the purpose of calculating Chc(Tp), Tp need not be taken less than 0.5 sec.

When calculating CHi from NZS 1170.5 for walls spanning vertically and held at the top, hi should be taken as the average of the heights of the points of support (typically these will be at the heights of the diaphragms). In the case of vertical cantilevers, hi should be measured to the point from which the wall is assumed to cantilever. If the wall is sitting on the ground and is laterally supported above, hi may be taken as half of the height to the point of support. If the wall is sitting on the ground and is not otherwise attached to the building it should be treated as an independent structure, not as a part. This will involve use of the appropriate ground spectrum for the site.




Note:

The above substitution for Ci(Tp) has been necessary because the use of the tri-linear function given in NZS 1170.5 (Equations 8.4(1), 8.4(2) and 8.4(3) does not allow appropriate conversion from force to displacement demands. The revised Ci(Tp) converts to the following, with the numerical numbers available from NZS 1170.5 Table 3.1.

Ci(Tp) = 2.0 for Tp < 0.5sec

= 2.0(0.5/ Tp)0.75 for 0.5 < Tp < 1.5sec

= 1.32/ Tp for 1.5 < Tp < 3sec

= 3.96/ Tp2 for Tp > 3sec

Only 5% damping should be applied. Experiments show that expected levels of damping from impact are not realised: the mating surfaces at hinge lines tend to simply fold onto each other rather than impact.

Step 9

Calculate , the participation factor for the rocking system. This factor may be taken as:

…10.17

Note:

The participation factor relates the response deflection at the mid height of the wall to the response deflection for a simple oscillator of the same period and damping.

Step 10

From Cp(Tp), Tp, Rp and calculate the displacement response, Dph (in mm) as:

/2 . . ...10.18

where: Cp(Tp) = the design response coefficient for face-loaded walls (refer Step 8

above, and for more details refer to Section 10.10.3) Tp = Period of face-loaded wall, sec Rp = the part risk factor as given by Table 8.1, NZS 1170.5 Cp(Tp) Rp ≤ 3.6.

Note that with Tp expressed in seconds, the multiplied terms (Tp/2π)2 × Cp(Tp) × g may be closely approximated in metres by:

/2 MIN /3, 1 …10.19




Step 11

Calculate

% 100 ∆ / 60 ∆ / …10.20

Note:

The 0.6 factor applied to i reflects that response becomes very dependent on the characteristics of the earthquake for deflections larger than 0.6i.

The previous version of these guidelines allowed a 20% increase in %NBS calculated by the above expression. However that is not justified now that different displacements are used for capacity and for the period and the subsequent calculation of demand. Note:

Steps 12 to 14 are only required for anchorage design.

Step 12

Calculate the horizontal accelerations that would just force the rocking mechanism to form. The acceleration may be assumed to be constant over the height of the panel, reflecting that it is associated more with acceleration imposed by the supports than with accelerations associated with the wall deflecting away from the line of the supports. Express the acceleration as a coefficient, Cm, by dividing by g.

Note:

Again, virtual work proves the most direct means for calculating the acceleration. Appendix 10B shows how and derives the following expression for Cm, in which the ancillary masses are assumed part of Wb and Wt.

…10.21

Note:

To account for the initial enhancement of the capacity of the rocking mechanism due to tensile strength of mortar and possible rendering, we recommend that Cm be cautiously assessed when mortar and rendering are present or in the case of retrofit likely to be added. The value of Cm may also be too large to use for the design of connections. Accordingly, it is recommended that Cm need not be taken greater than the maximum part coefficient determined from Section 8 NZS 1170.5 setting Rp and p =1.0.

Step 13

Calculate Cp(0.75), which is the value of Cp(Tp) for a part with a short period from NZS 1170.5 and define a seismic coefficient for the connections which is the lower of Cm, Cp(0.75) or 3.6




Note:

Cp(0.75) is the short period ordinate of the design response coefficient for parts from NZS 1170.5, and 3.6g is the maximum value of Cp(Tp) required to be considered by NZS 1170.5 when Rp and p =1.0 .

Step 14

Calculate the required support reactions using the contributing weight of the walls above and below the connection (for typical configurations this will be the sum of Wb and Wt for the walls above and below the support accordingly) and the seismic coefficient determined in Step 13.

Step 15

Calculate

%NBS = Capacity of connection from Section 10.8.4 x 100 …10.22 Required support reaction from Step 14

Note:

If supports to face-loaded walls are being retrofitted, we recommend that the support connections are made stronger than the wall(s) and not less than required using a seismic coefficient of Cp(0.75), i.e. do not take advantage of a lower Cm value.

Simplifications for regular walls

You can use the following approximations if wall panels are uniform within a storey (approximately rectangular in vertical and horizontal section and without openings) and the inter-storey deflection does not exceed 1% of the storey height. The results are summarised in Table 10.12. The steps below relate to the steps for the general procedure set out above.

Step 1 Divide the wall as before.

Step 2 Calculate the weight of the wall, W (in N), and the weight applied at the top of the storey, P (in N).

Step 3 Calculate the effective thickness as before, noting that it will be constant.

Step 4 Calculate the eccentricities, eb, et and ep. Each of these may usually be taken as either t/2 or 0.

Step 5 Calculate the instability deflection, i from the formulae in Table 10.12 for the particular case.

Step 6 Assign the maximum usable deflection, m, for capacity as 60% of the instability deflection.

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The eccentricshould be ent

mic Assessmen

.

which h is ex

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culate the paexpression

ue of 1.5 or le 10.12.

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0

0

0

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t/2

2)[h2 +7t2]

+Pt2}/g

4P/W)t/h

the piers shown

ot related to a bom the central p

the sketches isive number.

ed Masonry Buil

metres.

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1

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(W/2+P

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boundary condipivot point to t

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10.16.

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Vertical cantilevers

Parameters for assessing vertical cantilevers, such as partitions and parapets are derived in Appendix 10B. Please consult this appendix for general cases. For parapets of uniform rectangular cross-section, you may use the following approximations. These steps relate to the steps set out earlier for the general procedure for walls spanning between vertical diaphragms.

Step 1 You do not need to divide the parapet. Only one pivot is assumed to form: at the base.

Step 2 The weight of the parapet is W (in N). P (in N) is zero.

Step 3 The effective thickness is t (in mm) = 0.98tnom.

Step 4 Only eb is relevant. It is equal to t/2.

Step 5 The instability deflection measured at the top of the parapet i = t.

Step 6 The maximum usable deflection measured at the top of the parapet m = 0.3i = 0.3t.

Step 7 The period may be calculated from the assumption that t = 0.8m = 0.24i.

0.65 1 …10.24

in which h, the height of the parapet above the base pivot, and t, the thickness of the wall, are expressed in metres. The formulation is valid for P = 0, eb = t/2, yb = h/2 and approximating t = tnom.

Step 8 Calculate Cp(Tp) (refer to Step 8 of the general procedure for walls spanning vertically between diaphragms).

Step 9 Calculate = 1.5/[1+(t/h)2] ≤ 1.5. …10.25

Step 10 Calculate Dph from Cp(Tp), Tp and and as before.

Step 11 Calculate %NBS as for the general procedure for walls spanning between a floor and an upper floor or roof, from;

% 100∆ / 30∆ / 30 / . ...10.26

Note:

Steps 12 to 14 are only required for anchorage design.

Step 12 Calculate Cm = t/h. …10.27

Step 13 Calculate Cp(0.75) which is the value of Cp(Tp) for a part with a short period from NZS 1170.5. and define a seismic coefficient for the connections which is the lower of Cm, Cp(0.75) and 3.6.




Step 14 Calculate the base shear from W, Cm and Cp(0.75). This base shear adds to the reaction at the roof level restraint.

Note:

Charts are provided in Appendix 10C that allow the %NBS to be calculated directly for various boundary conditions for regular walls cantilevering vertically, given h/tGross for the wall, gravity load on the wall and factors defining the demand.

Gables

Figure 10.63(a) shows a gable that is:

free along the vertical edge

simply supported along the top edge (at roof level), and

continuous at the bottom edge (ceiling or attic floor level). This somewhat unusual case is useful in establishing parameters for more complex cases. The following parameters can be derived from this gable:

2 3 …10.28

32 …10.29

Note:

In the above equations, W and P are total weights, not weights per unit length. Also note that the participation factor now has a maximum value of 2.0 (t << h, P = 0). These results can be used for the gable in Figure 10.63(b) to provide a cautious assessment that does not recognise all of the factors that could potentially enhance the performance of such gables, such as the beneficial effects of membrane action Note:

There are several factors that enhance performance in gables like those shown in Figure 10.63(a), all of which relate to the occurrence of significant membrane action. Guidance on this aspect will be provided in future versions of this document when the necessary research (including testing) has been undertaken. (Please also refer to the following section on walls spanning horizontally and vertically.)




(a) Basic gable wall for defining parameters

(b) Typical gable for which results from (a) can be applied

Figure 10.63: Gable configurations

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Refer to Figure 10.65 for the definition of heff. This failure mode occurs when the diagonal tensile strength of a wall or pier is exceeded by the principal stresses. It is one of the undesirable failure modes as it causes a rapid degradation in strength and stiffness after the formation of cracking, ultimately leading to loss of load path. For this reason a deformation limit of y for this failure mode is recommended. This failure mode is more common where axial stresses are high, piers are squatter and the tensile strength of masonry is low. Diagonal tension failure leads to formation of an inclined diagonal crack that commonly follows the path of bed and head joints through the masonry, because of the lower strength of mortar compared to brick. However, cracking through brick is also possible if the mortar is stronger. In New Zealand masonry, the crack pattern typically follows the mortar joint. For conditions where axial stresses on walls or piers are relatively low and the mortar strengths are also low compared to the splitting strengths of the masonry units, diagonal tension actions may be judged not to occur prior to bed-joint sliding. However, there is no available research to help determine a specific threshold of axial stress and relative brick and mortar strengths that differentiates whether cracking occurs through the units or through the mortar joints (ASCE, 2013).

Toe crushing capacity

The maximum toe crushing strength, Vtc, of a wall, pier or spandrel can be calculated using Equation 10.31 if no flanges are present or if you have decided to ignore them. If flanges are to be accounted for, refer to the section below.

0.5 1.

…10.31

where:

α = factor equal to 0.5 for fixed-free cantilever wall/pier, or equal to 1.0 for fixed-fixed wall/pier

P = superimposed and dead load at top of the wall/pier Pw = self-weight of wall/pier Lw = length of the wall/pier, mm heff = height to resultant of seismic force (refer to Figure 10.65), mm fa = axial compression stress due to gravity loads at mid height of

wall/pier, MPa f’m = masonry compression strength, MPa (refer to Section 10.7.3).

SeUpd

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d relationshgure 10.66.

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recommend

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n strength an

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The capacity for bed-joint sliding in masonry elements is a function of bond and frictional resistance. Therefore, Equation 10.33 includes both factors. However, with increasing cracking, the bond component is progressively degraded until only the frictional component remains. The probable residual wall sliding shear capacity, Vs,r, is therefore found from Equation 10.33 setting the cohesion, c, equal to 0.

Note:

It is recommended that the bed joint sliding capacity of a rocking wall/pier be limited to a lateral drift of 0.003. The lateral performance of a wall/pier is considered to be unreliable and not able to provide the level of resilience considered appropriate when the deflections exceed this value. Wall/pier elements that are not part of the seismic resisting system are expected be able to provide reliable vertical load carrying capacity at higher drifts, approaching 0.075. These greater limits can also be used for all wall/pier elements when cyclic stiffness and strength degradation are included in the analysis method used. Such an analysis will automatically include redistribution of the lateral loads between elements when this is necessary.

Slip plane sliding

A DPC layer, if present, will be a potential slip plane, which may limit the capacity of a wall. The capacity of a slip plane for no slip can be found from Equation 10.34:

…10.34

where: dpc = DPC coefficient of friction. Typical values are 0.2-0.5 for

bituminous DPC, 0.4 for lead, and higher (most likely governed by the mortar itself) for slate DPC

Other terms are as previously defined. Note:

Where sliding of a DPC layer is found to be critical, testing of the material in its current/in situ state may be warranted. Alternatively, parametric checks, where the effects of low/high friction values are assessed, may show that the DPC layer is not critical in the overall performance. Sliding on a DPC slip plane does not necessarily define the deformation capacity of this behaviour mode. Evaluating the extent of sliding may be calculated using the Newmark sliding block (Newmark, 1965) or other methods. However, exercise caution around the sensitivity to different types of shaking and degradation of the masonry above/below the sliding plane. Where sliding is used in the assessment to give a beneficial effect, this should be subject to peer review.

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Expected in-plane strength of URM spandrels should be the lesser of the flexural and shear strengths. is the chord rotation of the spandrel, relative to the piers.

Note:

It is considered prudent to limit the deformation capacity of a spandrel panel to a panel drift of 3y if its capacity is to be relied on as part of the seismic resisting system. Panel chord rotation capacities beyond 0.02 or 0.01 for rectangular and arched spandrels respectively, for panels that are not assumed to be part of the lateral seismic resisting system, are not recommended as the performance of the spandrel (ie ability to remain in place) could become unreliable at rotations beyond these limits. These greater limits can also be used for all spandrel elements when cyclic stiffness and strength degradation are included in the analysis method used. Such an analysis will automatically include redistribution of the lateral loads between elements when this is necessary and therefore the need to distinguish, in advance, between elements of the lateral and non-lateral load resisting systems is not required. Two generic types of spandrel have been identified: rectangular and those with shallow arches. Recommendations for the various capacity parameters for these two cases are given in the following sections. Investigations are continuing on appropriate parameters for deep arched spandrels. In the interim, until more specific guidance is available, it is recommended that deep arched spandrels be considered as equivalent rectangular spandrels with a depth that extends to one third of the depth of the arch below the arch apex. The geometrical definitions used in the following sections are shown on Figure 10.69.

Figure 10.69: Geometry of spandrels with timber lintel (a) and shallow masonry arch (b)

(Beyer, 2012)

Rectangular spandrels

The expected in-plane strength of URM spandrels with and without timber lintels can be determined following the procedures detailed below.




Note:

There is limited experimental information on the performance of URM spandrels with lintels made from materials other than timber. However, URM spandrels with steel lintels are expected to perform in a similar manner to those with timber lintels.

When reinforced concrete lintels are present the capacity of the spandrel can be calculated neglecting the contribution of the URM.

Peak flexural strength

The peak flexural strength of rectangular spandrels can be estimated using Equation 10.35 (Beyer, 2012). Timber lintels do not make a significant contribution to the peak flexural capacity of the spandrels so can be ignored.

…10.35

where: ft = equivalent tensile strength of masonry spandrel psp = axial stress in the spandrel hsp = height of spandrel excluding depth of timber lintel if present bsp = width of spandrel lsp = clear length of spandrel between adjacent wall piers.

Unless the spandrel is prestressed the axial stress in the spandrel can be assumed to be negligible when determining the peak flexural capacity. Equivalent tensile strength of masonry spandrel, ft, can be estimated using Equation 10.36:

1.3 0.5 …10.36

where: pp = mean axial stress due to superimposed and dead load in the

adjacent wall piers f = masonry coefficient of friction c = masonry bed-joint cohesion.

Residual flexural strength

Residual flexural strength of rectangular URM spandrels can be estimated using Equation 10.37 (Beyer, 2012). Timber lintels do not often make a significant contribution to the residual flexural capacity of URM spandrels so they can be ignored.

, 1

. ...10.37

where: psp = axial stress in the spandrel fhm = compression strength of the masonry in the horizontal direction

(0.5f’m).




Axial stresses are generated in spandrel elements due to the restraint of geometric elongation. Results from experimental research indicate that negligible geometric elongation can be expected when peak spandrel strengths are developed (Beyer, 2012; and Graziotti et al., 2012), as this is at relatively small spandrel rotations. As a result, there is little geometric elongation. Significant geometric elongation can occur once peak spandrel strengths have been exceeded, and significant spandrel cracking occurs within the spandrel, as higher rotations are sustained in the element. An upper bound estimate of the axial stress in a restrained spandrel, psp, can be determined using Equation 10.38 (Beyer, 2014):

1 ...10.38

where: fdt = masonry diagonal tension strength s = spandrel aspect ratio (lsp/hsp).

Equation 10.38 calculates the limiting axial stress generated in a spandrel associated with diagonal tension failure of the spandrel. The equation assumes the spandrel has sufficient axial restraint to resist the axial forces generated by geometric elongation. In most typical situations you can assume that spandrels comprising the interior bays of multi-bay pierced URM walls will have sufficient axial restraint such that diagonal tension failure of the spandrels could occur. Spandrels comprising the outer bays of multi-bay pierced URM walls typically have significantly lower levels of axial restraint. In this case the axial restraint may be insufficient to develop a diagonal tension failure in the spandrels. Sources of axial restraint that may be available include horizontal post-tensioning, diaphragm tie elements with sufficient anchorage into the outer pier, or substantial outer piers with sufficient strength and stiffness to resist the generated axial forces. For the latter to be effective the pier would need to have enough capacity to resist the applied loads as a cantilever. It is anticipated that there will be negligible axial restraint in the outer bays of many typical unstrengthened URM buildings. In this case you can assume the axial stress in the spandrel is nil when calculating the residual flexural strength.

Peak shear strength

Peak shear strength of rectangular URM spandrels can be estimated using Equations 10.39 and 10.40 (Beyer, 2012). Timber lintels do not make a significant contribution to the peak shear capacity of URM spandrels so can be ignored.

...10.39

.1 ...10.40

Unless the spandrel is prestressed you can assume the axial stress in the spandrel is negligible when determining the peak shear capacity. Equation 10.39 is the peak shear

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cos ...10.45

Unless the spandrel is prestressed you can assume the axial stress in the spandrel is negligible when determining the peak flexural capacity.

Residual flexural strength

You can estimate the residual flexural capacity of a URM spandrel with a shallow arch using Equation 10.46 (Beyer 2012) and by referring to Figure 10.69.

, 1

. …10.46

where dimension htot is defined in Figure 10.69. You can calculate spandrel axial stresses, psp, with the procedures set out in the previous section.

Figure 10.71: Spandrel with shallow arch. Assumed load transfer mechanism after flexural

(a) and shear (b) cracking. (Beyer, 2012)

Peak shear strength

You can estimate peak shear strength of a URM spandrel with a shallow arch using Equations 10.47 and 10.48 (Beyer, 2012):

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Figure 10.75. The global strength capacity can be referred to in terms of base shear capacity. The deformation capacity will be the lateral displacement at heff for the building consistent with the base shear capacity accounting for non-linear behaviour as appropriate. This section provides guidance on the assessment of the global capacity for both basic and complex buildings. It also provides guidance on methods of analysis and modelling parameters.




Figure 10.75: Global capacity assessment approach for URM buildings

Determine component capacity for each "line" of the seismic system

Compare the seismic shear distribution and the shear capacity of each component

to determine the criticality of each component

Scale the base shear from the analysis to to the shear capacity of the critical

component to determine (Vprob)global, base

Check that the implied shears can be transferred by the diaphragms and the

shear connections between the diaphragms and each component

Carry out a lateral load analysis for each "line" of the seismic system to determine

the shear distibution between components and over the height of each

component

Diaphragm stiffness?

RIGID

FLEXIBLE

Carry out a lateral load analysis to determine the seismic shear distribution

over the height of each component, taking into account accidental eccentricities and any strength/stiffness eccentricities from

the centre of mass

Diaphragm and

connections adequate?

NO

YES

Factor down (Vprob)global, base accordingly

(Vprob )global, base

Check the horizontal deformation of the diaphragms

Horizontal diaphragm

deformation limits met?

NO

YES

Factor down (Vprob)global, base accordingly

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Figure 10.77: Relationship between demand and capacity for a basic building with rigid diaphragms

In the above discussion it has been assumed that the diaphragms are stiff enough to provide the required support to the face-loaded walls orientated perpendicular to the direction of loading. Diaphragms are considered as primary structural components for the transfer of these actions and their ability to do so may affect the global capacity of the building in that direction. Limits have been suggested in Section 10.8.3.2 for the maximum diaphragm

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where: Ch(T1) = the spectral shape factor determined from Clause 3.1.2,

NZS 1170.5 for the first mode period of the building, T1, g Z = the hazard factor determined from Clause 3.1.4, NZS 1170.5 Ru = the return period factor, Ru determined from Clause 3.1.5,

NZS 1170.5 N(T1,D) = the near fault factor determined from Clause 3.1.6, NZS 1170.5 KR = the seismic force reduction factor determined from Table 10.14.

Table 10.14: Recommended force reduction factors for linear static method

Seismic performance/ controlling parameters

Force reduction factor, KR

Notes

Pier rocking, bed joint sliding, stair-step failure modes

3 Failure dominated by strong brick-weak mortar

Pier toe failure modes 1.5

Pier diagonal tension failure modes (dominated by brick splitting)

1.0 Failure dominated by weak brick-strong mortar

Spandrel failure modes 1.0

Note:

The concept of a ductility factor (deflection at ultimate load divided by the elastic deflection) can be meaningless for most URM buildings. The introduction of KR primarily reflects an increase in the damping available and therefore reduced elastic response rather than ductile capability assessed by traditional means. Therefore the displacements calculated from the application of C(T1) are the expected displacements and should not be further modified by KR. These force reduction factors apply in addition to relief from period shift (if any). Redistribution of seismic demands between individual elements of up to 50% is permitted when KR = 3.0 applies, provided that the effects of redistribution are accounted for in the analysis. When there are mixed behaviour modes among the walls/piers in a line of resistance, you can ignore the capacity of any piers for which KR is less than the value that has been adopted for the line of resistance. Otherwise, consider lower force reduction factors. If you have adopted higher force reduction factors, carefully evaluate the consequences of loss of gravity load support from any walls/piers that have been ignored. If there are mixed failure modes among the walls and piers in a line of resistance, the displacement compatibility between these piers and walls should be evaluated. For the case of perforated walls when a strong pier – weak spandrel mechanism governs the wall behaviour KR = 1.0 shall be adopted for the wall line as a whole, or the capacities of the spandrels can be ignored. When the contribution of the spandrels is ignored the higher KR factors detailed in Table 10.14 may be used provided the consequences of loss of the ignored spandrels are considered.


10.10.3

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Ghobarah, A. and El Mandooh Galal, K. , 2004, Out-of-plane strengthening of unreinforced masonry walls with openings. Journal of Composites for Construction, 8(4), 298-305.

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Lam, N.T.K., Griffith, M., Wilson, J., and Doherty, K. (2003). Time-history analysis of URM walls in out-of-plane flexure. Engineering Structures, 25(6), 743-754.

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Magenes G. and Calvi G. M., 1992, Cyclic behavior of brick masonry walls. In: Tenth world conference on earthquake engineering. 1992. p. 3517–22.

Magenes G., della Fontana, A., 1998, Simplified non-linear seismic analysis of masonry buildings. Proceedings of the Fifth International Masonry Conference. British Masonry Society, London.

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Chena, S.-Y. Moona, F.L. Yib, T. 2008, A macroelement for the nonlinear analysis of in-plane unreinforced masonry piers, Engineering Structures, 30 (2008) 2242–2252.

Simsir, C.C. (2004). Influence of diaphragm flexibility on the out-of-plane dynamic response of unreinforced masonry walls. PhD Thesis, University of Illinois at Urbana-Champaign, United States-- Illinois.




STM. (2002). Standard Test Method for Conducting Strength Tests of Panels for Building Construction (No. E72-02). ASTM International.

Wilson, A., 2012, (2012).’ Seismic assessment of timber floor diaphragms in unreinforced masonry buildings’, Doctoral dissertation, The University of Auckland, Auckland, New Zealand, March, 568p. https://researchspace.auckland.ac.New Zealand/handle/2292/14696

Yi, T., Moon, F. L., Leon, R. T., and Kahn, L. F _2006a. Lateral load tests on a two-storey unreinforced masonry building.” J. Struct. Eng., 132_5_ 643–652.

Seismic Assessment of Unreinforced Masonry Buildings – Appendix 10A On-site Testing

Section 10 - Seismic Assessment of Unreinforced Masonry Buildings App-1 Updated 22 April 2015 ISBN 978-0-473-26634-9

Appendix 10A: On-site Testing

10A.1 General

While the seismic response of URM buildings is significantly influenced by characteristics such as boundary conditions and the behaviour of inter-element connections, on-site testing of material properties improves the reliability of the seismic assessments, the numerical models that describe the seismic behaviour of URM buildings and may lead to less conservative retrofit designs. However, the non-homogenous nature of masonry combined with the age of URM buildings make it difficult to reliably predict the material properties of masonry walls. It is recommended that field sampling or field testing of URM elements is conducted. Field sampling refers to the extraction of samples from an existing building for subsequent testing offsite, while field testing refers to testing for material properties in-situ. A set of techniques are described in subsequent sections that can be used to determine masonry material properties. Before proceeding to on-site testing, it is important to sensibly understand what information will be collected from the investigation, how that would be used and what value the information will add to reliability of the assessment. Before deciding an investigation programme, sensitivity analyses should be undertaken to determine what assessment parameters are more important and likely to influence the assessment result and whether the default parameter values given are likely to be appropriate/sufficient. Only rarely should on-site testing be considered necessary for basic buildings.

10A.2 Masonry Assemblage (Prism) Material Properties

If masonry assemblage (prism) samples are to be extracted for laboratory testing they should be single leaf and at least three bricks high. If they are two leafs thick or more, cut them into single leaf samples. If present, remove rendering plaster from both sides of the samples. Cap the prepared samples using gypsum plaster to ensure uniform stress distribution. Test individual brick units and mortar samples as per Section 10A.3 when sampling of larger assemblages is not permitted or practical. Masonry properties can then be predicted using the obtained brick and mortar properties as set out in Section 10.7.

10A.2.1 Masonry compressive strength

Determine masonry compressive strength in accordance with ASTM C 1314-03b (ASTM 2003a). Figure 10A.1 shows a typical prism sample before testing. Aluminium frames are attached to the sample ends and a displacement gauge spans between the frames to measure the sample displacement.



ASTM C 1314-03b (ASTM 2003a) also enables you to determine the masonry modulus of elasticity (further detailed in Section 10A.2.2.1).

Figure 10A.1: Example of extracted sample with test rig attached for the prism

compression test

10A.2.2 Masonry modulus of elasticity

10A.2.2.1 Laboratory calibrated displacement measurement

Laboratory calibrated displacement measurement devices may be attached to the masonry prisms during the compression tests detailed in Section 10A.2.1. Incorporate a minimum of two measurement devices to record displacements at opposing sample faces. Their gauge lengths should cover the distance from the middle of the top brick to the middle of the bottom brick. Use the recorded measurement to derive the masonry stress-strain relationship and subsequently the masonry modulus of elasticity, Em. The stress and strain values considered in the calculation of Em are those between 0.05 and 0.70 times the masonry compressive strength (f’m).

10A.2.2.2 In situ deformability test incorporating flat jacks

Flat jack testing is a versatile and effective technique that provides useful information on the mechanical properties of historical constructions. In-situ measurements of masonry modulus of elasticity should be performed in accordance with the ASTM C 1197 - 04 (ASTM 2004) in situ deformability test. Note:

Extensive studies have been conducted to confirm the reliability of this test, including the work by Noland J.L., Atkinson R.H., Schuller M.P. (1991), Gregorczyk and Lourenço (2000); Parivallal et al. (2011); and Simões (2012).



The in-situ deformability test is moderately destructive as it requires the removal of horizontal mortar joints (bed-joint) for the insertion of the two flat jacks (Figure 10A.2a). The horizontal slots are separated by at least five courses of brickwork, but the separation distance should not exceed 1.5 times the flat jack length. A pressure controlled hydraulic pump is used to inflate the flat jacks, applying vertical confinement pressure to the masonry between the two jacks. To monitor displacement, typically three measurement devices are attached between the two flat jacks (Figure 10A.2b). These flat jacks need to be calibrated, following ASTM C 1197 - 04 (ASTM 2004).

(a) Cutting mortar bed-joints and insertion of

flat jacks into clay brick masonry

(b) In-situ deformability test set-up under

preparation in clay brick masonry

Figure 10A.2: In situ deformability test preparation (EQ STRUC Ltd)

10A.2.3 Masonry flexural bond strength

Extract masonry prisms two bricks high and a single brick wide, and subject these to the flexural bond test of AS 3700-2001 (Australian Standards, 2001). Remove any rendering plaster from the sides of the sample before performing this test. Cut any samples that are two leafs thick or more into single leaf masonry prism samples. Alternatively, you may conduct the flexural bond test in situ if this is more practical.

(a) Plan view

Measurement device



(b) Elevation view

Figure 10A.3: Flexural bond test-set-up (AS 3700-2001)

10A.2.4 Masonry bed-joint shear strength

Conduct the ASTM C 1531-03 (ASTM 2003b) in-situ bed-joint shear test to determine masonry bed-joint properties. This type of test is moderately destructive as it requires the removal of at least one brick on one side of the test specimen to allow for insertion of a hydraulic jack, as well as the removal of a vertical mortar joint on the opposite side to allow horizontal bed joint movement to occur. The hydraulic jack is then loaded, using a pressure controlled hydraulic pump, until visible bed-joint sliding failure occurred. You can then derive the bed-joint shear strength from the peak pressure records. Alternatively, extract three brick high masonry prisms for laboratory testing following the triplet shear test BS EN 1052-3 (BSI 2002). This test should be conducted while applying axial compression loads of approximately 0.2 MPa, 0.4 MPa and 0.6 MPa. At least three masonry prism samples should be tested at each level of axial compression. Remove any rendering plaster from both sides of the sample before testing. Cut any masonry samples that are two leafs thick or more into single leaf samples. Bed-joint shear tests performed in the laboratory and in situ are shown in Figure 10A.4.

(a) Laboratory shear triplet test (b) In situ shear test without flat jacks (EQ

STRUC Ltd)

Figure 10A.4: In situ and laboratory bed- joint shear test



The in situ bed-joint shear test is limited to tests of the masonry face leaf. When the masonry unit is pushed in a direction parallel to the bed joint, shear resistance is provided across not only the bed-joint shear planes but also the collar joint shear plane. Because seismic shear is not transferred across the collar joint in a multi-leaf masonry wall, the estimated shear resistance of the collar joint must be deducted from the test values. This reduction is achieved by including a 0.75 reduction factor in Equation 10.33, which is the ratio of the areas of the top and bottom bed joints to the sum of the areas of the bed and collar joints for a typical clay masonry unit. The term P in Equation 10.33 represents the axial overburden acting on the bed joints. This value multiplied by the bed-joint coefficient of friction, (µf), allows estimation of the frictional component contributing to the recorded bed-joint stress. Due to the typical large variation of results obtained from individual bed-joint shear strength tests, the equation conservatively assumes µf = 1.0 for the purposes of determining cohesion, c. Therefore, for simplicity, the µf term has been omitted from the equation.

10A.3 Constituent Material Properties

10A.3.1 Brick compressive strength

Extract individual brick units for the ASTM C 67-03a (ASTM 2003a) half brick compression test. Cut these brick units into halves and cap them using gypsum plaster before compression testing (Figure 10A.5). Note that it is possible to obtain half brick units from the residual samples of the Modulus of Rupture test described in Section 10A.3.2.

Figure 10A.5: Brick and mortar sample and compression test set-up (EQ STRUC Ltd)

10A.3.2 Brick modulus of rupture

Extract individual brick units from the building and subject these to the modulus of rupture (MoR) test ASTM C 67-03a (ASTM 2003a). The tested brick specimens from the MoR test may be subjected to the half brick compression test ASTM C 67-03a (ASTM 2003a) in order to obtain a direct relationship between the brick MoR and compressive strength, f’b. Previous experimental investigation has confirmed that the brick unit MoR can be approximated to equal 0.12f’b.



10A.3.3 Mortar compressive strength

Extract irregular mortar samples for laboratory testing. As it is common for URM walls to have eroded mortar joints that were later repaired using stronger mortar, take care when selecting the location for mortar sample extraction to ensure that your samples are representative. The method to determine mortar compressive strength is detailed in ASTM C 109-08 (ASTM 2008). This method involves testing of 50 mm cube mortar samples, which generally are not attainable in existing buildings as most mortar joints are only 10 to 18 mm thick. Therefore, cut the irregular mortar samples into approximately cubical sizes with two parallel sides (top and bottom). The height of the mortar samples should exceed 15 mm in order to satisfactorily maintain the proportion between sample size and the maximum aggregate size. Cap the prepared samples using gypsum plaster to ensure a uniform stress distribution and testing in compression (Valek and Veiga, 2005): see Figure 10A.6 for examples. Measure the height to minimum lateral dimension (h/t) ratio of the mortar samples and use this to determine the mortar compressive strength correction factors. Divide the compression test result by the corresponding correction factors in Equation 10A.1. The average corrected strength is equal to the average mortar compressive strength, f’j.

…10A.1

where:

= normalised mortar compressive strength = t/l ratio correction factor = t/l ratio correction factor

= measured irregular mortar compressive strength. Equation 10A.1 normalises the measured compressive strength of irregular mortar samples to the compressive strength of a 50 mm cube mortar. Factors and are calculated as per Equations 10A.2 and 10A.3 (where . should be calculated as per Equation 10A.4) respectively. Factor is required in order to normalise the sample t/l ratio to 1.0, while factor is required in order to normalise the sample h/t ratio to 1.0, corresponding to a cubic mortar sample that is comparable to a 50 mm cube. These factors were derived based on the study detailed in Lumantarna (2012).

0.42 0.58 …10A.2

. …10A.3

. 2.4 5.7 4.3 …10A.4

When conducting tests on laboratory manufactured samples make 50 mm mortar cubes, leave these to cure under room temperature (±20 °C) for 28 days, and test them in compression following the mortar cube compression test ASTM C 109-08 (ASTM 2008).



(a) Example of typical extracted mortar samples

(b) Example of typical mortar sample preparations

(c) Example of typical test set-up

Figure 10A.6: Determination of mortar compression strength (EQ STRUC Limited)

10A.4 Proof Testing of Anchor Connections

An epoxied or grouted anchorage system is a typical method of connecting the floor and roof diaphragms of the building to masonry walls. Reliable anchor pull-out and shear strength is important for assessment or design of anchors and the specification of anchor spacing. Standard installation procedures of embedded anchors involve drilling the masonry wall, cleaning the drilled hole and epoxying or grouting threaded steel bars to the specified embedment depth, typically 50 mm less than the wall thickness. Two-part epoxy or high strength grouts are typically used with surface preparation conducted in accordance with the manufacturer’s specifications. On-site quality control and proof testing should be undertaken on at least 15% of all installed adhesive anchors, of which 5% should be tested prior to the installation of more than 20% of all anchors. Testing is required to confirm workmanship (particularly the mixing of epoxy and cleaning of holes) and anchor capacity against load requirements. If more than 10% of the tested anchors fail below a test load of 75% of the nominated probable capacity, discount the failed anchors from the total number of anchors tested as part of the quality assurance test. Test additional anchors to meet the 15% threshold requirements. Failures that cannot be attributed to workmanship issues are likely to be indicative of an overestimation of the available capacity and a reassessment of the available probable capacity is likely to be required.

10A.4.1 Anchors loaded in tension

Once the adhesive is cured (typically over 24 hours), the steel anchors can be loaded in tension using a hydraulic jack until ultimate carrying capacity is reached (ASTM, 2003) or when the load exceeds two times the specified load. The typical test set-up is shown in Figure 10A.7. A 600 mm clear span of reaction frame allows testing of up to 300 mm embedment depth without exerting any confining pressures onto the test area, as the reaction frame supports are outside the general zone of influence. On completing the test, the anchor stud is typically cut flush with the wall surface.



(a) Typical anchor pull test

set up

(b) Close up of the typical test set-up with an

alternative test frame

Figure 10A.7: Typical anchor pull-out test set-up (EQ STRUC Ltd)

10A.4.2 Anchors loaded in shear

The test set-up that could be adopted for in situ testing of anchors loaded in shear is shown in Figure 10A.8. Monotonic shear loading can be applied by using a single acting hydraulic actuator, with the external diameter of the actuator selected to be as small as possible. The bracket arrangements should minimise the tension loads in the anchors. The aim is to determine the shear capacity in the absence of tension.

(a) Typical anchor shear tests set-up (push cycle)

(b) Typical anchor shear tests set-up (pull cycle)

Figure 10A.8: Shear tests set-up used (EQ STRUC Ltd)

10A.5 Investigation of Collar Joints and Wall Cavities

Investigation of collar joints quality and wall cavities can be undertaken using a Ground Penetrating Radar (GPR) structural scanner (Figure 10A.9a). The scanner is capable of accurately determining the member thickness, metallic objects, voids and other

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Seismic Assessment of Unreinforced Masonry Buildings – Appendix 10A On-site Investigation

Appendix 10A–Detailed Assessment of Unreinforced Masonry Buildings App-10 Updated 22 April 2015 ISBN 978-0-473-26634-9

References ASTM (2003). “Standard Test Methods for Strength of Anchors in Concrete and Masonry Elements.” E488-96. ASTM International. Pennsylvania, United States.Se

ASTM. (2000). "Standard Test Method for Measurement of Masonry Flexural Bond Strength." C 1072 - 00a. ASTM International, Pennsylvania, United States.

ASTM. (2003a). “Standard Test Methods for Sampling and Testing Brick and Structural Clay Tile.”, C 67-03a. ASTM International, Pennsylvania, United States.

ASTM. (2003b). "Standard Test Methods for In Situ Measurement of Masonry Mortar Joint Shear Strength Index." C 1531 - 03. ASTM International, Pennsylvania, United States.

ASTM (2004). "Standard Test Method for In Situ Measurement of Masonry Deformability Properties Using the Flatjack Method." C 1197 - 04, ASTM International, Pennsylvania, United States.

ASTM. (2008). "Standard Test Method for Compressive Strength of Hydraulic Cement Mortars (Using 2-in. or [50-mm] Cube Specimens).”, C 109/C 109M - 08, ASTM International, Pennsylvania, United States.BSI (2002). "Methods of test for masonry. Determination of initial shear strength." BS EN 1052-3:2002, British Standards Institution, United Kingdom.

Noland J.L., Atkinson R.H., Schuller M.P. (1991) "A Review of the Flat Jack Method for Non-destructive Evaluation of Civil Structures and Materials.” The National Science Foundation. Grant No. MSM 9005818.

Gregorczyk P. and Lourenço P. (2000). “A Review on Flat-Jack Testing.” Universidad do Minho, Departamento de Engenharia Civil Azurém, P – 4800-058 Guimarães, Portugal.

Parivallal S., K. Kesavan, K. Ravisankar, B Arun Sundram and A K Farvaze Ahmed (2011). “Evaluation of In-situ Stress in Masonry Structures by Flat Jack Technique.” Proceedings of the National Seminar & Exhibition on Non-Destructive Evaluation. NDE 2011, December 8-10, 2011 CSIR-Structural Engineering Research Centre, Chennai 600 113.

Lumantarna, R. (2012). “Material Characterisation of New Zealand’s Clay Brick Unreinforced Masonry Buildings.” Doctoral Dissertation, University of Auckland, Department of Civil and Environmental Engineering Identifier: http://hdl.handle.net/2292/18879.

Simões A., A. Gago, M. Lopes & R. Bento (2012). “Characterization of Old Masonry Walls: Flat-Jack

Method.” 15th World Conference on Earthquake Engineering (15WCEE) Lisbon, Portugal 24-28 September 2012.

Standards Australia (2001). "Appendix D: Method of Test for Flexural Strength." AS 3700 - 2001, Standards Australia, Homebush, New South Wales, Australia.

Valek, J., and Veiga, R. (2005). ”Characterisation of Mechanical Properties of Historic Mortars - Testing of Irregular Samples”, Structural studies, repairs and maintenance of heritage architecture XI: Ninth international conference on structural studies, repairs and maintenance of heritage architecture, Malta, 22-24 June.

Seismic Assessment of Unreinforced Masonry Buildings – Appendix 10B Derivation of Instability Deflection and Fundamental Period for Masonry Buildings

Appendix 10B–Detailed Assessment of Unreinforced Masonry Buildings App-11 Updated 22 April 2015 ISBN 978-0-473-26634-9

Appendix 10B: Derivation of Instability Deflection and Fundamental Period for Face-Loaded Masonry Walls

10B.1 General considerations and approximations

There are many variations that need to be taken into account when considering a general formulation for URM walls that might fail out-of-plane. These include:

Walls will not usually be of a constant thickness in a building, or even within a storey.

Walls will have embellishments, appendages and ornamentation that may lead to eccentricity of masses with respect to supports.

Walls may have openings for windows or doors.

Support conditions will vary.

Existing buildings may be rather flexible, leading to possibly large inter-storey displacements that may adversely affect the performance of face-loaded walls.

You can use the following approximations to simplify your analysis while still accounting for some the key important factors. 1 Deformations due to distortions (straining) in the wall can be ignored. Assume

deflections to be entirely due to rigid body motion.

Note:

This is equivalent to saying that the change in potential energy from a disturbance of the wall from its initial position is mostly due to the movement of the masses of the elements comprising the wall and the movements of the masses tributary to the wall. Strain energy contributes less to the change in potential energy.

2 Assume that potential rocking occurs at the support lines (e.g. at roof or floor levels)

and, for walls that are supported at the top and bottom of a storey, at the mid-height. The mid-height rocking position divides the wall into two parts of equal height: a bottom part (subscript b) and a top part (subscript t). The masses of each part are not necessarily equal.

Note:

It is implicit within this assumption and (1) above that the two parts of the wall remain undistorted when the wall deflects. For walls constructed of softer mortars or walls with little vertical pre-stress from storeys above, this is not actually what occurs: the wall takes up a curved shape, particularly in the upper part. Nevertheless, errors occurring from the use of the stated assumptions have been found to be small and you will still obtain acceptably accurate results.



3 Assume the thickness to be small relative to the height of the wall. Assume the slope, A, of both halves of the wall to be small; in the sense that cos(A) ≈ 1 and sin(A) ≈ A.

Note:

The approximations for slope are likely to be sufficiently accurate for reasonably thin walls. For thick walls where the height to thickness ratio is smaller, the formulations in this appendix are likely to provide less accurate results and force-based approaches provide an alternative.

4 Inter-storey slopes due to deflection of the building are assumed to be small.

Note:

Approximate corrections for this effect are noted in the method. 5 In dynamic analyses, the moment of inertia is assumed constant and equal to that

applying when the wall is in its undisturbed position, whatever the axes of rotation.

Note:

The moment of inertia is dependent on the axes of rotation. During excitation, these axes continually change position. Assuming that the inertia is constant is reasonable within the context of the other approximations employed.

6 Damping is assumed at the default value in NZS 1170.5:2004, which is 5% of

critical.

Note:

For the aspect ratio of walls of interest, additional effective damping due to loss of energy on impact is small. Furthermore, it has been found that the surfaces at rocking (or hinge) lines tend to fold onto each other rather than experience the full impact that is theoretically possible, reducing the amount of equivalent damping that might be expected. However, for in-plane analysis of buildings constructed largely of URM, adopting a damping ratio that is significantly greater than 5% is appropriate.

7 Assume that all walls in storeys above and below the wall under study move “in

phase” with the subject wall.

Note:

Analytical studies have found this to be the case. One reason for this is that the effective stiffness of a wall as it moves close to its limit deflection (e.g. as measured by its period) becomes very low, affecting its resistance to further deflection caused by accelerations transmitted to the walls through the supports. This assumption means that upper walls, for example, will tend to restrain the subject wall by exerting restraining moments.



10B.2 Vertically spanning walls

10B.2.1 General formulation

Figures 10B.1 and 10B.2 show the configuration of a wall panel within a storey at two stages of deflection. The wall is intended to be quite general. Simplifications to the general solutions for walls that are simpler (e.g. of uniform thickness) are made in a later section. Figure 10B.1 shows the configuration at incipient rocking. Figure 10B.2 shows the configuration after significant rocking has occurred, with the wall having rotated through an angle A and with mid-height deflection, , where = Ah/2. In Figure 10B.1 the dimensions eb and et relate to the mass centroids of the upper and lower parts of the panel. The dimension ep relates to the position of the line of action of weights from upper storeys (walls, floors and roofs) relative to the centroid of the upper part of the panel. The arrows on the associated dimensioning lines indicate the positive direction of these dimensions for the assumed direction of motion (angle A at the bottom of the wall is positive in the anti-clockwise sense). Under some circumstances the signs of the eccentricities may be negative; for example for ep when an upper storey wall is much thinner than the upper storey wall represented here, particularly where the thickness steps on one face. When the lines of axial force from diaphragm and walls from above are different, the resultant force should be calculated.

Figure 10B.1: Configuration at incipient rocking

ICR

Wt

Wb

P

eb eo

eo+ebeP

et

yt

yb

h2

h

ICR



The instantaneous centres of rotation (ICR) are also marked on these figures. These are useful in deriving virtual work expressions.

10B.2.2 Limiting deflection for static instability

You can write the equation of equilibrium directly by referring to Figure 10B.2 and using virtual work expressions. For static conditions this is given by:

0 …10B.1

The final term represents the effect of any inter-storey drift. In the derivation presented, the total deformation has been assumed to be that resulting from the summation of the drift and the rocking wall. Writing:

…10B.2

and:

…10B.3

and collecting terms in A, the equation of equilibrium is rewritten as:

0 …10B.4

from which:

…10B.5

when the wall becomes unstable.



Figure 10B.2: Configuration when rotations have become significant and there is inter-

storey drift

Therefore, the critical value of the deflection at mid-height of the panel, at which the panel will be unstable, is:

∆ …10B.6

It is assumed that m, a fraction of this deflection, is the maximum useful deflection. Experimental and analytic studies indicate that this fraction might be assumed to be about 0.6. At larger displacements than 0.6i, analysis reveals an undue sensitivity to earthquake spectral content and a wide scatter in results.

10B.2.3 Equation of motion for free vibration

When conditions are not static, the virtual work expression on the left-hand side in the equation above is unchanged, but the zero on the right-hand side of the equation is replaced by mass x acceleration, in accordance with Newton’s law. This gives:

…10B.7

This uses the usual notation for acceleration (a double dot to denote the second derivative with respect to time; in this case indicating angular acceleration), and J as the rotational inertia. The rotational inertia can be written directly from Figures 10B.1 and 10B.2, noting that the centroids undergo accelerations vertically and horizontally as well as rotationally, and these accelerations relate to the angular acceleration in the same way as the displacements relate to the angular displacement. While the rotational inertia is dependent on the

support reaction

ICR

Wt

Wb

P

eb eo

eo+ebeP

et

yt

yb

h2

h

Wt Wt

Wb

support reaction

ICR

Wt

Wb

P

eb

eP

et

yt

yb

h2

h

Wt

eb - Ayb

eo +eb - Ah2

eo +eb +et +ep - Ah

ICRP

h +

A(e

0+

eb

+ e

t+

eP)

ICR

Wt

Wb

A, A

A, A

eo +eb +et - A(h -yt)

Drift (max = 0.025)

+

Rocking Wall



displacements, the effects of this variation are ignored. Therefore, the rotational inertia is taken as that when no displacement has occurred. This gives the following expression for rotational inertia.

…10B.8

where Jbo and Jto are the mass moments of inertia of the bottom and top parts respectively about their centroids, and Janc is the inertia of any ancillary masses, such as veneers, that are not integral with the wall but contribute to its inertia. For a wall with unit length, held at the top and bottom, and rocking crack at mid-height, with a density of ρ per unit volume, the mass moment of inertia about the horizontal axis through the centroid is given by:

kgm …10B.9

The corresponding mass moment of inertia about the vertical axis through the centroid is:

kgm …10B.10

The polar moment of inertia through the centroid is the sum of these, or:

kgm

…10B.11

where m is the mass (kg) and W (N) is the weight of the whole wall panel and g is the acceleration of gravity. Note that in this equation the expressions in square brackets are the squares of the radii from the instantaneous centres of rotation to the mass centroids, where the locations of the instantaneous centres of rotation are those when there is no displacement. Some CAD programs have functions that will assist in determining the inertia about an arbitrary point (or locus), such as about the ICR shown in Figure 10B.2. Collecting terms and normalising the equation so that the coefficient of the acceleration term is unity gives the following differential equation of free vibration:

…10B.12



10B.2.4 Period of free vibration

The solution of the equation for free vibration derived in the previous section is:

…10B.13

The time, , is taken as zero when the wall has its maximum rotation, A (=/2h). Using this condition and the condition that the rotational velocity is zero when the time = 0, the solution becomes:

∆ ….10B.14

Take the period of the “part”, Tp, as four times the duration for the wall to move from its position at maximum deflection to the vertical. Then the period is given by:

4 ∆ …10B.15

This can be simplified further by substituting the term for i found from the static analysis and putting the value of used for the calculation of period as Δt to give:

4 ∆∆

…10B.16

If we accept that the deflection ratio of interest is 0.6 (i.e. ∆ ∆ = 0.6), then this becomes:

6.27 …10B.17

as in the 2006 guidelines. However, research (Derakhshan et al, (2014a)) indicates that the resulting period and responding displacement demand is too large if a spectrum derived from linear elastic assumptions is used. Rather, this research suggests that an effective period calculated from an assumed displacement of 60% of the assumed displacement capacity should be used. Therefore, the period is based on Δt = 0.36Δi so that:

4.07 …10B.18

10B.2.5 Maximum acceleration

The acceleration required to start rocking of the wall occurs when the wall is in its initial (undisturbed) state. This can be determined from the virtual work equations by assuming that A=0. Accordingly:

…10B.19



However, a more cautious appraisal assumes that the acceleration is influenced primarily by the instantaneous acceleration of the supports, transmitted to the wall masses, without relief by wall rocking. Accordingly:

…10B.20

where Cm is the acceleration coefficient to just initiate rocking.

10B.2.6 Participation factor

The participation factor can be determined in the usual way by normalising the original form of the differential equation for free vibration, modified by adding the ground acceleration term. For the original form of the equation, the ground acceleration term is added to the RHS. Written in terms of a unit rotation, this term is (Wbyb + Wtyt) times the ground acceleration. The equation is normalised by dividing through by J, and then multiplied by h/2 to convert it to one involving displacement instead of rotation. The participation factor is then the coefficient of the ground acceleration. That is:

…10B.21

10B.2.7 Simplifications for regular walls

You can make simplifications where the thickness of a wall within a storey is constant, there are no openings, and there are no ancillary masses. Further approximations can then be applied:

The weight of each part (top and bottom) is half the total weight, W.

yb = yt = h/4

The moment of inertia of the whole wall is further approximated by assuming that all e are very small relative to the height (or, for the same result, by ignoring the shift of the ICR from the mid-line of the wall), giving J = Wh2/12g. Alternatively, use the simplified expressions for J given in Table 10B.1.

10B.2.7.1 Approximate displacements for static instability

Table 10B.1 gives values for a and b and the resulting mid-height deflection to cause static instability when eb and/or ep are either zero or half of the effective thickness of the wall, t. In this table eo and et are both assumed equal half the effective wall thickness. While these values of the eccentricities are reasonably common, they are not the only values that will occur in practice. The effective thickness may be assumed as follows:

0.975 0.025 …10B.22

where tnom is the nominal thickness of the wall.



Experiments show that this is a reasonable approximation, even for walls with soft mortar. In that case, there is greater damping and that reduces response, which compensates for errors in the expression for effective thickness. 10B.2.7.2 Approximate expression for period of vibration Noting that:

…10B.23

and using the approximation for J relevant to a wall with large aspect ratio, the expression for the period is given by:

4.07 …10B.24

where it should be noted that the period is independent of the restraint conditions at the top and bottom of the wall (i.e. independent of both eb and ep). If the height is expressed in metres, this expression simplifies to:

.

/ …10B.25

It should be appreciated that periods may be rather long. This approximation errs on the low side, which leads to an underestimate of displacement demand and therefore to slightly incautious results. The fuller formulation is therefore preferred. 10B.2.7.3 Participation factor Suitable approximations can be made for the participation factor. This could be taken at the maximum value of 1.5. Alternatively, the numerator can be simplified as provided in the following expression, and the simplified value of J shown in Table 10B.1 can be used. 10B.2.7.4 Maximum acceleration By making the same simplifications as above, the maximum acceleration is given by:

…10B.26

Or, more cautiously, the acceleration coefficient, Cm, is given in Table 10B.1 for the common cases regularly encountered.

Appendix 10B–Updated 22 April 20

10B.2.8

Using the h/4, b is percentage31% for Caffects shosteel mome Table 10B.1

Boundary condition nu

ep

eb

b

a

i = bh/(2

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Note: 1. The boun2. The top e

et, is the ha boundar

3. The eccenshould be

10B.3

10B.3.1

Figure 10Boverburdenload does horizontallextent thanthe wall ina continuouapplication

–Detailed Asses015

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App-0-473-26634

lacemen

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SBN 978-0-4

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– Appendix 10Basonry Buildings

App-21473-26634-9

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r it are not

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…10B.27

…10B.28

…10B.29

10B.3.

B s

9

s y t

e n



10B.3.2 Limiting deflection for static instability

When the wall just becomes unstable, the relationship for A remains the same as before but the deflection is Ah. Thus, the limiting deflection is given by:

∆ ...10B.30

For the case where P=0 and yb=h/2 this reduces to i = 2eb = t.

10B.3.3 Period of vibration

If Δt = 0.36Δi as for the simple case, the general expression for period would remain valid. However, cantilevers are much more susceptible to instability under real earthquake stimulation than wall panels that are supported both top and bottom. Therefore, the maximum useable displacement for calculation of capacity, Δm, is reduced from 0.6Δi to 0.3Δi and the displacement for calculation of period changes from 0.6Δm to 0.8Δm = 0.24Δi

so that:

3.1 …10B.31

Where P=0, eb=t/2, yb=h/2, approximating t=tnom and expressing h in metres, the period of vibration is given by:

0.65 1 …10B.32

Note that P, whether eccentric or not, will not affect the static instability displacement, and therefore neither the displacement demand (by affecting the period), nor the displacement capacity.

10B.3.4 Participation factor

The expression for the participation factor remains unaffected; that is, = Wh2/2J. This may be simplified for uniform walls with P=0 (no added load at the top) by inserting the specific expression for J. This gives:

…10B.33

10B.3.5 Maximum acceleration

Using the same simplifications as above:

…10B.34

Seismic Assessment of Unreinforced Masonry Buildings – Appendix 10C Charts for Assessment of Out-of-Plane Walls

Appendix 10C– Charts for Assessment of Out-of-Plane Walls App-23 Updated 22 April 2015 ISBN 978-0-473-26634-9

Appendix 10C: Charts for Assessment of Out-of-Plane Walls

10C.1 General

This appendix presents simplified ready-to-use charts for estimation of %NBS for face-loaded unreinforced masonry walls with uniform thickness. The charts have been developed for walls with various slenderness ratios (wall height/thickness) vs Basic Performance Ratio (BPR). The BPR can be converted to %NBS after dividing it by the product of the appropriate spectral shape factor (Ch(0), required to evaluate C(0) for parts), return period factor (R), hazard factor (Z), near-fault factor (N(T, D)), and part risk factor (Rp) which have been assigned unit values for developing the charts. The charts are presented for various boundary conditions and ratio of load on the wall to self-weight of the wall. Refer to Section 10 and Appendix 10B for symbols and sign conventions. This appendix includes charts for the following cases:

one-way vertically spanning walls laterally supported both at the bottom and the top with no inter-storey drift

one-way vertically spanning walls laterally supported at the top and the bottom with inter-storey drift of 0.025

vertical cantilever walls. The following section presents how these charts should be used.

10C.2 One-way Vertically Spanning Face-Loaded Walls

Charts for one-way vertically spanning walls are presented in Figures 10C.1a-f, 10C.2a-f and 10C.3a-f for 110 mm, 230 mm and 350 mm thick walls respectively for inter-storey drift of 0.00. Similarly, charts for an inter-storey drift of 0.025 are presented in Figures 10C.4a-f, 10C.5a-f and 10C.6a-f for 110 mm, 230 mm and 350 mm thick walls respectively. The charts have been developed for et = eo = t/2 and various values for ep. Follow the following steps for estimation of %NBS for a vertically spanning face-loaded wall:

Identify thickness, tGross and height, h of the wall.

Calculate slenderness ratio of the wall (h/tGross).

Calculate the total self-weight, W of the wall.

Calculate vertical load, P on the wall. This should include all the dead load and appropriate live loads on the wall from above.

Calculate P/W.

Calculate eccentricities (eb and ep). eb could be t/2 or 0, whereas ep could be ±t/2 or 0. To assign appropriate values, check the base boundary condition and location of P on the wall. Calculation of effective thickness, t is not required.

Seismic Assessment of Unreinforced Masonry Buildings – Appendix 10C Charts for Assessment of Out-of-Plane Walls

Appendix 10C– Charts for Assessment of Out-of-Plane Walls App-24 Updated 22 April 2015 ISBN 978-0-473-26634-9

Refer to the appropriate charts (for appropriate eb and ep, P/W and inter-storey drift).

Estimate Basic Performance Ratio (BPR) from the charts. Linear interpolation between plots may be used as necessary for inter-storey drifts between 0 and 0.025.

Refer NZS 1170.5 for Ch(0) required to evaluate C(0) for parts, R, Z, N(T, D), CHi and Rp. For estimation of CHi, hi is height of the mid-height of the wall from the ground.

% /

,

10C.3 Vertical Cantilevers

Charts for one-way vertically spanning walls are presented in Figures 10C.7a-c, 10C.8a-c and 10C.9a-c for 110 mm, 230 mm and 350 mm thick walls respectively. Follow the following steps for estimation of %NBS of a face-loaded cantilever wall:

Identify thickness, tGross and height, h of the wall.

Calculate slenderness ratio of the wall (h/tGross).

Calculate total self-weight, W of the wall above the level of cantilevering plane.

Calculate vertical load, P on the wall, if any. This should include all the dead load and appropriate live loads on the wall from above.

Calculate P/W.

Calculate eccentricity, ep, for loading P(ep). ep could be ±t/2 or 0, which depends upon location of P on the wall. Calculation of effective thickness, t is not required.

Refer to the appropriate charts (for appropriate ep, and P/W).

Estimate Basic Performance Ratio (BPR) from the charts. Interpolation between plots may be used as necessary.

Refer NZS 1170.5 for Ch(0) required to evaluate C(0) for parts, R, Z, N(T, D), CHi and Rp. For estimation of CHi, hi shall be taken as height of the base of the cantilever wall.

% /

,

ApUpd

ppendix 10C– Cdated 22 April 2015

Charts for Asses

a

sment of Out-of

a) For eb = +

b) For eb =

Seismic Ass

f-Plane Walls

+ t/2 and ep =

+ t/2 and ep

sessment of Unr

= + t/2 (tGross

p = 0 (tGross =

reinforced MasoCharts for Asse

IS

=110 mm)

110 mm)

onry Buildings –sessment of Out

SBN 978-0-4

– Appendix 10Ct-of-Plane Walls

App-25473-26634-9

C s

5 9

Appendix 10C–Updated 22 April 20

– Charts for Ass015

sessment of Ou

c) For eb =

d) For e

Seismic A

ut-of-Plane Wall

= + t/2 and e

eb = 0 and ep

Assessment of

s

ep = -t/2 (tGro

p = t/2 (tGross

Unreinforced MCharts for A

oss = 110 mm

= 110 mm)

Masonry BuildingAssessment of O

ISBN 978-

m)

gs – Appendix 1Out-of-Plane W

App-0-473-26634

10C Walls

p-26 4-9

ApUpd


Figure 10

Charts for Asses

C.1: 110

sment of Out-of

e) For eb =

f) For eb =

0 mm thick o

Seismic Ass

f-Plane Walls

= 0 and ep =

0 and ep = -

one-way ve

sessment of Unr

= 0 (tGross = 1

-t/2 (tGross =

rtically span


IS

10 mm)

110 mm)

nning face-l


SBN 978-0-4

oaded walls


App-27473-26634-9

s ( = 0)

C s

7 9



sessment of Ou

a) For eb =

b) For eb

Seismic A

ut-of-Plane Wall

= + t/2 and e

b = + t/2 and

Assessment of

s

ep = + t/2 (tGr

d ep = 0 (tGros


ross =230 mm

ss =230 mm)


ISBN 978-

m)


App-0-473-26634

10C Walls

p-28 4-9

ApUpd


Charts for Asses

c

sment of Out-of

c) For eb = +

d) For eb =

Seismic Ass

f-Plane Walls

+ t/2 and ep =

= 0 and ep =

sessment of Unr

= -t/2 (tGross =

t/2 (tGross = 2


IS

= 230 mm)

230 mm)


SBN 978-0-4


App-29473-26634-9

C s

9 9


Figure 1


10C.2: 2

sessment of Ou

e) For e

f) For eb

230 mm thic

Seismic A

ut-of-Plane Wall

eb = 0 and e

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ck one-way v

Assessment of

s

ep = 0 (tGross =

= -t/2 (tGross

vertically sp


= 230 mm)

= 230 mm)

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ISBN 978-

e-loaded wa


App-0-473-26634

alls ( = 0 )

10C Walls

p-30 4-9

ApUpd


Charts for Asses

a

sment of Out-of

a) For eb = +

b) For eb =

Seismic Ass

f-Plane Walls

t/2 and ep =

+ t/2 and ep

sessment of Unr

= + t/2 (tGross

p = 0 (tGross =


IS

= 350 mm)

350 mm)


SBN 978-0-4


App-31473-26634-9

C s

9



sessment of Ou

c) For eb =

d) For e

Seismic A

ut-of-Plane Wall

= + t/2 and e

eb = 0 and ep

Assessment of

s

ep = -t/2 (tGro

p = t/2 (tGross


oss = 350 mm

= 350 mm)


ISBN 978-

m)


App-0-473-26634

10C Walls

p-32 4-9

ApUpd


Figure 10C

Charts for Asses

C.3: 350

sment of Out-of

e) For eb =

f) For eb =

0 mm thick o

Seismic Ass

f-Plane Walls

= 0 and ep =

0 and ep = -

one-way ver

sessment of Unr

= 0 (tGross = 3

-t/2 (tGross = 3

rtically span


IS

350 mm)

350 mm)

nning face-lo


SBN 978-0-4

oaded walls


App-33473-26634-9

s ( = 0 )

C s

3 9



sessment of Ou

a) For e

b) For

Seismic A

ut-of-Plane Wall

eb = + t/2 and

r eb = + t/2 an

Assessment of

s

d ep = + t/2 (

nd ep = 0 (tG


tGross = 110

Gross = 110 m


ISBN 978-

mm)

m)


App-0-473-26634

10C Walls

p-34 4-9

ApUpd


Charts for Assessment of Out-of

c) For eb =

d) For e

Seismic Ass

f-Plane Walls

= + t/2 and e

eb = 0 and ep

sessment of Unr

ep = -t/2 (tGro

p = t/2 (tGross


IS

oss = 110 mm

= 110 mm)


SBN 978-0-4

m)


App-35473-26634-9

C s

5 9


Figure 10


C.4: 110

sessment of Ou

e) Fo

f) For

0 mm thick o

Seismic A

ut-of-Plane Wall

or eb = 0 and

r eb = 0 and

one-way ver

Assessment of

s

d ep = 0 (tGro

ep = -t/2 (tGr

rtically span


oss = 110 mm

ross = 110 mm

nning face-l


ISBN 978-

m)

m)

oaded walls


App-0-473-26634

s ( = 0.025

10C Walls

p-36 4-9

)

ApUpd



a) For eb =

b) For eb

Seismic Ass

f-Plane Walls

= + t/2 and e

b = + t/2 and

sessment of Unr

ep = + t/2 (tGr

d ep = 0 (tGros


IS

ross =230 mm

ss =230 mm)


SBN 978-0-4

m)

)


App-37473-26634-9

C s

7 9



sessment of Ou

c) For e

d) Fo

Seismic A

ut-of-Plane Wall

eb = + t/2 an

or eb = 0 and

Assessment of

s

d ep = -t/2 (t

d ep = t/2 (tGro


tGross = 230 m

oss = 230 mm


ISBN 978-

mm)

m)


App-0-473-26634

10C Walls

p-38 4-9

ApUpd

F


Figure 10C.5

Charts for Asses

5: 230 m

sment of Out-of

e) For e

f) For e

mm thick on

Seismic Ass

f-Plane Walls

eb = 0 and e

eb = 0 and ep

e-way vertic

sessment of Unr

ep = 0 (tGross =

p = -t/2 (tGross

cally spanni


IS

= 230 mm)

s = 230 mm)

ing face-loa


SBN 978-0-4

aded walls (


App-39473-26634-9

= 0.025 )

C s

9 9



sessment of Ou

a) For e

b) For

Seismic A

ut-of-Plane Wall

eb = + t/2 and

r eb = + t/2 an

Assessment of

s

d ep = + t/2 (

nd ep = 0 (tG


tGross = 350

Gross = 350 m


ISBN 978-

mm)

m)


App-0-473-26634

10C Walls

p-40 4-9

ApUpd



c) For eb =

d) For e

Seismic Ass

f-Plane Walls

= + t/2 and e

eb = 0 and ep

sessment of Unr

ep = -t/2 (tGro

p = t/2 (tGross


IS

oss = 350 mm

= 350 mm)


SBN 978-0-4

m)


App-41473-26634-9

C s

9


Figure 10


C.6: 350

sessment of Ou

e) Fo

f) For

0 mm thick o

Seismic A

ut-of-Plane Wall

or eb = 0 and

r eb = 0 and

one-way ver

Assessment of

s

d ep = 0 (tGro

ep = -t/2 (tGr

rtically span


oss = 350 mm

ross = 350 mm

nning face-l


ISBN 978-

m)

m)

oaded walls


App-0-473-26634

s ( = 0.025

10C Walls

p-42 4-9

)

ApUpd


Charts for Asses

a

sment of Out-of

a) For eb = +

b) For eb =

Seismic Ass

f-Plane Walls

t/2 and ep =

+ t/2 and ep

sessment of Unr

= + t/2 (tGross

p = 0 (tGross =


IS

= 110 mm)

110 mm)


SBN 978-0-4


App-43473-26634-9

C s

3 9



sessment of Ou

c) For eb =

Figure 10C

a) For eb =

Seismic A

ut-of-Plane Wall

= + t/2 and e

C.7: 110

= + t/2 and e

Assessment of

s

ep = -t/2 (tGro

mm thick c

ep = + t/2 (tGro


oss = 110 mm

cantilever wa

oss = 230 mm


ISBN 978-

m)

all

m)


App-0-473-26634

10C Walls

p-44 4-9

ApUpd


Charts for Asses

c

F

sment of Out-of

b) For eb =

c) For eb = +

igure 10C.8

Seismic Ass

f-Plane Walls

+ t/2 and ep

+ t/2 and ep =

8: 230 m

sessment of Unr

p = 0 (tGross =

= -t/2 (tGross =

m thick can


IS

230 mm)

= 230 mm)

ntilever wall


SBN 978-0-4


App-45473-26634-9

C s

5 9



sessment of Ou

a) For eb =

b) For eb

Seismic A

ut-of-Plane Wall

= + t/2 and e

b = + t/2 and

Assessment of

s

ep = + t/2 (tGro

ep = 0 (tGross


oss = 350 mm

s = 350 mm)


ISBN 978-

m)

)


App-0-473-26634

10C Walls

p-46 4-9

ApUpd


Charts for Asses

c

F

sment of Out-of

c) For eb = +

igure 10C.9

Seismic Ass

f-Plane Walls

+ t/2 and ep =

9: 350 m

sessment of Unr

= -t/2 (tGross =

m thick can


IS

= 350 mm)

ntilever wall


SBN 978-0-4


App-47473-26634-9

C s

7 9

New Zealand SocietyFor Earthquake

Engineering

Assessment and Improvementof the Structural Performance

of Buildings in Earthquakes

Section 14 New SectionGeotechnical Considerations

Recommendations of a NZSEE Project Technical GroupIn collaboration with SESOC and NZGS

Supported by MBIE and EQCJune 2006

Issued as part of Corrigendum No. 4ISBN 978-0-473-32280-9 (pdf version)

This document is a new section which now forms part of an amendment to the NZSEE Guidelines which were published in 2006. Any comments will be gratefully received. Please forward any comments to NZSEE Executive Officer at [email protected] April 2015

SeUpd

S

ection 14 - Geotdated 22 April 2015

Section

14.1 Int

14.2 Sc

14.3 Th

14.4 Ro

14.

14.

14.

14.

14.5 As

14.6 Ma

14.7 Ge

14.8 As

14.9 Ide

14.10 So

14.11 Re

14.12 Su

technical Consid

14 - Ge

roduction ..

ope .............

e Holistic E

oles and Res

.4.1 Gene

.4.2 The S

.4.3 The G

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sessment P

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ations (24 March

SBN 978-0-4

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14-i473-26634-9

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....... 14-2

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Geotechnical Considerations (24 March 2015 revision)

Section 14 - Geotechnical Considerations 14-13


Allowing for soil-foundation-structure interaction can be very complicated and difficult to model. Precision should not be assumed in any assessment of the interaction, but the sensitivity to the expected response of the various assumptions should be understood. The process will require close collaboration between the structural engineer and the geotechnical engineer with each having an understanding of the issues faced by the other. For assessments of earthquake performance of buildings both the structural and geotechnical engineer must recognise and accommodate the potential for non-linear behaviour of the structure, foundations and the ground. Principles to work by include:

The ground’s behaviour cannot be represented by unique parameter values with uniform distributions (e.g. linear springs).

With close collaboration, the old fear of possible misinterpretations and abuse of numbers (e.g. spring stiffness, modulus of subgrade reaction) can be significantly reduced and possibly averted.

An iterative process between structural and geotechnical designers has to be established as soil behaviour is non-linear, spring stiffness depends on load, and load depends on structure (including foundation) stiffness.

Sensitivity to variations in assumptions should always be checked. There can be some beneficial influence of soil-foundation structure interaction on a building’s life-safety performance (e.g. elongation of building period, concentration of displacement demands in ‘ductile’ foundation rotation, damping resulting from plastic soil behaviour etc.). However, these beneficial influences are the subject of on-going research. The assessing engineer should be cautioned in adopting the various ‘benefits’ of SFSI if considering possible mechanisms that may significantly reduce the assumed seismic demands on the structure. The following are considerations for SFSI modelling:

Soil structure interaction is modelled directly by soil springs, because the structural model needs to be supported on something:

- The behaviour of springs is predictable and easy to understand.

- Springs are easy to incorporate into the software most structural engineers use.

- In a lot of cases structure response is not that sensitive to the spring values used (sensitivity test – 50% to 200% x spring value). If insensitivity is confirmed, this in itself is a useful finding.

Pinned or fixed supports are not necessarily realistic.

Load transfer and shearing depends on relative stiffness of both structural elements and supporting ground.

Multiple load cases to be considered (permanent, temporary, dynamic, different combinations, load factors etc.).

Serviceability deflections are often critical for the design of new structures but not for the assessment of existing structures. Therefore, bearing capacities capped to limit settlements to meet serviceability conditions are not appropriate for assessments of structures for earthquake life-safety protection.

Cost and time associated with more rigorous analysis methods.


14.11

ASCE 41 (20site characteand foundatiNew Zealand

Bradley (201Wellington, M

Clayton, Kamof existing bu

Larkin & Van

NZS 1170.5

14.12

Boulanger & 14/01, CenteCalifornia, Da

Bray & Das12:1129-1156

Day (1996) G

Kramer (2002

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NZGS Guide

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m & Beer (201uildings. 2014

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ww.nzgs.org/p

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2008). Soil liqtitute, Oakland

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ZSEE Bull. Vo

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index.aspx

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des guidance ic earth presscompatible w

2015 conferen

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ol. 47, No. 1.

No. UCD/CGng, University

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assessment a

undation desi

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